Offshore Foundations - PSA

OFFSHORE FOUNDATIONS: TECHNOLOGIES, DESIGN AND APPLICATION Pedro Gomes Simões de Abreu Thesis to obtain the Master of Science Degree in Civil Engineering

Masters in Civil Engineering Supervisor: Dr. Peter Bourne-Webb

Examination Committee: Chairperson: Prof. Jaime Santos Supervisor: Dr. Peter Bourne-Webb Members of the Committee: Prof. Alexandre Pinto

JULY 2014

Offshore Foundations: Technologies, Design and Application

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ACKNOWLEDGMENTS First and foremost I would like to take this t his opportunity to express my sincere gratitude to Dr Peter Bourne-Webb for encouraging me to pursue this research. Dr Peter was always available to discuss various aspects of the work and remained an important source of guidance throughout this project. It would be remiss of me not to thank my dearest colleague Ana Tavares. Without her I would not accomplish this entire process, so for that, thank you for your support during these last 5 years. A special thank you also to my family, particularly my sister and my parents, your guidance, encouragement, love and understanding, not only over the past year but throughout my life has been an inspiration. I would like to thank my grandfather Antonio for his effort over the last 70 years to help all my family to pursue p ursue our studies as far as we all wanted. Finally, all my friends that faced by my side my academic path throughout these last seven years, thank you for your unwavering u nwavering support has been a source of strength and encouragement. This work is dedicated to my parents (Virgílio and Lurdes) and my aunt Ana. Pedro G. Simões de Abreu

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ABSTRACT The offshore oil industry started over 60 years ago, since then it evolved immensely. This evolution was forced by the need of exploiting oil and gas reserves in more challenging regions. The purpose of this study was to gather information about the foundation structures used in the offshore industry, and to assess the applicability of two types of foundation in a real scenario. São Tome & Principe (STP) was selected as the case-study for this paper because it is a member of the Community of Portuguese Language Countries, and has recently been subjected to several studies in its offshore region to evaluate its potential as an oil & gas supplier. This paper described the geotechnical characterisation of the offshore of STP based on investigations performed in the Gulf of Guinea (GoG) for more than 10 years. The results of the study were that the soil is probably a highly sensitive clay (St=2 to 4), and the shear strength profile presents a gradient of about 1.5 kPa/m. Another conclusion is that many sites in the GoG exhibit a greater resistance (up to about 15 kPa) in the first 2 m, this phenomenon is called a “crust”. This work also describes design principles for two anchoring systems: the Torpedo Anchors and Suctions Embedded Plate Anchors (SEPLA). For Torpedo, the results revealed that the pull-out resistance, after reconsolidation, is expected to be 8.7 MN. Whereas, the results for SEPLA holding capacity is expected to be 10 MN. For both systems the calculations were made for the largest of their solutions available in the market. Keywords: Offshore Foundations, ultra-deep water, Torpedo Anchors, SEPLA, São Tomé & Princípe.

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TABLE OF CONTENTS Acknowledgments ..................................................................................................................................................... ii Abstract.......................................................................................................................................................................... iv List of Figures.............................................................................................................................................................. ix List of Tables ..............................................................................................................................................................xii List of Symbols .........................................................................................................................................................xiv 1.

2.

Introduction & Motivation............................................................................................................................1 1.1.

Context & Motivation ...............................................................................................................................................1

1.2.

Thesis structure .......................................................................................................................................................... 4

Offshore production facilities......................................................................................................................5 2.1. Introduction ........................................................................................................................................................................ 5 2.2.

Fixed Platforms ........................................................................................................................................................... 5

2.3.

Compliant Tower........................................................................................................................................................ 7

2.4.

Tension Leg Platform ...............................................................................................................................................8

2.5.

Semi-Submersible Floating Production Systems ........................................................................................ 8

....................................................................................................................................................................................................... 9

3.

2.6.

SPAR Platform .......................................................................................................................................................... 10

2.7.

Floating Production Storage and Offloading Facility ............................................................................. 11

2.8.

Subsea System .......................................................................................................................................................... 13

Shallow Water Foundations......................................................................................................................14 3.1.

Spudcans ..................................................................................................................................................................... 14

3.1.1. 3.2.

4.

Geotechnical Calculations ......................................................................................................................... 16

Pile Foundations ...................................................................................................................................................... 16

3.2.1.

Driven Piles ...................................................................................................................................................... 17

3.2.2.

Grouted Piles ................................................................................................................................................... 19

3.2.3.

Pile Resistance ................................................................................................................................................ 20

3.3.

Gravity Base Structures........................................................................................................................................ 22

3.4.

Concrete Caissons for Tension Leg Platforms ........................................................................................... 25

3.5.

Steel Buckets for Jackets ...................................................................................................................................... 27

3.5.1.

Installation in Clay........................................................................................................................................28

3.5.2.

Installation in Sand....................................................................................................................................... 28

Deep and Ultra-Deep Water Foundations ..........................................................................................28 4.1.

Gravity Anchors ....................................................................................................................................................... 30

4.2.

Pile Anchors ............................................................................................................................................................... 31 vi


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4.3.

Suction Caissons ...................................................................................................................................................... 32

4.4.

Vertically Loaded Drag Anchor ........................................................................................................................ 34

4.5.

Suction Embedded Plate Anchor ..................................................................................................................... 37

4.6.

Dynamically Penetrated Anchor...................................................................................................................... 38

Geological Characterization ......................................................................................................................43 5.1.

Topographical features of ocean floors ........................................................................................................ 44

5.2.

Seabed Geology ........................................................................................................................................................ 47

5.2.1.

Seabed sediments origin and classification...................................................................................... 48

5.2.2.

Geotechnical characteristics of some offshore regions .............................................................. 52

5.3.

6.

5.3.1.

Triggering Mechanisms for Submarine Slope Failures ............................................................... 55

5.3.2.

Geohazard identification ...........................................................................................................................56

5.3.3.

Geotechnical site investigation............................................................................................................... 59

5.3.4.

Submarine slope failures and slides .................................................................................................... 60

Case study: Geotechnical considerations for ultra-deep oil fields off São Tome e Principe 63 6.1.

Proposed development ........................................................................................................................................ 63

6.2.

Geotechnical Site Conditions ............................................................................................................................. 63

6.2.1.

Index Properties of Gog Sediments ...................................................................................................... 66

6.2.2.

In Situ Stresses and Stress History ....................................................................................................... 69

6.2.3.

Shear Strength Profiles............................................................................................................................... 72

6.3.

Design of anchor solutions ................................................................................................................................. 77

6.3.1.

Torpedo Anchors........................................................................................................................................... 77

6.3.2.

SEPLA .................................................................................................................................................................. 90

6.4. 7.

Geohazards ................................................................................................................................................................. 53

Discussion of Results.......................................................................................................................................... 106

Conclusion and Further Research ....................................................................................................... 109 7.1 Conclusion....................................................................................................................................................................... 109 7.2. Further Research........................................................................................................................................................ 110 Index .......................................................................................................................................................................................... 121

Appendix I – Details of Shallow Foundation Studies. ..........................................................................123 Appendix II – Calculation to determine Loss in Anchor Embedment. ......................................... 125 Appendix III – Calculations to determine the resistance of the SEPLA. ......................................127 Wilde et al. (2001) .............................................................................................................................................................. 127 Merifield et al. (2001) ....................................................................................................................................................... 127 DNV-RP-E302 (2002)........................................................................................................................................................ 129

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LIST OF FIGURES FIGURE 1 – TREND IN WATER DEPTH FOR EXPLORATION AND DEVELOPMENT DRILLING SINCE 1940. ................................................. 1 FIGURE 2 – MARINE FOUNDATIONS TYPE BY APPLICATION AND WATER DEPTH . FPSO- FLOATING PRODUCTION, STORAGE AND OFFLOADING VESSEL. VLA- DRAG EMBEDMENT VERTICALLY LOADED ANCHOR . SEPLA- SUCTION EMBEDDED PLATE ANCHOR . DPA- DYNAMIC PENETRATING ANCHOR. .......................................................................................................................................... 2 FIGURE 3 – MAP OF SÃO TOMÉ & PRINCIPE ISLANDS AND THE ECONOMIC EXCLUSIVE Z ONE, COURTESY OF AGENCIA NACIONAL DE PETRÓLEO DE STP. ............................................................................................................................................... 3 FIGURE 4 – OFFSHORE DEVELOPMENTS SYSTEMS : BOTTOM SUPPORTED, VERTICALLY MOORED STRUCTURES, FLOATING PRODUCTION AND SUBSEA SYSTEMS, FROM: HTTP ://I46.PHOTOBUCKET.COM/ALBUMS/F133/TAMAAA/UPLOAD1/PICTURE1.PNG. ....................... 5 FIGURE 5 – STEEL JACKET PLATFORMS AND CONCRETE GRAV ITY STRUCTURE , FROM: HTTP://PETROAHDAL.WEBS.COM/APPS/PHOTOS/PHOTO?PHOTOID=109394836 AND HTTP :// OFFSHOREMAG.COM/ARTICLES/2013/JAN/GO-AHEAD-FOR-14-BILLION-HEBRON-PROJECT-OFFSHORE -EASTERN-CANADA.HTML. ................ 6 FIGURE 6 – OFFSHORE PLATFORM ELEMENTS , FROM: HTTP://WWW.CONSERVATION.CA.GOV/DOG/PICTURE _A _WELL/PAGES/OFFSHORE _PLATFORM.ASPX. .................................... 6 FIGURE 7 – TRANSPORTATION OF THE BULLWINKLE STEEL SUBSTRUCTURE (JACKET), COURTESY OF SHELL INTL. HTTP://WWW.ESA.ORG/ESABLOG/WP-CONTENT/UPLOADS/2011/10/STEEL -JACKET-BEING-TOWED-OFFSHORE.JPG. ................. 7 FIGURE 8 – PETRONIUS COMPLIANT TOWER AND EMPIRE STATE BUILDING FOR COMPARISON, FROM: HTTP://PETROWIKI.ORG/FILE%3AVOL3_PAGE _530_IMAGE _0001.PNG. ....................................................................... 7 FIGURE 9 – WORLDWIDE FLEET OF INSTALLED AND SANCTIONED TLPS, COURTESY OF BRITISH PETROLEUM FROM: HTTP://PETROWIKI.ORG/FILE%3AVOL3_PAGE _532_IMAGE _0001.PNG. ....................................................................... 8 FIGURE 10 – SEMI-SUBMERSIBLE VESSEL WITH TWIN HULLS (COLUMNS AND PONTOONS). HTTP://OILANDGASPROCESSING.BLOGSPOT.PT /2009/02/OIL-RIG-OFFHORE-STRUCTURE.HTML ............................................. 9 FIGURE 11 - WORLDWIDE FLEET OF INSTALLED AND SANCTIONED SEMI -SUBMERSIBLE FPS, COURTESY OF BRITISH PETROLEUM FROM: HTTP://PETROWIKI.ORG/FILE%3AVOL3_PAGE _531_IMAGE _0002.PNG. ....................................................................... 9 FIGURE 12 – THREE DIFFERENT SPAR PROFILES, FROM: HTTP://IMAGES.PENNWELLNET.COM/OGJ/IMAGES/OGJ2/9644JSK02.GIF. ..... 10 FIGURE 13 – WORLDWIDE FLEET OF INSTALLED SPARS, COURTESY OF BRITISH PETROLEUM. ........................................................ 11 FIGURE 14 - FLOATING PRODUCTION STORAGE AND OFFLOADING FACILITY , FROM: HTTP://OILANDGASPROCESSING.BLOGSPOT.PT /2009/02/OIL-RIG-OFFHORE-STRUCTURE.HTML. .......................................... 12 FIGURE 15 – SUBSEA SYSTEM COMPONENTS. BOP - BLOWOUT PREVENTER. FROM: HTTP://WWW.RYANSRANTINGS.COM/?P=817. .... 13 FIGURE 16 – JACK-UP INSTALLATION PROCEDURE. FROM: HTTP://IGSDELHICHAPTER.COM/IGSDC2014_FOUNDATIONSFOROFFSHORE.PDF................................................................................................................. 15 FIGURE 17 – SOME EXAMPLE SPUDCAN SHAPES (RANDOLPH ET AL., 2005) .............................................................................. 15 FIGURE 18 – PILE DRIVING METHOD TROUGH A JACKET LEG (DEAN, 2009)............................................................................... 18 FIGURE 19 – DETAIL OF THE MUDMATS AT THE BOTTOM OF THE STEEL LATTICE STRUCTURE , HTTP://WWW.TENSIONTECH.COM/SERVICES/TEXTILE _SOLUTIONS.HTML ......................................................................... 18 FIGURE 20 - ARRANGEMENT FOR PILE GROUP AROUND A LEG (DEAN, 2009) ............................................................................ 19 FIGURE 21 – STAGES IN INSTALLATION OF AN OFFSHORE DRILLED AND GROUTED PILE (RANDOLPH ET AL., 2005)............................... 20 FIGURE 22 – (A) PLANT OF EKOFISK TANK AND (B) CONDEEP GRAVITY BASE DESIGNS (RANDOLPH ET AL., 2005)............................... 23 FIGURE 23 – CONDEEP STRUCTURES INSTALLED , FROM: HTTP://WWW.INRISK.UBC.CA/FILES/2012/11/CONDEEP COMPARISON.PNG .. 23 FIGURE 24 –YOLLA A HYBRID GRAVITY BASED STRUCTURE EXAMPLE , BASS STRAIT, AUSTRALIA (WATSON & HUMPHESON,2005) ....... 24 FIGURE 25 – PRINCIPAL DESIGN ISSUES FOR PARALLEL DESCENT INSTALLATION : (A) DOWEL PENETRATION , (B) SKIRT PENETRATION , (C) BASE SUCTION OR PRESSURE , (D) DOME CONTACT STRESSES, (E) GROUTING PRESSURES AND DENSITY , AND (F) SCOUR PROTECTION (DEAN, 2009). ............................................................................................................................................................ 25 FIGURE 26 – (A) INSET, FOUNDATION TEMPLATE SHOWING CLUSTER OF CONCRET E CAISSONS (B) PREDICTED AND MEASURED CYCLIC LOAD VS. DISPLACEMENT RESPONSE (CHRISTOPHERSEN, 1993)............................................................................................. 26 FIGURE 27 – STAGES OF INSTALLATION OF A BUCKET FOUNDATION FOR A JACKET STRUCTURE ....................................................... 27 FIGURE 28 – SEEPAGE PRESSURES SET UP DURING SUCTION INSTALLATION OF BUCKET FOUNDATION (ERBRICH & TJELTA, 1999). ......... 28 FIGURE 29 – LOCATIONS OF SOME CURRENT DEEP -WATER DEVELOPMENTS (DEAN, 2009) .......................................................... 29

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Offshore Foundations: Technologies, Design and Application FIGURE 30 – (A) CATENARY MOORING AND DRAG EMBEDMENT ANCHOR . (B) TAUT LEG MOORING AND VLA. FROM: HTTP://EVENTS.ENERGETICS.COM/DEEPWATER/PDFS/PRESENTATIONS/SESSION5/RODERICKRUINEN.PDF ............................... 30 FIGURE 31 – GRAVITY ANCHORS (RANDOLPH ET AL., 2005) ................................................................................................. 31 FIGURE 32 – (A) ANCHOR PILE AND CHAIN. (B) FORCES ON AN ELEMENT OF THE ANCHOR LINE (RUINEN, 2005) ............................... 32 FIGURE 33 – MOORING APPLIANCE POINT AND VARIATION OF PADEYE DEPTH WITH LOADING ANGLE FOR GIVEN CENTER OF ROTATION (RANDOLPH & HOUSE, 2002)............................................................................................................................... 33 FIGURE 34 – SUCTION CAISSON INSTALLATION STEPS, FROM: HTTP://WWW.EPD.GOV.HK/EIA/REGISTER/REPORT/EIAREPORT/EIA _1672009/HKOWF%20HTML%20EIA/MAIN%20TEXT%2 0FIGURES/FIG%202.36A.PNG. ............................................................................................................................. 34 FIGURE 35 – DRAG ANCHORS, (A) FIXED FLUKE ANCHOR . (B) VRYHOF STEVMANTA VLA. SOURCE: VRYHOF ANCHOR MANUAL (2005) .. 35 FIGURE 36 – INSTALLATION METHODS ............................................................................................................................. 36 FIGURE 37 – (A) ANCHOR DEPTH DETERMINATION. (B) VLA SIMPLE RECOVERY , VRYHOF ANCHOR MANUAL (2005). ......................... 36 FIGURE 38 – COMPONENTS OF A SUCTION EMBEDDED PLATE ANCHOR (GAUDIN ET AL., 2006)..................................................... 37 FIGURE 39 –SCHEMATIC OF SEPLA INSTALLATION (YANG ET AL., 2012; AND COURTESY OF INTERMOOR). ..................................... 38 FIGURE 40 – RADIUS COMPARISON BETWEEN FLOATING UNITS LINKED TO CONVENTIONAL DRAG ANCHORS AND TORPEDO ANCHORS, FROM: HTTP://WWW.HINDAWI.COM/JOURNALS/JAM/2012/102618.FIG.002.JPG. ................................................................. 39 FIGURE 41 – DYNAMICALLY PENETRATING ANCHORS (A) TORPEDO ANCHOR WITH FINS AND WITHOUT FINS (MEDEIROS, 2002); (B) INSTALLATION OF 4 FLUKES TORPEDO ANCHOR (MEDEIROS, 2002; O’LOUGHLIN ET AL., 2004)............................................ 40 FIGURE 42 – FULL SCALE TORPEDO PILE AND RELEASING SITUATION , LIENG ET AL . (1999) ............................................................ 41 FIGURE 43 – ADVANTAGES AND DISADVANTAGES OF DIFFERENT DEEP WATER ANCHOR TYPES (EHLERS ET AL ., 2004). ........................ 42 FIGURE 44 – TOPOGRAPHICAL FEATURES OF OCEAN FLOORS, AFTER HTTP://WWW.MARINEBIO.NET/MARINESCIENCE/01INTRO/WOIMG/XSECTOPO.JPG. ......................................................... 45 FIGURE 45 – PROFILES OF OCEAN FLOOR TOPOGRAPHY AT SELECTED LOCATIONS , RANDOLPH ET AL. 2011....................................... 45 FIGURE 46 – BATHYMETRIC SHADED RELIEF MAP SHOWING A DETAILED EXTENT OF CONTINENTAL MARGINS (NUMBER IN RED RELATES TO LOCATIONS IN FIGURE 45), US NATIONAL GEOPHYSICAL DATA CENTRE........................................................................... 46 FIGURE 47 – SEDIMENT THICKNESS OF THE WORLD ’S OCEANS AND MARGINAL SEAS. SOURCE: HTTP://WWW.NGDC.NOAA.GOV/MGGLSEDTHICK/SEDTHICK.HTML. ...............................................................................48 FIGURE 48 – DISTRIBUTION OF SEDIMENT ACROSS A PASSIVE CONTINENTAL MARGIN , FROM: HTTP://CLASSCONNECTION.S3.AMAZONAWS.COM/927/FLASHCARDS/68927/JPG/4-91305062763712.JPG. .................... 49 FIGURE 49 - SEDIMENTARY PROCESS OF MARINE DEPOSITS (AFTER SILVA 1974) ........................................................................ 49 FIGURE 50 – TURBIDITY CURRENT: (A) SEABED TOPOGRAPHY WHERE SLOPE BREAK MAY START A TURBIDITY CURRENT (B) TURBIDITY CURRENT PROGRESS (C) SETTLEMENT TURBIDITY PARTICLES . FROM: HTTP ://PEOPLE.MATHS.OX.AC.UK/FAY/RESEARCH.HTML ....... 50 FIGURE 51 – DISTRIBUTION OF THE DIFFERENT SEDIMENTS DEPOSITS ACROSS THE WORLD , FROM: HTTP://CLASSCONNECTION.S3.AMAZONAWS.COM/927/FLASHCARDS/68927/JPG/4-91305062763712.JPG. .................... 51 FIGURE 52 – MICROGRAPHS OF (A) CALCAREOUS SAND AND (B) SILICA SAND (DEAN, 2009). ....................................................... 52 FIGURE 53 – WORLDWIDE DISTRIBUTION OF OFFSHORE OIL AND GAS DEVELOPMENTS . DEAN (2009) BASED ON MCCLELLAND (1974) AND POULOS (1988). ................................................................................................................................................ 53 FIGURE 54 – SCHEMATIC DIAGRAMS SHOWING MAIN OFFSHORE GEOHAZARDS (A) WIDELY USED SUMMARY OF DEEPWATER GEOHAZARDS (POWER ET AL ., 2005) (B) MORE GEOHAZARDS (STROUT AND TJELTA, 2007). ................................................................. 54 FIGURE 55 – SUMMARY OF GENERALISED INVESTIGATION ELEMENTS (CAMPBELL ET AL. 2008) ..................................................... 57 FIGURE 56 – TESTS THAT SHOULD BE PERFORMED DURING A LABORATORIAL TESTING PROGRAM (RANDOLPH ET AL., 2011). ............... 60 FIGURE 57 – DEFINITION OF SLIDE MOBILITY...................................................................................................................... 61 FIGURE 58 – COMPARISON OF VOLUME AND RUN -OUT DISTANCE OF SUBMARINE AND SUB-AERIAL SLIDES (RANDOLPH ET AL. 2011 AFTER SCHEIDEGGER 1973, EDGERS AND KARLSRUD 1982, HAMPTON ET AL. 1976, DADE AND HUPPERT 1998, DE BLASION ET AL . 2006) ............................................................................................................................................................. 61 FIGURE 59 – SÃO TOMÉ & PRINCIPE LOCATION AND SURROUNDING GEOLOGY, COURTESY OF AGENCIA NACIONAL DO PETROLEO OF STP FROM: HTTP ://WWW.STP-EEZ.COM/DOWNLOADS/POSTERS/3_STP_REGIONALGEOL.PDF . .............................................. 64 FIGURE 60 – CROSS SECTION FROM NIGERIA COST TO PRINCIPE ISLAND, COURTESY OF AGENCIA NACIONAL DO PETROLEO OF STP FROM: HTTP://WWW.STP-EEZ.COM/DOWNLOADS/POSTERS/3_STP_REGIONALGEOL.PDF ......................................................... 64 FIGURE 61 – CROSS SECTION FROM PRINCIPE ISLAND TO EQUATORIAL GUINEA COST, COURTESY OF AGENCIA NACIONAL DO PETROLEO OF STP FROM: HTTP://WWW.STP-EEZ.COM/DOWNLOADS/POSTERS/3_STP_REGIONALGEOL.PDF ......................................... 65

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Offshore Foundations: Technologies, Design and Application FIGURE 62 – THE RED CIRCLE LIMITS THE EXTENT OF GULF OF GUINEA. .................................................................................... 65 FIGURE 63 – DEEP-WATER SEDIMENTS PHYSICAL PROPERTIES ON GULF OF GUINEA (PUECH, 2004)............................................... 66 FIGURE 64 – DEEP-WATER SEDIMENTS PHYSICAL PROPERTIES , DATA FROM INTERNATIONAL OCEAN DISCOVERY PROGRAM TESTING RESULTS, AFTER MASCLE ET AL . (1998). SITES ARE LOCATED OFFSHORE OF IVORY COAST AND WITH DEPTHS RANGING FROM 2090 TO 4637 METERS, FIGURE 66................................................................................................................................ 67 FIGURE 65 – COMPARISON OF THE IODP TESTING RESULTS (OFFSHORE OF IVORY COAST) AND PUECH (2004) SEDIMENT PHYSICAL PROPERTIES IN GOG. ........................................................................................................................................... 67 FIGURE 66 – MAP SHOWING THE DRILLING LOCATIONS OF IODP IN GULF OF GUINEA, FROM: HTTP://WWWODP.TAMU.EDU/PUBLICATIONS/159_IR/IMAGES/MAP.JPG . ....................................................................................... 68 FIGURE 67 - PLASTICITY INDEX OF GOG SEDIMENTS IN DEPTH (PUECH, 2004). ......................................................................... 68 FIGURE 68 – PLASTICITY CHART OF GOG DEEP-WATER SEDIMENTS, CASAGRANDE DIAGRAM (PUECH, 2004). .................................. 69 FIGURE 69 – CV VERSUS W L CHART (KULHAWY & MAYNE, 1990) ........................................................................................... 69 FIGURE 70 – TYPICAL IN SITU STRESS PROFILES FOR GULF OF GUINEA DEEP-WATER SOILS AND THE COMPARISON WITH GULF OF MEXICO (PUECH, 2004).................................................................................................................................................. 70 FIGURE 71 – VERTICAL YIELD PRESSURE AND YSR VERSUS PENETRATION DEPTH FOR TYPICAL GOG SITES (J.L. COLLIAT & H. DENDANI ET AL., 2011)........................................................................................................................................................ 71 FIGURE 72 - UNDRAINED SHEAR STRENGTH VARIATION WITH DEPTH IN GULF OF GUINEA, DETERMINED BY INTERNATIONAL OCEAN DISCOVERY PROGRAM USING VANE SHEAR AND PENETROMETER TESTS , AND TEST LOCATIONS............................................... 73 FIGURE 73 – TYPICAL CONE RESISTANCE PROFILE FOR GOG DEEP WATER SEDIMENTS (WD – WATER DEPTH), PUECH (2004). ............. 74 FIGURE 74 – UNDRAINED SHEAR STRENGTH PROFILES OF GULF OF MEXICO, BRAZIL AND GOG PROPOSED BY PUECH (2004). .............. 75 FIGURE 75 – SHEAR STRENGTH PROFILES OF DIVERSE MARINE SITES ........................................................................................ 76 FIGURE 76 –NIGERIAN CONTINENTAL SLOPE STUDY AREA , SULTAN ET AL . (2007). ..................................................................... 76 FIGURE 77 – SHEAR STRENGTH PROFILE OF NIGERIAN CONTINENTAL SLOPE OBTAINED IN LABORATORY GEOTECHNICAL TESTS AND ITS COMPARISON WITH SHEAR STRENGTH PROFILES PROPOSED BY OTHERS ............................................................................. 77 FIGURE 78 – DESIGN PROCESS FOR TORPEDO ANCHORS. ...................................................................................................... 78 FIGURE 79- SCHEMATIC LONGITUDINAL SECTION DRAWING OF THE T-98, BRANDÃO ET AL. (2006). .............................................. 78 FIGURE 80 – PHOTOS OF THE T-98 BODY SECTIONS WELDING AND ITS FINAL ADJUSTMENTS , BRANDÃO ET AL. (2006). ...................... 79 FIGURE 81 – SENSITIVITY OF EMBEDMENT DEPTH PREDICTIONS TO SHAFT ADHESION FACTOR ACCORDING TO RICHARDSON ET AL. (2008) AND INTERPOLATION FOR ADHESION FACTOR OF 0.8 AND IMPACT VELOCITY OF 40M/S. ...................................................... 83 FIGURE 82 – SENSITIVITY OF EMBEDMENT DEPTH PREDICTIONS TO UNDRAINED SHEAR STRENGTH GRADIENT , RICHARDSON ET AL. (2008). ...................................................................................................................................................................... 84 FIGURE 83 – COMPARISON OF CENTRIFUGE AND FIELD TEST EMBEDMENT DATA , O’LOUGHLIN ET AL . (2013). .................................. 84 FIGURE 84 – CROSS SECTION OF A TORPEDO ANCHOR WITH FOUR FLUKES (AGUIAR EL AL., 2009). ................................................ 87 FIGURE 85 – T-98 HOLDING CAPACITY FOR 4 TILTS AND 3 DIFFERENT LOAD DIRECTIONS IN CAMPOS BASIN, BRANDÃO ET AL. (2006). .. 87 FIGURE 86 – TYPICAL SEPLA WITH KEYING FLAP , WANG ET AL. (2012) .................................................................................. 91 FIGURE 87 – PLATE ANCHOR SETUP BEFORE KEYING PROCESS . ............................................................................................... 93 FIGURE 88 – GRAPH SHOWING THE RESULTS ACHIEVED BY O’LOUGHLIN ET AL. (2006) FOR ECCENTRICITY RATIOS BETWEEN 0.17AND 1.0. ...................................................................................................................................................................... 94 FIGURE 89 – COMBINED LOADING PATHS FOR HIGH AND LOW ECCENTRICITY PLATE ANCHORS , O’LOUGHLIN ET AL . (2006). ................ 94 FIGURE 90 – LOSS IN ANCHOR EMBEDMENT DURING KEYING VERSUS ANCHOR GEOMETRY FACTOR , SONG ET AL. (2009). .................... 97 FIGURE 91 – PROBLEM NOTATION ................................................................................................................................ 100 FIGURE 92 – SHALLOW AND DEEP ANCHOR BEHAVIOUR , MERIFIELD ET AL. (2001) .................................................................. 101 FIGURE 93 – DESIGN CHART FOR RECTANGULAR ANCHORS IN CLAY , ALLOWS DETERMINING ANCHOR BREAK -OUT FACTOR NCO AT VARIOUS EMBEDMENT RATIOS (MERIFIELD ET AL ., 2003). ...................................................................................................... 102 FIGURE 94 – EQUATION AND CURVE DESCRIBING THE VARIATION OF NC VALUE IN SHALLOW FAILURE ZONE , DAHLBERG ET AL . (2004). . 104

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LIST OF TABLES TABLE 1 –DESIGN PARAMETERS FOR STEEL PILES IN SILICEOUS SANDS IN ISO (2004) .................................................................. 21 TABLE 2 – PILE DESIGN MULTIPLIER PARAMETER ( ) FOR CLAYS BASED ON OR OCR (SEMPLE AND GEMEINHARDT, 1983). .... 21 TABLE 3 – SHAFT FRICTION AND END BEARING CAPACITY ON CARBONATE SANDS BASED ON KOLK (2000). AND ARE RESPECTIVELY THE UNIT SHAFT FRICTION AND UNIT END BEARING OF SILICEOUS SAND , WHILE AND ARE RESPECTIVELY THE UNIT SHAFT FRICTION AND UNIT END BEARING OF CALCAREOUS SAND ......................................................................... 22 TABLE 4 – SURVEYS AND THEIR PURPOSE (DEAN, 2009) ...................................................................................................... 43 TABLE 5 – DRAG COEFFICIENTS ACCORDING TO DIFFERENT PENETROMETER SHAPE AND REFERENCE , ON FREEMAN AND HOLLISTER FORMULA L IS THE PENETROMETER LENGTH AND D IS THE DIAMETER. .............................................................................. 80 TABLE 6 – ADOPTED VALUES FOR DIFFERENT SOIL CHARACTERISTICS AND REFERENCES FROM WHICH THEY WERE BASED ON . ................. 81 TABLE 7 – MODEL AND PROTOTYPE ANCHOR DIMENSIONS AND SOIL PROPERTY . ........................................................................ 83 TABLE 8 – SOIL PROPERTIES, PENETRATION AND VERTICAL RESISTANCE OF CAMPOS BASIN IN BRAZIL AND STP OFFSHORE. AVERAGE UNDRAINED SHEAR STRENGTH GRADIENT WAS PROPOSED BY MEDEIROS (2001)................................................................ 89 TABLE 9 - ADVANTAGES AND DRAWBACKS OF SUCTION PILES AND DRAG EMBEDDED PLATE ANCHORS (VLAS), WILDE ET AL . (2001) ..... 90 TABLE 10 – PARTIAL SAFETY FACTOR FOR ANCHOR RESISTANCE , DAHLBERG ET AL. (2004). ........................................................ 105 TABLE 11 – PLATE ANCHOR HOLDING CAPACITIES IN THE IDEALIZED STP OFFSHORE CONDITIONS ACCORDING TO THREE DIFFERENT AUTHORS. ....................................................................................................................................................... 106 TABLE 12 – SEPLA AND TORPEDO ANCHOR ADVANTAGES. ................................................................................................. 108 TABLE 13 – SEPLA AND TORPEDO ANCHOR DISADVANTAGES. ............................................................................................. 108

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LIST OF SYMBOLS

 



σ’vy

α

ze , H Wsub wL w

vT vimpact v t su,o

St Rs Rb PI p’c p’0

Nco NC ma M0 m’

m M L

k,ρ h g Fs,max Fs Fn,max Fn Fd F f

Etotal e

deff D* d cv CD B Ap

a

∆  − 

angle Empirical reduction factor coefficient of lateral earth pressure Relative density yield vertical effective stress Adhesion factor Embedment depth submerged weight of the anchor in the water Liquid limit Water content Terminal velocity Impact velocity velocity thickness undrained shear strength at the ground surface sensitivity friction resistance End bearing resistance Plasticity index Maximum effective stress previously applied effective stress currently applied break-out factor end bearing capacity factor Mass of ballast Initial moment Effective mass mass Moment lenght Undrained shear strength gradient heigth Acceleration of gravity Maximum Parallel force Parallel force Maximum normal force Normal force drag force Force Shank resistance Total energy Loading eccentricity Effective diameter Optimal depth diameter Coefficient of consolidation Drag coefficient breadth Sectional area acceleration Loss in anchor embedment limiting unit skin friction Unit shaft friction xiv


,,  ′, 0   ,           ̅′ 

, su,avg

unit shaft friction of siliceous sands unit shaft friction of calcareous sand Cyclic shear strength actual vertical effective stress vertical effective stress Water density soil-pile friction angle unit weight of the soil Partial safety factor Installation penetration depth Mean static undrained shear strength Shape factor unit end bearing of siliceous sand unit end bearing of calcareous sand limiting bearing pressure on the base bearing pressure on the base Gradient for estimating anchor loss in anchor embedment Loading eccentricity for anchor weight Loading eccentricity for friction resistance Equivalent plate width Cyclic loading factor Bearing capacity factor Equivalent plate length Potential energy Kinetic energy Transversal area Shaft area average shaft undrained shear strength Overall submerge anchor weight undrained shear strength

xv

1. INTRODUCTION & MOTIVATION 1.1. CONTEXT & MOTIVATION The offshore oil industry started in 1947 with the installation of the first oil rig in just 6 metres of water, off the coast of Louisiana in the United States. Nowadays, there are over 7000 offshore platforms around the world, located in a large range of water depths which since the late1990s, are starting to exceed 2000 m. This progression forced a change in the concept of “deep water”; in the 1970s deep-water meant depths of 50 m to 100 m while now deep-water refers to

water depths around 800 m. Further to this, when referring to water depths greater than about 1000 m, the phrase “ultra-deep” water is now used. Figure 1 shows the evolution of exploration water depths through time, from the mid-20th century on.

Figure 1 – Trend in water depth for exploration and development drilling since 1940.

Over the past century, many advances have been made in the development of offshore technology, including the foundations of the structures that support the working platforms. There are currently, several potential foundation solutions for any given offshore site conditions. The circumstances of a particular site may be either fair or harsh depending on: the weather conditions such as wind, currents and waves; or the local conditions such as the water-depth, seabed geotechnical quality and topography, among other particular circumstances. Nowadays offshore platforms can be positioned either in shallow, deep or ultra-deep water columns, which results in very different foundation solutions. These definitions will be advanced in the following chapters. Not only the foundation systems vary with water depth, but also the offshore platform selected to explore the oil & gas reserve may differ. Figure 2 shows a diagram where the different foundation solutions are classified by water depth.

1


Marine Foundations

Shallow (<500 m)

Spudcans

Gravity Based Structures Tank

Condeep

Steel Buckets

Deep & Ultra-Deep Concrete Caissons Driven Piles

Piles

Embedded Anchors

Gravity

Grouted Piles

Box

Grillage and berm

Piles

Suction Caissons

VLA

SEPLA

DPA Torpedo Anchors

Figure 2 – Marine Foundations type by application and water depth. FPSO- floating production, storage and offloading vessel. VLA- drag embedment vertically loaded anchor. SEPLA- suction embedded plate anchor. DPA - dynamic penetrating anchor.

2


As near-shore and thus, shallow water resources have largely been exploited, the industries´ attention has turned to the exploration of more oil and gas fields that are further offshore and in deeper water. As a consequence, some countries where Portuguese is the official language have gained attention, and as a result some are currently under investigation (e.g. Mozambique, Guinea Bissau and São Tomé & Principe), while in others investments have already been made (e.g. Brazil and Angola). São Tome & Principe is a member of the group of countries which belong to the Community of Portuguese Language Countries (CPLP) and over the past five years has been subject to many field tests in order to quantify the potential oil & gas reserves, and to evaluate the quality of the potential extractable product (oil) of those reserves as well. Therefore, the chosen scenario to assess the feasibility of employing two different foundation systems for possible future production developments is off the coast of Sao Tome & Principe (STP). The exclusive economic zone (EEZ) of STP is now divided into several blocks which can be seen in Figure 3. Those blocks are licensed to Oil & Gas companies, so they can develop investigation work and evaluate the potential of the reserves.

Figure 3 – Map of São Tomé & Principe islands and the economic exclusive zone, courtesy of Agencia Nacional de Petróleo de STP.

3


As a final course thesis of Civil Engineering specializing in geotechnics, the work presented in this thesis has the purpose of gathering information about the foundation structures used in the offshore petroleum industry and the assessment of the application of two types of foundation in a realistic scenario – in this instance the developments proposed for STP described above.

1.2. THESIS STRUCTURE Following this introduction, Chapters 2 through 4 present a review of the literature relating to existing offshore foundation systems. In Chapter 2, offshore platforms are discussed; it is from these facilities where the operations for oil and gas extraction are developed. They can either be fixed platforms that stand above the water and are directly fixed t o the seabed or floating platforms that are fixed to the seabed by means of mooring systems. The review continues with the presentation of the two main foundation groups, i.e. “shallow” water (Chapter 3) and “deep &

ultra-deep” water (Chapter 4) foundations. Within each grouping the various foundation concepts are presented, and the design of these foundation systems is explained. Chapter 5 summarizes the different aspects involved in marine geotechnical characterization. First, the topographical features of ocean floors are explained and then the genesis of the seabed geology is described, i.e. the origin of the sediments that form the seabed. Lastly, geohazards that are commonly found in the seabed are described, as well as their triggering mechanisms. Chapter 6 presents a case-study developed for the offshore region of the Western African country São Tomé & Principe. The chapter starts by explaining which foundation systems will be assessed for their feasibility of use in that region; as this region typically has water depths greater than 1000 m, the proposed solutions will be based on anchor type systems. Then, the geotechnical site conditions are characterized based on the findings from the Gulf of Guinea published by other authors. Finally, a conceptual design of the foundations is made and the results obtained are discussed in terms of which might be more suitable for use.

4


2. OFFSHORE PRODUCTION FACILITIES 2.1. INTRODUCTION Many factors may influence the selection of an appropriate development strategy for offshore hydrocarbon deposits. Among those factors considered are (French et al., 2006): 

Water depth



Reservoir size



Proximity to existing infrastructure



Number of wells



Operating considerations



Economic factors



Anticipated well intervention frequency (e.g. frequency of well replacement)

Offshore structures can be broadly categorised as fixed platforms, compliant towers, floating structures and subsea systems. The various types of offshore development systems currently in use are shown in Figure 4.

Figure 4 – Offshore developments systems: Bottom supported, vertically moored structures, floating production and subsea systems, from: http://i46.photobucket.com/albums/f133/tamaaa/upload1/Picture1.png.

2.2.

FIXED PLATFORMS

Fixed platforms are working decks supported by legs directly connected onto the seabed foundations. They can be tubular steel jackets, concrete or hybrid gravity structures. Steel jackets 5


are primarily pile supported, while gravity structures achieve stability by virtue of their immense structural weight and large diameter base. Additional stability may be provided by use of base skirts which penetrate several meters into the seabed. The limitation of this kind of structures is related with economic issues; therefore they are not used in water depths greater than 500 m. Figure 7 shows the jacket of the Bullwinkle platform located in Gulf of Mexico, the platform has total height of 529 m and it extends 412 m below the waterline.

Figure 5 – Steel jacket platforms and concrete gravity structure, from: http://petroahdal.webs.com/apps/photos/photo?photoid=109394836 and http:// offshoremag.com/articles/2013/jan/go-ahead-for-14-billion-hebron-project-offshore-easterncanada.html.

Figure 6 – Offshore Platform elements, from: http://www.conservation.ca.gov/dog/picture_a_well/Pages/offshore_platform.aspx.

6


Figure 7 – Transportation of the Bullwinkle steel substructure (jacket), courtesy of Shell Intl. http://www.esa.org/esablog/wp-content/uploads/2011/10/Steel-jacket-being-towedoffshore.jpg.

2.3.

COMPLIANT TOWER

A compliant tower is a slender steel space-frame tower with a piled foundation. These towers are designed to be more flexible in bending than fixed structures and, therefore, more “compliant” to the environment . This flexibility means that the platform can withstand significant

lateral loads by sustaining large lateral deflections. Compliant towers are typically applicable in water depths ranging from 500 m to 600 m. The tallest in the world is the Petronius tower, illustrated in Figure 8, which stands in 535 m of water and has a total height of 609.9 m, making it one of the highest freestanding structures ever built (Chevron 2000).

Figure 8 – Petronius compliant tower and Empire State Building for comparison, from: http://petrowiki.org/File%3AVol3_Page_530_Image_0001.png.

7


2.4.

TENSION LEG PLATFORM

A Tension Leg Platform (TLP) consists of a semi-submersible platform moored by vertical tendons connected to the seafloor. The excess buoyancy provided by various hull components maintains the tension in the mooring system even during storm loading conditions. TLPs are capable of being used in water depths up to 2000 m. Figure 9 illustrates the worldwide fleet of installed TLPs, with special attention to Kizomba A platform (see arrow) which is located in Angola (country member of CPLP), in nearly 1200 m of water.

Figure 9 – Worldwide fleet of installed and sanctioned TLPs, courtesy of British Petroleum from: http://petrowiki.org/File%3AVol3_Page_532_Image_0001.png.

2.5.

SEMI-SUBMERSIBLE FLOATING PRODUCTION SYSTEMS

A semi-submersible floating production system (FPS) typically comprises parallel pontoons connected to the topside by numerous vertical columns, Figure 10. Semi-submersible platforms can be moved from place to place, and the pontoons and columns can be filled with water to alter the buoyancy of the system for improved stability under wave and wind loading. Semi-submersibles can be deployed in a wide range of water depths for both temporary and permanent operations. Figure 11 shows the FPSs used worldwide and the wide range of water depths (300 m to 2000 m) where the system has been deployed.

8


Figure 10 – Semi-Submersible Vessel with twin hulls (columns and pontoons). http://oilandgasprocessing.blogspot.pt/2009/02/oil-rig-offhore-structure.html

Figure 11 - Worldwide fleet of installed and sanctioned semi-submersible FPS, courtesy of British Petroleum from: http://petrowiki.org/File%3AVol3_Page_531_Image_0002.png.

9


2.6.

SPAR PLATFORM

A SPAR consists of a large diameter, vertical, cylindrical hull which supports the platform by means of excess buoyancy. Buoyancy chambers located near the top of the hull enable the buoyancy of the structure to be controlled thereby maintaining platform stability. In addition, spiral strakes fitted to the hull minimize lateral movement due to vortex shedding, improving lateral stability. The SPAR can be anchored to the seabed by vertical tethers but catenary or taut mooring lines are more common. These details can be seen in Figure 12, where two different SPAR profiles are illustrated, i.e. a classic SPAR with soilid hull and truss SPAR. SPARs are theoretically capable of being deployed in water depths up to 3000 m (Figure 13), and currently the deepest platform in the world is the Perdido SPAR in the Gulf of Mexico, which floats in 2438 meters of water.

Figure 12 – Three different SPAR profiles, from: http://images.pennwellnet.com/ogj/images/ogj2/9644jsk02.gif .

10


Figure 13 – Worldwide fleet of installed SPARs, courtesy of British Petroleum.

2.7.

FLOATING PRODUCTION STORAGE AND OFFLOADING FACILITY

Floating Production Storage and Offloading (FPSO) facilities comprise a large tanker type vessel fitted with production and storage facilities, Figure 14. The storage capabilities of FPSO mean that it may be suitable for economically marginal fields located in remote areas in which pipeline infrastructure does not exist. Smaller shuttle tankers may be used to transport the hydrocarbons to an onshore processing facility. FPSOs can be fixed in position or comprise multiple mooring lines meeting at a single point. The single point mooring allows the tanker to weathervane to achieve an optimal orientation with regard to the prevailing environmental conditions. The key advantages of FPSOs relate to their ability to operate on short term or permanent developments in water depths up to and exceeding 3000 m. For example, the P-50 FPSO is moored in approximately 1240 m of water in the Albacora Leste field in the deepwater Campos Basin, Brazil (Brandão et al. 2006).

11


Figure 14 - Floating production storage and offloading facility, from: http://oilandgasprocessing.blogspot.pt/2009/02/oil-rig-offhore-structure.html.

12


2.8.

SUBSEA SYSTEM

Subsea systems typically comprise either a single subsea well, whose production is piped to a nearby platform, or multiple wells producing through a manifold and pipeline system to a distant production facility. Multi-component seabed facilities such as subsea wells, manifolds, control umbilicals and flowlines allow subsea systems to recover hydrocarbons in water depths and conditions that would normally preclude the installation of a conventional fixed or floating platform. Subsea systems are capable of operating in any water depth (Richardson et al., 2008). Figure 15 illustrates several subsea system components.

Figure 15 – Subsea system components. BOP - Blowout preventer. From: http://www.ryansrantings.com/?p=817.

13


3. SHALLOW WATER FOUNDATIONS Water depth is considered shallow if the seabed depth does not exceed about 500 meters. The principal types of foundation used in this situation are: 

Spudcans



Piles



Gravity base structures (GBS)



Concrete caissons



Steel Buckets

All these solutions have advantages and disadvantages, some cannot be applied in all seabed soil conditions and some are more favourable than others, e.g. GBS in soft clay soils induce higher settlements than spudcans do, due to its enormous self-weight. Hence, the next sections will describe their principal properties as well as their pros and cons.

3.1. SPUDCANS In the offshore industry an important role is played by self-elevating mobile drilling units, commonly known as jack-ups, due to their flexibility and cost-effectiveness (Randolph et al., 2005). It has proved to be a very useful construction “tool”, especially when working in turbulent

sea areas, or breaking waves such as shoal or coastal waters, and in swift currents (Gerwick Jr. et al., 2007). These structures consist of a buoyant triangular unit resting on three or more retractable legs. This unit supports drilling and other topside equipment; it moves onto the intended location with legs retracted, then releases the legs onto the seabed, and raises the hull out of the water, as shown on Figure 16. On jack-ups, the foundation legs operate independently of each other, and their foundations are usually known as “spudcans”.

14


Figure 16 – Jack-up installation procedure. From: http://igsdelhichapter.com/IGSDC2014_FoundationsforOffshore.pdf.

These foundations have a unique geometry, since they are installed relying only on the structure’s self -weight plus an additional designed preload, which is intended to minimize

settlement and improve resistance to environmental solicitations. Spudcans are roughly circular in plan, typically they have a shallow conical underside (in the order of 15 to 30 degrees to the horizontal) with a sharp protruding spigot. In the larger jack-ups operating today, the spudcan diameter can exceed 20 meters, with the shapes varying with the manufacturer and rig (Randolph et al., 2005). Figure 17 illustrates two different spudcan shapes. The usual height of a jack-up structure is over 160 m.

Figure 17 – Some example spudcan shapes (Randolph et al., 2005)

Since jack-ups started to be applied in deeper water and harsher environments, preloading also started to play a more important role in design. Preloading induces bearing capacity failure in the soil beneath and around each spudcan, causing the spudcan to penetrate into the seabed until the soil resistance equals the applied load. Using the jacks on one leg at a 15


time, the barge acting as the reaction, the legs are forced into the soil. After these a pile hammer may be used on the top of the legs to gain even greater penetration. If the same penetration was given to all legs a punch-through failure or a damage on the foundation might occur during preloading, which could result in the jackup toppling over and one or more legs being bent or broken. So, to avoid these risks, modern jackups are able to preload the spudcans individually (Dean, 2009). 3.1.1. GEOTECHNICAL C ALCULATIONS Jackups are mobile units, and when designed the foundation conditions where it will be applied are not usually known. So, for each new site, SNAME (2002) recommends that a sitespecific assessment be done. This may include: 

Assessment of geohazards



Foundation assessment for installation, commonly including a preload check



Foundation assessment for operations, including a sliding check and an overturning check based on assessed vertical and horizontal actions



Assessment of effects of the jackup on nearby structures



Leg extraction assessment, for when the jackup is moved off site to another location These analyses are preformed to predict the footing penetration during installation and

preloading, and the capacity to withstand a design storm.

3.2. PILE FOUNDATIONS Piles are slender columnar elements in a foundation which have the function of transferring load from the superstructure through weak compressible strata or through water, onto stiffer or more compact and less compressible soils or onto rock. They may be required to carry uplift loads when used to support tall structures subjected to overturning forces from winds or waves. Piles used in marine structures are subjected to lateral loads from the impact of berthing ships and from waves (M.J. Tomlinson, 1977). Pile foundations can be used either in shallow or deep water, the link to the working platform is what differs. In shallow water, the connection is typically made by a steel lattice structure commonly called a jacket. This is the most used structure for fixed offshore platforms. Piles can also be used as anchors in moored floating facilities, and this application will be discussed further in Chapter 4. There are two construction methods used for piles that are constructed offshore: driven and grouted. The most commonly used are metallic driven piles because they are the most reliable 16


and have the easiest construction path. Although we are in a marine environment, there is no problem with corrosion as the steel pile embedded in the seabed, has no contact with oxygen. 3.2.1. DRIVEN P ILES Offshore, the most frequently used pile type is the open-ended steel pipe, which is driven into the seabed by a hammer. Pile diameters range from 0.76 m up to 2.5 m, but exceptionally a diameter of 5.1 m has been successfully used on offshore wind turbines. The wall thickness may vary along the pile length, so it will be thicker where moment is greater (near the pile head). Typical diameter to wall thickness ratios (d/t) are between 20 and 60. The lower value represents the greatest curvature that can normally be achieved in a steel rolling machine. The highest value represents a curvature beyond which wall-buckling or section ovalisation effects can be common (Barbour and Erbrich, 1994; MSL, 2001; Aldridge et al., 2005; Randolph et al., 2005). The process of installing an offshore driven pile through a steel jacket leg is illustrated in Figure 18. First, the steel jacket is released onto the required location, where it will be supported by mudmats. Mudmats are templates used in the bottom of the steel jacket to avoid its undesirable penetration into soft soils, Figure 7 and Figure 19 shows four mudmats at the bottom of a steel jacket structure. Then the first section of pile is lowered trough the leg. A hammer is installed on the head of the pile, and is used to drive the first segment until its limit, the pilling equipment is removed and another pile segment is lifted on and welded in place. This weld is normally subjected to non-destructive testing, after which the hammer is lifted back on, and the whole procedure starts again until the designed pile penetration is achieved. Unless the jacket confines the pile, grout is injected into the annular space to provide the structural connection between them (Dean, 2009).

17


Figure 18 – Pile driving method trough a jacket leg (Dean, 2009)

Figure 19 – Detail of the mudmats at the bottom of the steel lattice structure, http://www.tensiontech.com/services/textile_solutions.html

When more than one pile is required per leg, sleeves may be attached to the jacket around the base of its legs. These sleeves operate both as guides and pile pooler. Piling is done in the same way as for a leg pile, with each pile installed in several segments, if necessary. Figure 20 shows the sleeve arrangement for pile group.

18


Figure 20 - arrangement for pile group around a leg (Dean, 2009)

3.2.2. GROUTED P ILES Even though driven piles are the most common type used in the offshore environment, there is also the equivalent of a bored pile. It involves the grouting of a steel section, which is inserted in a previously drilled hole. Figure 21 shows the stages in construction of a drilled and grouted pile. In order to avoid collapse of loose uncemented material near the seabed, it is often necessary to drive a primary pile first; alternatively stabilizing mud can also be used. This solution is only used if an adequate drilling barge is already on the site, since it is more expensive to install and has longer construction period than driven piles (Randolph et al, 2005). Whenever the seabed is composed of calcareous sediments, and other potentially crushable material, where the shaft friction obtained with driven piles has been found to be very low, drilled and grouted piles are more reliable. The low shaft friction is associated with very low radial effective stresses around the pile, a situation remedied by drilled and grouted pile construction, where the original horizontal effective stresses in the ground can be restored by appropriate grouting design (Randolph et al., 2005).

19


Figure 21 – stages in installation of an offshore drilled and grouted pile (Randolph et al., 2005)

3.2.3. P ILE RESISTANCE Although many studies have been made into the understanding of end-bearing and shaft friction resistance of piles, design methods still rely on empirical methods. The determination of the soil resistance can be made applying current offshore guidelines, e.g. API RP2A and ISO 19902, or CPT- based methods. The latter’s advantage is t hat it takes account of the detailed stress history of the soil around the pile (Randolph et al., 2005; Dean, 2009). Unit parameters

Table 1 summarizes the unit parameters recommended by API RP2A and ISO 19902, and also gives some data for carbonate sands. For granular material, drained conditions are assumed to apply. The parameters are based on relative density and silt content. The design approach for shaft friction is expressed as,

 =   ≤ − where K is a coefficient of lateral earth pressure, and the recommended values are between 0.7 and 0.8 for open ended piles loaded in compression and 0.5 to 0.7 for tension piles, vertical effective stress,



is the soil-pile friction angle, and

which varies with soil type and density.

−

is the

is a limiting unit skin friction,

The limiting bearing pressure on the base of the pile is expressed as, 20




Where



 =  ≤ − ranges from 12 to 50 according to the grain size and relative density of the material. All

parameters needed to determine pile resistance in sand are given in Table 1. Siliceous Sands Soil description

Sand

Silty Sand Clayey Sand

Sandy Silt

 − (˚)

(kPa)

 −

Soil density

Dr(%)

Loose

15-35

20

65

12

3

Medium

35-65

25

80

20

5

Dense

65-85

30

95

40

10

Very Dense

85-100

35

115

50

12

Loose, Med

15-65

20

65

12

3

Dense

65-85

25

80

20

5

Very Dense

85-100

30

95

40

10

Loose

15-35

15

45

8

2

Med. Dense

35-85

20

65

12

3

Very dense

85-100

25

80

20

5

(MPa)

Table 1 –Design parameters for steel piles in siliceous sands in ISO (2004)

For cohesive material, axial pile failure is assumed to occur in undrained conditions. Shaft

  = 0.2 × .

friction is calculated as multiple, α of the undrained shear strength the ratio

/

. The multiplier depends on

or in terms of the overconsolidation ratio (OCR), using Semple and Gemeinhardt’s

(1983) relation. This last interpretation also expresses, recommends that the unit end bearing be taken as 9 determine the adhesion multiplier, α. ISO 19902 and API RP2A



. API RP2A

in clay. Table 2 provides the way to

Interpreted in terms of OCR

   / / ≤ .   ≤ 1. 3 . ≤ / ≤  0.5/ /.. 1.3 ≤  ≤ 6.6 1.11/..  ≤ / 0.5/ / 6.6 ≤   0.74/ Range of

Approximate range of OCR

1

Table 2 – Pile design multiplier parameter ( Gemeinhardt, 1983).

1



) for clays based on

/

or OCR (Semple and

It is now recommended that grouted piles are better suited to soil profiles consisting primarily of calcareous and carbonate materials (Kolk, 2000). For soil profiles that contain thin 21


calcareous and carbonate sand layers, driven piles may still be feasible. Kolk’s recommendat ions

for open-ended driven piles in calcareous soils are summarized in Table 3. Range of carbonate contents: CC

   − 

,

Unit end bearing,

 

Notes

 ≤ % % ≤  ,  ,  ,       = log20 ≤ % l o g4 % ≥  min0.14 ,15 0.7  = 10 0.3   = 3   As siliceous sand

As siliceous sand

-

for coring,

and

for plugged.

determined from CPT cone (MPa).

if

no CPT data available

Table 3 – Shaft friction and end bearing capacity on carbonate sands based on Kolk (2000).



are respectively the unit shaft friction and unit end bearing of siliceous sand, while

are respectively the unit shaft friction and unit end bearing of calcareous sand.

,

and

,  and

3.3. GRAVITY BASE STRUCTURES Gravity base structures (GBS) are designed to be founded at or just below the seafloor, transferring their loads to the soil by means of shallow footings. Usually these structures are made of reinforced and prestressed concrete, but some were built of steel or a hybrid of concrete and steel. These structures have a large base “footprint” with purpose of minimizing soil-bearing loads. An important advantage of these solutions is the possibility of oil storage within the base structure, i.e. the base operates both as foundation and storage facility. GBS are also used for offshore wind power plants. By the end of 2010, 14 of the world's offshore wind farms were supported by gravity-base structures. Design loads for an offshore GBS are superior to onshore conditions. Due to its large volume, inertial forces under waves, earthquake, and impact from vessel or icebergs are much greater than usual. Thus, sliding tends to become the dominant mode of failure. So, to prevent this possibility, concrete or steel skirts and dowels are employed; these are designed to penetrate the seabed and thus force the failure surface deeper below the seafloor. Skirts also provide protection against scour and piping (Gerwick Jr. et al., 2007). These structures have evolved over time, the first of its kind was the Ekofisk tank, which was installed in the North Sea in 1973 (Clausen et al., 1975). The experience gained on this first project lead to conceptualization of a better, and now common, concrete deep water structure 22


called Condeep. A Condeep gravity base comprises a number of cylindrical cells usually displayed in a hexagonal arrangement, the underside of the cells has a convex profile and half a metre inside the concrete skirts the top of the dome touches down on the seabed, as illustrated in Figure 22(b). In Figure 22 it is visible the more complex design of Condeep foundation relative to the Ekofisk tank is apparent. While the Ekofisk tank had short (40cm) concrete skirts, the Condeep has steel skirts (to 3.5m), that project from concrete skirts. The condeep design has much smaller wave forces acting on the structure as the major volume is located below the water surface. In Figure 23, it is possible to see this characteristic on some different condeep platforms.

Figure 22 – (a) Plant of Ekofisk tank and (b) Condeep gravity base designs (Randolph et al., 2005).

Figure 23 – Condeep Structures installed, from: http://www.inrisk.ubc.ca/files/2012/11/CondeepComparison.png

Another unusual group of GBS has been developed, and it is a hybrid concrete-steel solution, i.e. a mixed structure which has a concrete base and a steel lattice structure, Figure 24. They have been applied at sites where calcareous muddy silts and sands dominate, because of their lighter weight when compared to concrete GBS. The rapid construction time is a very important reason for choosing this kind of solution (Dean, 2009). 23


Figure 24 –Yolla A Hybrid gravity based structure example, Bass Strait, Australia (Watson & Humpheson,2005)

Gravity foundations are often considered to be more complicated than jackets because the soil behaviour must be considered in a three-dimensional volume that stretches one or more base diameters below, and several diameters to either side of the base. The main design requirement is to determine the foundation footprint, and the skirt depth and spacing. There are three design codes for gravity platforms, ISO 19903:2006, API RP2A and DNV-OS-C502:2012. Besides structural design another really important aspect of the implementation of these structures is the foundation construction method. The foundation is constructed in a dry dock, but the most difficult phase of the construction is when fixing it to the sea ground. Most gravity platforms are kept level as they are lowered to the seabed. This allows the dowels and then the skirts to penetrate the seabed almost simultaneously across the footprint of the base. Dowels are steel pipes up to about 2 m in diameter, they contact the seabed first and pin the platform to the seabed and penetrate a few meters below all other components. As for the skirts, they can be of steel (a few centimeters thick) or concrete (about a meter thick). Their height is determined by the need to transfer vertical load to competent layers and provide shear against horizontal forces, and design is similar to piles. In Figure 25, all the issues that must be considered while designing and installing the platform are illustrated (Dean, 2009).

24


. Figure 25 – Principal design issues for parallel descent installation: installation: (a) dowel penetration, (b) skirt penetration, penetration, (c) base b ase suction or pressure, (d) dome contact stresses, (e) grouting pressures and density, and (f) scour protection (Dean, 2009).

3.4. CONCRETE CAISSONS FOR TENSION LEG PLATFORMS Concrete caissons evolved from deep skirted concrete base foundations, and comprise individual or clusters of small concrete caissons or “bucket” foundations. Figure 26(a) 26(a) illustrates

a TLP and its foundation system. This foundation system has the particularity that the resistance is provided by a combination of concrete self-weight and the interaction between the caisson and seabed (Randolph et al., 2005). The average static tension is counteracted by the weight of concrete foundation, while load originating from cyclic waves and wind (design storm) are transferred to the soil by skirt friction and suction under the top cap (Stove et al., 1992). During a storm, the TLP will drift out of alignment with the foundations introducing a moment action into them, which gives the most critical load situation (Christophersen, 1993). Figure 1993). Figure 26(b) 26(b) presents the limit equilibrium (upper solid line) and three dimensional finite element (lower solid line) analysis predictions, as well as measured values of bearing capacity and displacement for the concrete caisson cluster with tension loading (Christophersen, 1993).

25


Figure 26 – (a) Inset, Foundation template showing showing cluster of concrete caissons (b) Predicted and measured cyclic load vs. displacement displacement response (Christophersen, 1993)

The concrete caisson installation procedure, involves: 

The free release of the concrete caisson(s) onto the seabed, this step is very important to ensure a minimum penetration by means of the caisson self-weight, so a water tight seal is provided to withstand the differential pressures applied during the subsequent application of suction (Stove et al., 1992).



Water evacuation from the inside of the caisson is commenced by means of pumps located on its top. As the soil responds under undrained conditions, the caisson skirts are forced to penetrate the soil due to t o the internal negative pressure (suction).



Afterwards the top of the foundation that has not penetrated into the seabed can be covered with ballast to increase the weight and confinement.



Finally, the connection and tensioning to the TLP can be made. The resistance calculation for the skirts is similar to that for piles; from observation the

adhesion coefficient α is in the range of 0.15 and 0.30 depending on the material (Randolph et al., 2005). The resistance available due to suction has to be analyzed on a site-by-site basis, since it depends on the soil characteristics, such as permeability, and load magnitude. The greater the cyclic tension loads, the greater the suctions that will be developed. Caissons have certain advantages over piles as anchors for deeper water moorings, if they can provide enough tensile capacity. For example, the pumps used for caisson installation do not have the same problems as piling hammers at great working depths (even though new systems are being developed for the latter to allow operations in water depths of 3 km). Also, the larger 26


diameter of caisson foundations provides a larger area for ballast and can also mobilize greater reverse end bearing or passive suction during uplift compared to a pile foundation (Clukey et al., 1995).

3.5. STEEL BUCKETS FOR JACKETS Steel buckets (also known as suction cans) are used as an alternative to pile foundations for jackets. They have also been used extensively for offshore wind turbine foundations. These suction foundations are steel cylindrical structures, closed on one end and open on the other. Bucket foundations often exceed 5 m diameter, some reach 10 or 20 meters. The T he jackets, to which the buckets are attached, have much larger dimensions, d imensions, so the installation of these t hese structures can prove to be quite difficult to perform with the use of only a crane. Therefore, pontoons are used to bring the structure to site, from which it is launched into the water. Then, the structure dives by slowly filling the hollow sections with water and the entire structure is aligned with the help of a crane so it is correctly placed on the seabed. In the same manner as for concrete caissons, for installation, inst allation, the open end of the bucket is placed on the seabed and the water contained within the cylinder and the floor is pumped out. This creates a vertical load on the structure, pulling it into the ground. Figure ground. Figure 27 illustrates the three steps of the installation, first the structure touches down on the seabed, then the ballast containers are filled so the structure penetrates the soil under its own weight, and finally, water is pumped from the caissons producing suction penetration. Suction foundations can be applied in sands as well as in clays, and in softer clays they work very well (Randolph et al., 2005). Depending on which soil the steel buckets are installed, different failures in the soil around the t he bucket can occur as discussed below.

Figure 27 – Stages of installation of a bucket Foundation for a jacket structure

27


3.5.1. I NSTALLATION IN C LAY One of the failure mechanisms is that when there is a lot of friction on the outside of the can, too high a suction is developed and plastic failure of the soil occurs. In this case, instead of penetration of the can, the soil can be sucked up into the can instead. This phenomenon is called reverse end bearing failure. Normal end bearing failure occurs when after installation, too large a load is placed on the soil inside and right underneath the can. The reverse can happen if either the suction applied or the tension load on the can is too large. 3.5.2. I NSTALLATION IN S AND In sands a different failure mechanism can occur, and it is related to the inflow of water. In Figure 28 the flow net into the suction can has been sketched; during the suction penetration phase, pumping creates an under-pressure across the foundation baseplate and, and more importantly, sets up seepage flow that reduces tip resistance and internal skirt friction. The inflow will always be present because of the pressure differential between the sea and the inside of the can and the fact that sands are relatively permeable. If the suction produces a very fast flow, a large flow gradient can occur, which will decrease effective stresses and ultimately liquefaction can occur. If this does not occur in all the cans at the same time, accidents may happen.

Figure 28 – Seepage pressures set up during suction installation of bucket foundation (Erbrich & Tjelta, 1999).

4. DEEP AND ULTRA -DEEP WATER FOUNDATIONS The demand for oil products and natural gas, has forced companies to search for resources in increasingly remote sites. Many of these sites are offshore and have water depths in excess of 1000 m, Figure 29. Water depths in excess of 5 00 m are considered to be “deep”, and “ ultra-deep” when greater than 1000 m. The economic investment associated with developing these sites is huge, so it is vital to produce solutions with an optimal balance between reliability and economy.

28


Clearly, at these great depths it becomes increasingly impracticable to build load transfer structures such as jackets or gravity based structures. Therefore, different foundation solutions have been adopted, i.e. anchors with mooring systems. There is a vast range of solutions which are sub-divided between gravity anchors and embedded anchors. Gravity anchor types include: 

Boxes



Grillage and Berm

Embedded anchor types include: 

Anchor piles



Suction caissons



Drag anchors (fixed fluke)



Vertically loaded drag anchors (VLA)



Suction embedded plate anchors (SEPLA)



Dynamically penetrated anchors (DPA)

Figure 29 – Locations of some current deep-water developments (Dean, 2009)

Many current deepwater developments are close to the continental rise, and so are subject to additional potential geohazards associated with possible land sliding (Dean, 2009). These aspects are presented and discussed in Section 5. The mooring system plays an important role in deep and ultra-deep waters, since it is not viable to build and install load transfer structures from the waterline onto the seabed in such deep

29


waters. To overcome this issue, several types of anchors were created to be attached to mooring systems, there are two main types of mooring: 1. Catenary mooring systems are generally used in shallow to deep water (up to 1000 metre). Chain or wire rope mooring lines are used, of which a significant length lies on the seabed, and the anchor is loaded in a horizontal direction. Typically conventional drag embedment anchors are used in these systems. Figure 30(a) shows the geometry of a catenary mooring system and a common drag anchor in profile. 2. Taut leg mooring systems are typically used in deep and ultra-deep water (greater than 1000 metre). Mooring lines used are light weight (synthetic rope or wire rope), and they enter the seabed at a significant angle. The anchor is loaded in the horizontal and vertical direction. Vertical loaded anchors (VLA) are commonly used, although there are diverse solutions. Figure 30(b) illustrates the straight alignment of the taut leg mooring system and an exemplar of a VLA.

Figure 30 – (a) Catenary mooring and drag embedment anchor. (b) Taut leg mooring and VLA. From: http://events.energetics.com/deepwater/pdfs/presentations/session5/roderickruinen.pdf

4.1. GRAVITY ANCHORS An anchoring system is normally required to provide resistance forces that are primarily horizontal, with cyclic as well as static components. Gravity anchors consist of heavy weight steel structures (box, or grillage), filled or covered with granular fill (either rock-fill, or heavier material

30


such as iron ore), and placed on the seafloor. Simply, the structural element is placed first, and then the bulk fill is added (Randolph et al., 2005; Dean, 2009). Two different structures are represented in Figure 31, on the left is a conventional box anchor filled with iron ore, which provides ballast, and on the right a covered grillage. The latter is considerably more efficient in terms of quantity of steel for a given holding capacity, but is much less efficient in terms of the quantity of ballast required. Design of this type of anchor is also more complex since a variety of failures mode must be considered, ranging from sliding of the complete berm, pulling out of the grillage, or combinations involving asymmetric mechanisms (Randolph et al., 2005).

Figure 31 – Gravity anchors (Randolph et al., 2005)

4.2. PILE ANCHORS Pile anchors have a similar behaviour to pile foundations (i.e. little skin friction developed in calcareous soil), but the construction method and the forces they need to withstand are different. They are very effective in many soils, and can either be drilled in and grouted using an offshore mobile drilling rig, or driven in with an underwater hammer. Advances have been made to allow hydraulic hammers to work in deep waters and with greater power, so the driving of the piles does not become an issue (Gerwick et al., 2007). The anchor pile system consists of a mooring chain or cable, and the pile. Figure 32(a) shows a simple system of an anchor pile and chain. In a catenary mooring system, the chain is laid along the seabed describing a smooth curve, and the anchoring force that it provides includes the weight of the line, the friction on the seabed, and the frictional resistance from the soil on the buried part of the anchor line, as well as the pull-out resistance of the pile itself. The soil resistance along the length of a buried chain or cable can be a significant proportion of the overall anchoring 31


resistance provided by the system. Figure 32(b) shows the forces on an element of the anchor line in a catenary mooring system. In a taut mooring system, the line is tensioned, and rises from the seabed without passing along the seafloor. The concept of an anchor pile starts with a pad-eye that is attached to a pile and a line is attached to the pad-eye, Figure 32(a). The location of the pad-eye is designed with the purpose of reducing potential rotation of the pile when loaded. Then the pile is driven into the seabed, and may be left protruding slightly above the seafloor, so as to be retrieved later. After installation, the chain is attached to the floating platform and tightened. The most difficult anchoring soil of all is a soft mud, silt, or loose sand overlying a hard material such as conglomerate or very dense sand and silt. So, a conventional pile may be placed in holes excavated by clamshell bucket and then back filled with dumped rock (Gerwick et al., 2007). However, this operation gets increasingly difficult with larger water depths.

Figure 32 – (a) Anchor pile and chain. (b) Forces on an element of the anchor line (Ruinen, 2005)

4.3. SUCTION CAISSONS Although concrete caissons have been used, the majority of suction caissons are fabricated from steel, which have a similar concept to steel buckets on shallow foundations. Suction caissons operate as anchors, and vertical capacity is granted by the weight of the plug of soil inside and the 32


friction on the outer surfaces, and in addition, the characteristic negative end-bearing resistance. The latter, as in a steel bucket, is the force required to separate the lower end of the soil p lug from the undisturbed soil. Typically, suction anchors are open at the bottom and closed at the top. They have large diameters, typically more than 5 meters in diameter and are 20 to 30 meters in length, with a length to diameter ratio (L/d) in the range of 3 to 6. Normally the cylinders have very high ratios of diameter to wall thickness (d/t ~100 to 250), that require internal stiffeners to prevent structural buckling during installation, and due to the large lateral loads imposed by taut moorings. Mooring loads are applied by an anchor line attached to the side of the caisson at a depth that optimises the holding capacity. Usually this requires the line of action of the load to pass through a point on the axis at a depth of 60% to 70% of the embedded depth. Figure 33 illustrates the optimal depth (D*) of the padeye, from which it is possible to realise that a taut wire does not require as deep a padeye as a catenary mooring.

Figure 33 – Mooring appliance point and variation of padeye depth with loading angle for given center of rotation (Randolph & House, 2002 )

Suction caissons have an identical installation process to buckets and concrete caissons. The four major installation steps are illustrated in Figure 34. Firstly, the caisson is delivered to the required location, and then it is initially penetrated into the seabed under self-weight with the top-vent opened. Afterwards, the remaining penetration is completed by pumping water from inside the caisson, using demountable pumps connected to a valve in the lid and operated by ROVs. When the desired skirt penetration is achieved, the pumps are stopped and the vent is closed to stop water flow into the interior. This allows internal suctions to develop under vertical loading (uplifting), hence maximising the end-bearing resistance (Randolph et al., 2005). 33


Figure 34 – Suction Caisson Installation Steps, from: http://www.epd.gov.hk/eia/register/report/eiareport/eia_1672009/HKOWF%20HTML%20EIA/ Main%20Text%20figures/Fig%202.36a.png.

4.4. VERTICALLY LOADED DRAG ANCHOR High capacity drag anchors evolved from conventional ship anchors. Traditionally, drag anchors comprise a broad fluke rigidly connected to a shank, as shown in Figure 35(a). The angle between shank and fluke is pre-determined, though may be adjusted prior to anchor placement on the seabed. This angle is typically around 50° for clay conditions and 30° in sand or where clay of high strength occurs at the seabed. For installation, the anchor is positioned on the seabed correctly orientated and it is then embedded by pre-tensioning the chain to an appropriate proof load. Depending on soil conditions, penetration depths usually range from 1 to 5 fluke lengths (typical fluke lengths being 1 to 8 m), and anchors can be dragged through a distance of 10 to 20 times the fluke length, typically a holding capacity of 20 to 50 times the anchor weight is mobilized.

34


Figure 35 – Drag anchors, (a) fixed fluke anchor. (b) Vryhof Stevmanta VLA. Source: Vryhof anchor manual (2005)

Vertically loaded drag anchors (VLA), also known as drag-in plate anchors, were developed to overcome the existing limitations on fixed-fluke anchors, which could not withstand significant vertical load components at the seabed. Actually, fixed-fluke anchors are removed by applying vertical load to the anchor chain. Therefore, common drag anchors cannot be used for deep-water foundations using taut or semi-taut polyester rope moorings. The VLA is similar to the conventional drag anchor except the fluke is a plate with a much slender profile and the shank is replaced by either a much thinner shank or a wire harness, Figure 35(b). It is installed like a conventional drag anchor with a horizontal chain load at the mudline and then different mechanisms are used to allow the fluke to rotate until it is perpendicular to the applied load [angle adjuster in Figure 35(b)], mobilizing the maximum possible soil resistance, and enabling the anchor to withstand both horizontal and vertical loading. There are two different installation methods; it can be either single line or double line installation. Figure 36 illustrates both of the installation methods. For each installation method a recovery method for the VLA is available, typically by detaching the front chains or wires and pulling the anchor in the opposite direction to the installation direction with a fraction of the installation load. Figure 37(b) shows the VLA being pulled out of the seabed after detaching the front mooring.

35


Figure 36 – Installation methods

Figure 37 – (a) Anchor depth determination. (b) VLA simple recovery, Vryhof anchor manual (2005).

VLAs achieve the desired effect of installing a plate anchor at a sufficient depth below the seabed in order to resist the mooring loads, but there is an inherent problem with these kinds of anchors in knowing exactly where they are in the soil. Usually, the depth is determined by measuring the angle of the anchor line makes with the seabed, and the length of wire in the soil, but for many reasons this angle may not correspond to reality, as it may have a differential angle along its embedded length. In Figure 37(a), a scheme to determine the approximate anchor depth is indicated. Initially, VLA utilization was predominantly for semi-permanent moorings, for example, for mobile offshore drilling units (MODUs). Thus, it can also be considered a temporary foundation. Nowadays, VLAs are commonly used to moor permanent floating facilities, such as a FPSO in the Campos Basin, Brazil, which is located in 1600 m of water with a taut-line mooring system secured to the seabed by nine 14 m² Vryhof Stevmanta VLAs.

36


4.5. SUCTION EMBEDDED PLATE ANCHOR A new system, called a suction embedded plate anchor (SEPLA), was developed to overcome the problems of the conventional plate anchor (e.g. VLA), achieving greater and more precise depth location below the seabed (Dove et al., 1998; Wilde et al., 2001). The SEPLA uses a suction caisson (or “follower”) to embed a rectangular plate anchor,

providing a known initial penetration depth for the anchor, at a specified geographical location. The components of a SEPLA are illustrated in Figure 38.

Figure 38 – Components of a suction embedded plate anchor (Gaudin et al., 2006)

SEPLA installation consists of 3 steps: caisson penetration, caisson retraction, and anchor keying. These steps are shown schematically in Figure 39. First, the caisson with a plate anchor slotted vertically in its base is lowered to the seafloor and penetrated into the soil under its dead weight until the skin friction and end-bearing resistance of the soil on the caisson equal the caisson’s dead weight. The vent valve on the top of caisson is then closed and the water trapped

inside is pumped out. The ensuing differential pressure at the top drives the caisson to the design depth. The plate anchor is then released and the water is pumped back into the caisson, causing the caisson to move upward, leaving the plate anchor in place in a vertical orientation. The caisson is retracted from the seabed and prepared for the next installation. As the anchor chain is tensioned, it cuts into the soil. Simultaneously, the anchor line applies a load to the anchor’s offset padeye causing it to rotate or “key”. In order to achieve the maximum mobilized capacity, the plate must be as close to perpendicular to the direction of loading as possible (Yang et al., 2012).

37


Figure 39 –Schematic of SEPLA installation (Yang et al., 2012; and courtesy of InterMoor).

SEPLA installation accuracy represents a great improvement over that for drag embedment anchors, however two questions emerge (these questions are applied to all offshore plate anchors such as VLAs). Firstly, the caisson penetration and anchor keying provokes a disturbance in the soil mass around the SEPLA, which leads to a decrease of the soil strength in the region. Secondly, when keying is being initiated, a loss of embedment depth occurs. While, the first question can be solved as the soil strength is largely recovered over time by soil reconsolidation, the second problem cannot because loss of embedment depth is permanent. This is a very important issue, since SEPLA capacity significantly depends on its embedment depth when the soil has increasing strength with depth (which is a typical in the off shore environment). Therefore, it becomes very important to accurately estimate the loss of embedment depth during the keying process. This estimate can then be factored into the design; Wilde et al. (2001) report upward movements ranging between 0.5 and 1.7 times the plate height, which is a wide range when plate heights of 2.5 m to 4.5m are used in practice. Even though the undrained capacity of plate anchors has been extensively investigated by means of analytical and experimental methods; for SEPLA, there are a limited number of reported studies and therefore the keying process is not yet well understood. However, Yang el al. (2012) present a theoretical model to predict the trajectory and corresponding capacities of SEPLA during the keying process based on plastic limit analysis.

4.6. DYNAMICALLY PENETRATED ANCHOR As offshore exploitation moves to water depths of around 3000 m, new technologies have had to be developed in order to reduce installation costs, and facilitate construction. Moreover, the high number of floating production and drilling units in operation may provoke the congestion of the sea bottom due to the high number of risers and mooring lines employed. In this scenario, dynamically penetrated anchors (DPA), and in particular Torpedo anchors, have proven to be a 38


reliable alternative used in Brazilian offshore fields (Aguiar et al., 2009). The reduced mooring line radius employed on torpedo anchors relative to catenary mooring systems with drag anchors, reduces sea bottom congestion, Figure 40.

Figure 40 – Radius comparison between floating units linked to conventional drag anchors and torpedo anchors, from: http://www.hindawi.com/journals/jam/2012/102618.fig.002.jpg.

Torpedo anchors (TA) are the most applied type of DPA and they have been developed by the Brazilian oil company Petrobras. TAs are cone-tipped, cylindrical steel pipes filled with concrete and scrap metal. They penetrate the seabed relying on the kinetic energy they acquire while free falling from heights of between 30 m and 150m above the seabed. Torpedo anchors come in various sizes from 0.76 m to 1.07 m in diameter, 12 m to 17 m in length, and 241 kN to 961 kN in weight. The inside of the anchor shaft is filled with ballast to increase the weight and maintain the centre of gravity below the centre of buoyancy for stability. Some versions of the TA have been fitted with 4 flukes at the trailing edge, ranging in width from 0.45 m to 0.9 m, and 9 m to 10m long (Raie, 2009; Medeiros et al.,1997, 2001, 2002). Two different DPAs, with and without fins are pictured in Figure 41(a). Torpedoes can easily reach velocities of 25 m/s to 35 m/s at the seabed after being released from a height of 20 m to 40 m above the seabed, allowing tip penetrations up to 3 times

39


the anchor length and holding capacities after consolidation that are expected to be in the range of 5 to 10 times the weight of the anchor (Randolph et al., 2005).

Figure 41 – Dynamically penetrating anchors (a) Torpedo anchor with fins and without fins (Medeiros, 2002); (b) installation of 4 flukes torpedo anchor (Medeiros, 2002 ; O’Loughlin et al., 2004)

The installation procedure for DPA has developed from its original method. Instead of using only one anchor-handling vessel (AHV) to lower the anchor to a predetermined height above the seabed, using the permanent mooring line, now two AHV are used. The installation process was modified to minimize the effect of drag force on the mooring line on the free falling motion of the anchor. Accordingly, the anchor is lowered using an installation wire from the first AHV while the second AHV holds the permanent mooring line that is attached to the anchor and for ms a loop. A remote release system is used at the end of installation wire to release the anchor (Araujo et al., 2004). A chain segment is recommended for the lower portion of the mooring line because model tests of the anchor installation (Lieng et al., 2000) have shown that chain drag does not reduce the velocity of the anchor during free fall. Figure 41(b) demonstrates the lowering of two model scale torpedo anchors to position them before free-fall releasing. A full scale torpedo pile and the situation immediately prior to TA release, in which it is possible to see the loop between the permanent mooring line and the installation line, is illustrated in Figure 42.

40


Figure 42 – Full scale torpedo pile and releasing situation, Lieng et al. (1999)

The main reason for using this type of mooring solution is its simplicity and speed of installation. With regard to the equipment required for installation, the torpedo anchor installation is depth-independent. Moreover, torpedo piles are cost-effective throughout fabrication, transportation, and installation. Fabrication is easy and inexpensive due to the simple design of the torpedo anchors. The cost of transportation is low because the compact design of the torpedo anchor allows more anchors to be transported per voyage of the AHV than, for example, suction caissons. Also, the installation is economical because an external source of energy is not required for installation and a quick installation is possible using one or two AHVs and limited use of ROVs. Finally, the predicted holding capacity is less dependent on the precise evaluation of the soil shear-strength profile. Since higher strength profiles reduce the penetration and lower strength profiles increase penetration, therefore the holding capacity is mainly a function of the kinetic energy obtained during free falling. Nevertheless, torpedo anchors have the disadvantage of the uncertainty in verticality of the anchor, which affects the holding capacity (O’Loughlin et al.,

2013; Raie, 2009). Figure 43 presents a table, where the advantages and disadvantages of the suction caisson, VLA, SEPLA and torpedo anchor systems discussed in this work are summarised.

41


Figure 43 – Advantages and disadvantages of different deep water anchor types (Ehlers et al., 2004).

42


5. GEOLOGICAL CHARACTERIZATION Before exploration, governments, energy companies and academic institutions study what seem to be the most likely areas to look for evidence of energy sources. When the objective is to find hydrocarbon resources, the geology of the area gives indications of the likelihood of the existence of hydrocarbon traps and other oil and gas sources. But if the objective is renewable energy then metocean data gives information of certain locations where the appropriate environmental conditions occur, e.g. areas where strong winds constantly blow will be appropriate to install wind turbines. If potentially commercial resources are found, the government will determine the boundaries of offshore exploration lots, and tender exploration licenses. In exploration for hydrocarbon sources, geophysical surveys of licensed areas are run to determine whether and where the oil or gas is likely to be. These surveys typically penetrate several kilometers into the seabed. For renewable energy, the focus is on wind, wave, current, or tidal characteristics. Once a resource has been found and government consents have been obtained, further survey work is carried out to determine engineering and other design conditions and parameters (Dean, 2009). Table 4 lists some of the surveys that may be done. Table 4 – Surveys and their purpose (Dean, 2009)

Survey

Bathymetry survey

Purpose

To measure water depths, map the seafloor, identify seafloor hazards such as unevenness, slopes, fluid expulsion features, and collapses To determine wind, wave, and current characteristics at the platform

Metocean study

sites; these will form the basis of estimates of platform loading and seabed scour

Environmental baseline survey

To identify environmental issues and measure environmental flora and fauna populations, so that the site can be returned to its original condition after the production operations have finished

Geohazards

To identify and plan the mitigation of geological and geotechnical

assessment

hazards

Shallow geophysical

To identify soil layering and sub-bottom hazards such as geological

survey

faults, in-filled ancient riverbeds, voids, shallow gas, and rock

Geotechnical survey

To verify soil layering and determine soil properties at the platform sites

43


Pipeline or cable route surveys

To identify bathymetry, soil conditions, and hazards along pipeline routes between different offshore platforms, and between platforms and onshore

Seismic risk

To identify the seismicity of the area and determine spectra and/or

assessment

acceleration time histories to be used in seismic design Done just before installing a platform at a site, to check that the seafloor

Seafloor survey

has not changed and hazards such as dropped objects, sand dunes, shipwrecks have not occurred there

In order to evaluate, which foundation solution is best for each site, the geological characterization must answer important questions such as: 

What is the water depth?



What is the composition of the mudline?



What soils do we find?



What are the layer thicknesses?



What geohazards do we find there?

5.1. TOPOGRAPHICAL FEATURES OF OCEAN FLOORS Figure 44 illustrates a typical marine topography of the ocean bottom, the features common to all oceans: the continental margin, the continental rise and the abyssal plain of the deep ocean. Trenches and seamounts (oceanic ridge or rise) may also be present in the deep ocean. The continental margin comprises the continental shelf and slope. The shelf is the submerged continuation of the adjacent land. The seaward extent of the continental shelf is the shelf break, or the continental ridge, leading into the continental slope. The toe of the continental slope is marked by the continental rise leading into the abyssal plain. The continental margins (covering approximately 20% of the total ocean floor) are extremely important as oil reservoirs and as such, are of most interest to engineers concerned with harnessing offshore oil and gas resources.

44


Figure 44 – Topographical features of ocean floors, after http://www.marinebio.net/marinescience/01intro/woimg/xsectopo.jpg.

While ocean topography can be described in general terms, regional variations naturally exist. Figure 46 shows a shaded relief map of land topography and ocean bathymetry compiled by the United States National Geophysical Data Centre, it also shows the amplification of Southern Africa, where is possible to distinguish the continental shelf and slope, as well as the abyssal plain. In addition, Figure 45 shows the topographical profile of the continental margin in a selection of offshore locations, illustrating the potential diversity of shelf and slope bathymetry. These locations are indicated on the world map in Figure 46.

Figure 45 – Profiles of ocean floor topography at selected locations, Randolph et al. 2011

45


Figure 46 – Bathymetric shaded relief map showing a detailed extent of continental margins (number in red relates to locations in Figure 45), US National Geophysical Data Centre

46


The continental shelf can be virtually non-existent or extend several hundred kilometers offshore with the shelf break occurring between depths of 10 m and 500 m. Shelf breaks are deepest off glaciated areas and shallowest in areas with extensive coral growth. The average slope of the continental shelf is approximately 1:500 (0 ˚07’). Beyond the shelf break, the continental slope has a steeper gradient than the shelf, ranging from an average 1:40 (1.2˚) in delta regions but up to 1:10 (6˚) in faulted areas, and r eaches water depths of 2000 m to 3000 m. The

continental rise, at the foot of the continental slope, has gradients ranging from 1:1000 and 1:700. The abyssal plain adjacent to the continental rise is smooth with gradients between 1:1000 and 1:10000, and occurs from depths of 2500 m to 6000 m.

5.2. SEABED GEOLOGY In the majority of the ocean bottom, the first layers are thick compositions of marine sediments; Figure 47 shows a digital model compiled by the U.S. National Geophysical Data Centre showing the distribution of sediment in the world’s oceans and marginal seas. These sediments

are clearly thicker near continents, and thinner on newly formed mid-ocean ridges. In some areas, strong bottom currents are responsible for cleaning away any sediment, avoiding its settlement and consolidation. Even though, the continental margins cover only 20% of the seabed area, they contain nearly 75% of the total marine sediments. In many places, the continental rise is a depositional feature, formed mainly of sediment slurry, and reaching up to 1.6 km thickness. Canyons often cut across the rise and act as channels for the seaward transport of sediment. Abyssal plains are connected by canyons or other channels to landward sources of sediments, which are transported as dense slurries to the plains (Poulos, 1988; Randolp h et al., 2011).

47


Figure 47 – Sediment thickness of the world’s oceans and m arginal seas . Source: http://www.ngdc.noaa.gov/mgglsedthick/sedthick.html.

5.2.1. S EABED SEDIMENTS ORIGIN AND CLASSIFICATION Marine sediments are composed of detrital material either from land or the remains of marine organisms which, now, leads to the principal classification of sediment as terrigenous (transported from land) or pelagic (sediments that settle through the water column). Pelagic sediments are deposited so slowly that near shore and coastal areas are overwhelmed by terrigenous deposits. On one hand, terrigenous material comes from sediments carried by the flow of rivers, coastal erosion, aeolian or glacial activities. The size of the soil particles is used as the basis for describing terrigenous sediments, and then sub-categorizing them. These sediments tend to be grains of silicate-based minerals such as quartz and feldspar, and are principally formed from the erosion of rocks – leading to the term lithogenous. On the other hand, pelagic sediments are generally fine grained and are instead classified according to their composition. Organic, or biogenous, pelagic sediments derive from the insoluble remains of marine organisms, e.g. shells, skeletons and teeth. Lithogenous pelagic sediments are formed when particles are transported by wind into the ocean, before settling through the water column. Figure 48 illustrates a schematic cross-sectional view of the continental margin which shows various sediment types and their distribution across an idealized 48


passive margin. Passive margins are the margins which not extend down to subduction zones (active continental margins). Marine sediments can also form from biological and chemical reactions occurring in the water column or within sediments, and are referred to as hydrogenous sediments. Figure 49 summarises some of the processes involved in the formation of marine sediments.

Figure 48 – Distribution of sediment across a passive continental margin, from: http://classconnection.s3.amazonaws.com/927/flashcards/68927/jpg/4-91305062763712.jpg.

Figure 49 - Sedimentary process of marine deposits (after Silva 1974)

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Lithogenous Sediments

The number one suppliers of oceanic sediments are rivers, as they contribute 20 billion tonnes of sediment to the world’s oceans every year. The settlement of particles is controlled primarily by grain size, such that generally finer particles are found with increasing distance from shore, since these particles are lighter they remain suspended in the air and water longer and travel further than the heaviest ones. However, lithogenous sediments can also come from material transported during submarine slides such as debris flows and turbidity currents (heavy fluid flows). The latter, in particular, can transport sediment from the continental margins to the continental rise and onto the abyssal plain, Figure 50. Deposits formed by turbidity currents have graded bedding, with larger particles overlain by progressively smaller particles, Figure 50(c) shows this phenomena. Wind also plays an important role, transporting around 100 million tonnes of sediment into world’s oceans annually, and small particles can be carried for considerable distances before

being deposited in the sea as pelagic sediments (Keuen et al, 1950; Randolph et al., 2011).

Figure 50 – Turbidity current: (a) seabed topography where slope break may start a turbidity current

(b)

turbidity

current

progress

(c)

http://people.maths.ox.ac.uk/fay/research.html .

50

settlement

turbidity

particles.

From:


Biogenous Sediments

Biogenous sediments may either be siliceous or calcareous and are formed from planktonic organisms, which are the most abundant organism in the oceans (now and through geological history). The insoluble shell of these creatures can be formed of silica or calcite. While, silica organisms are more common in the Polar Regions or in the very deep waters of the equator, calcite organisms are more common in shallow water and temperate tropical climates. Calcium carbonate, the main constituent of calcareous deposits, is soluble at high pressure, that’s why they are not found in water depths greater than 4000 m (Randolph et al., 2011). Figure 51 shows the worldwide distribution of neritic (nearshore) and pelagic (open ocean) sediments, it is conclusive that neritic deposits are dominated by lithogeneous materials while pelagic deposits are dominated by various types of biogenous oozes and lithogeneous abyssal clay.

Figure 51 – Distribution of the different sediments deposits across the world, from: http://classconnection.s3.amazonaws.com/927/flashcards/68927/jpg/4-91305062763712.jpg.

All these pelagic sediments become from remains of ancient micro-organisms, such as shells or skeletons (typically from an organism called foraminifera), and they are often perforated and angular. These remains when coupled with the relative softness of calcium carbonate (of which they are made), leads to the fragility and high compressibility which are characteristic of calcareous sediments. Figure 52 compares micrographs of calcareous sand, from Goodwyn on the North-West shelf of Australia, with quartz-grained silica sand, showing clearly the difference in particle shape. Calcareous sand has fragile, angular and hollow particles as opposed to the hard, rounded silica sand grains.

51


Figure 52 – Micrographs of (a) calcareous sand and (b) silica sand (Dean, 2009).

5.2.2. GEOTECHNICAL CHARACTERISTICS OF SOME OFFSHORE REGIONS As a consequence of what was explained before, it is possible to say, in broad terms, that finer-grained sediments predominate further from shore and in deeper water. Because coarser terrigenous sediments cannot be transported this far, therefore pelagic sediments predominate. Certain geohazards are also more prevalent in deep water, including steep scarps that are often found at the margins of the continental shelf. Even though, many factors may affect the characterization of a specific area, in broad terms, some common seabed conditions in the major areas of oil and gas exploration can be characterized as (on the following list, the numbers at the end of each description refer to the sites on Figure 53), (Dean, 2009): 

Gulf of Mexico – soft, normally consolidated, medium high plasticity clays (30


Campos and Santos Basins offshore Brazil – sands and clays with high carbonate content (site 5).



Western Africa – soft, normally consolidated, very high plasticity clays (70


North Sea and other glaciated regions – stiff, overconsolidated clays and dense sands, with a recent drape of softer material (site 14).



South-East Asia – desiccated crusts of stiff soil, which are remnants of low sea levels during the Pleistocene (2,588,000 to 11,700 years ago), with strengths one or two orders of magnitude greater than the underlying soil (site 20). 52




Australia’s North-West and Timor Sea- carbonate sands, silts and clay-sized soils, often

with variable cementation (site 13).

Figure 53 – Worldwide distribution of offshore oil and gas developments. Dean (2009) based on McClelland (1974) and Poulos (1988).

5.3. GEOHAZARDS By definition, a geohazard is a geological and environmental condition, and involve longterm or short-term processes, that can lead to the movement of soil, rock, fluid or gas during sudden episodic events or slow progressive deformations (Randolph et al., 2011). For offshore oil and gas projects, geohazards have the potential to cause injury or loss of life, damage to the environment or infrastructure, and can impose significant additional project costs. Diverse geohazards are associated with engineering on the ocean floor and they become significantly more dangerous in deeper water. Geohazards are grouped in two general categories: 

Hazardous events – events that are infrequent and episodic in nature, such as phenomena associated with earthquakes, submarine slope movements, turbidity flows and gas expulsions.



Hazardous ground conditions – conditions that involve slow processes that are progressive in nature, such as soil creep, non-tectonic fault creep and mud or salt tectonics. 53


Geohazards may pose a threat to the integrity or serviceability of structures and their foundations over their design lifetime. Therefore, with the purpose of their identification on the site; studies of geology, geomorphology, and geography of a region, and t horough geophysical and geotechnical surveys and investigations are executed (Dean, 2009). Figure 54 illustrates the most typical geohazards to be considered, including many geological features as well as landslides, carbonate sands, unconsolidated soils, gas hydrates, and disturbed sediments. Submarine slope instabilities are discussed further in the following section, since these are the common geohazard that needs to be dealt with.

Figure 54 – Schematic diagrams showing main offshore geohazards (a) widely used summary of deepwater geohazards (Power et al., 2005) (b) more geohazards (Strout and Tjelta, 2007).

54


5.3.1. T RIGGERING M ECHANISMS FOR S UBMARINE S LOPE F AILURES There are a several prospective events capable of triggering submarine slope instabilities, flow slides, debris flows, and mudflows and such events can be either sudden, or progressive. Natural processes or anthropogenic actions cause an increment in soil stresses or decrease in soil strength, leading to failure of a soil mass. The comprehension of pore pressure conditions and the processes and mechanisms that lead to excess pore pressure generation are very important for assessing potential geohazard triggering mechanisms (Randolph et al., 2011). Kvalsta et al. (2001) identified a selection of triggering mechanisms (some events are illustrated in Figure 54b): A) Natural Processes 

Rapid deposition leading to excess pore pressures, under-consolidation and increased shear stresses in a slope



Base erosion or top deposition leading to higher slope inclination and increased gravity forces and shear stresses along potential failure surfaces



Melting of gas hydrates caused by temperature increase or pressure reduction leading to increased pore pressure and reduced soil strength



Active fluid or gas flow and expulsion



Mud volcano eruptions giving rise to mass wasting and soil displacements



Tectonic fault displacements generating earthquakes, near-field displacement pulses, and ground rupture



Earthquake strong ground shaking causing short-term inertia forces and increase in pore pressure



Long wavelength wave loading



Sea level lowering during glacial periods leading to lower hydrostatic pressure, free gas expansion and gas hydrate ex-solution



Increased seawater temperature at seabed level caused by changes in current regime leading to temperature increase in the soil mass and ex-solution of hydrates.



Sensitive (contractive) and collapsible soils increasing the risk of retrogressive sliding and increased areal extent of failure zones.

B) Human Activities 

Drilling wells, creating blowouts and cratering at the seabed



Underground blowouts changing the pore pressure regime in shallow layers and potentially creating instability in sloping areas

55




Oil production increasing heat flow and temperature around wells leading to hydrate ex-solution, increased pore pressures and strength loss of the adjacent soil



Depletion of reservoir pressure giving rise to reservoir collapse and changes in overburden stresses



Installation activities leading to increasing gravity forces



Mooring installations and anchoring forces imposing short and long-term lateral forces.

Marine geohazards generally involve mobilisation of the seabed and sub-bottom sediments. This mobilisation may impact against or bury infrastructure or lead to loss of foundation support. The volume of soil involved may range from a few cubic metres to thousands of cubic kilometres with consequences ranging from local over-stressing of subsea infrastructure to total loss of an installation with the associated human casualties and economic and environmental impacts. If a very large seabed displacement occurs it has the p otential to generate a tsunami, which can cause massive human, economic and environmental losses. Therefore platforms are designed to avoid generating a slope failure, and to resist the forces from turbidity currents and debris flows generated elsewhere (Dean, 2009; Randolph et al., 2011). 5.3.2. GEOHAZARD IDENTIFICATION Due to the large and complex nature of offshore projects, they often span a range of different environments, extending from deep water on the continental rise, up the continental slope, across the continental shelf through shallow water to the shoreline. Therefore, major projects are exposed to a wide range of geological and geotechnical conditions, so these projects evolve through time and may go through many design concepts and engineering stages. Projects involve three phases, moving from the general to the specific; Figure 55 summarises these phases. First, pre-exploration interpretations of potential site conditions are carried out, then post-discovery preliminary engineering evaluations are made, and finally, an integrated site characterisation is performed in order to support detailed design. The geological and geotechnical information collected during the development of projects needs to be sufficient in order to support each stage in a project. As a result, a phased approach often proves most effective in meeting engineering requirements for offshore developments (Randolph et al., 2011).

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First Phase. Pre-Drilling Activities

Second Phase. PostDiscovery

Third Phase. Integrated Site Characterisation

Screening of regional geological hazards

Preliminary engineering evaluation

Seismic inversion and development of final geotechnical criteria

Assess Geohazards in prospect area

Plan high-resolution geophysical survey programme

Detailed geohazard assessment, analyses for special engineering issues

Assess hazards for specific well sites Team meetings and reporting

Carry out high-resolution geophysical survey programme Prepare and process high-resolution geophysical data Complete preliminary site characterisation Plan geotechnical site investigation

Risk assessment Develop model with integrated site characteristics Prepare integrated report Team meetings and reporting

Carry out geotechnical investigation Sample preparation and shipping to laboratories

Geological lab testing

Geotechnical lab testing Team meetings and reporting

Figure 55 – Summary of generalised investigation elements (Campbell et al. 2008)

The Phase 1 works essentially as a desktop study, to evaluate general constraints from conditions such as extreme terrain, earthquake and fault activity, slope instability and broad geotechnical soil properties. Generally this first stage involves: compilation and review of published or unpublished data and reports, interpretation of exploration seismic testing results; development of a project geographic information system (GIS), identification of critical engineering issues and engineering support. The main product resulting from this initial stage of study is a regional scale geohazard map, which identifies landslides, fault crossings, areas of liquefaction potential, salt domes, mud volcano activity and areas of gas hydrates. However, these results retain large uncertainties and are not suitable for detailed design (Randolph et al., 2011). The next phase of work (Phase 2) occurs after a discovery has been made and builds upon the first phase baseline geohazards assessment to define geohazard issues that may affect specific components of a proposed system. The Phase 2 investigations will involve planning and execution

57


of detailed geophysical and geotechnical data acquisition programmes as well as preliminary site characterisation activities. Phase 2 investigations may include the following: 

Acquiring high resolution geophysical and geotechnical data sets



Developing a predictive soil model



Developing the project GIS further and mapping geological conditions in the foundation zone or along route alignments



Conducting detailed terrain analysis, and interpretation of high-resolution geophysical data to identify specific hazards within the project area



Constructing preliminary hazard susceptibility maps showing, for example, rugged terrain, faults, landslides, liquefiable terrain, and submarine canyon crossings



Identifying specific targets requiring investigation to develop final design parameters



Interacting with the engineering team to discuss geohazard constraints and impacts on design



Developing recommendations for special studies required to address specific technical issues.

After these works are done, geohazard and design teams need to get together and review the specific findings and determine whether additional investigations need to be completed (Randolph et al., 2011). Lastly, Phase 3 activities involve the final detailed integration of all of the various data sets. This is an intensive stage of many projects and involves close interaction between geologists, geotechnical engineers and owner’s representatives. During this stage, additional geophysical

processing may be carried out (e.g. seismic inversion), final sub-surface soil models are developed, and soil parameters are defined for use in specialty studies such as site amplification, liquefaction and slope stability analyses. Detailed geohazard assessments are also executed to address the distribution, severity and frequency of geohazards such as submarine slope failures, mass gravity flows, faulting, strong ground shaking, liquefaction, scour, gas hydrates and fluid expulsion. There are two general types of data acquisition tools required to complete deep-water developments. These include geophysical survey tools and geotechnical/geological sampling and in situ testing tools. The latter will be discussed in the following section.

58


5.3.3. GEOTECHNICAL SITE INVESTIGATION A geotechnical site investigation programme is most effectively carried out following acquisition of site-specific geotechnical data, development of the preliminary soil model and completion of the initial geohazard assessment. This enables the collection of targeted data and samples to support the foundation engineering stage of a project, but also provides an opportunity to acquire data for the assessment of specific seafloor features. The geotechnical investigation provides data for specific soil properties and pore pressure conditions of seabed deposits, involves borehole logging, field testing, sampling, and both geological and geotechnical laboratory testing. There are two main approaches for geotechnical site investigations: seabed mode and down-hole mode (Campbell et al., 2008). Seabed mode methods characterise the shallow part of the stratigraphic profile, which means depths less than 50 meters. These methods often include large diameter piston core sampling and in situ testing, piezocone penetrometer testing (PCPT), vane shear tests (VST) and T-bar and ball penetrometer testing (Peuchen and Rapp, 2007). Yet, some facilities, such as TLPs, compliant platforms and mooring systems for SPARs, semi submersibles and FPSOs vessels have driven pile foundations that require data on soils to a greater depth than can be reached using seabed investigations approaches. In these cases, sampling is completed using down-hole approaches. This involves using rotary d rilling techniques from a ship-based drilling platform. Sample types may include: piston samplers, push tubes, rotary corers, percussion corers and pressurised corers to sample gas hydrates. It is also important to execute field tests to identify pore pressure conditions, temperature distributions and remoulded and residual shear strength of soft sediments. High-quality samp ling techniques should be used to retrieve cores for laboratory testing. Figure 56 enumerates the tests that a testing programme should include (Randolph et al., 2011).

59


Soil texture and mineralogy Clay mineralogy Classification tests to identify

Age Pore water salinity Thermal properties

Laboratorial testing programme

Peak strength Critical state

Strength tests to identify

Remoulded and residual properties Stress tests to investigate effects

Stress dependency Strain rate Strength anisotropy Cyclic loading

Figure 56 – Tests that should be performed during a laboratorial testing program (Randolph et al., 2011).

5.3.4. S UBMARINE SLOPE FAILURES AND SLIDES Seafloor instability is the main geohazard threat encountered offshore and can have catastrophic effects on offshore developments. Seabed instability is an issue even on the continental shelf with seafloor gradients as low as 0.5 ᵒ. Submarine slides can range in size from relatively small coastal slides of less than a cubic kilometre to vast slides involving thousands of cubic kilometres of material. The mobility of submarine landslides can be characterised geometrically by the run-out ratio L/H where L is the horizontal distance from source to deposit and H is t he vertical elevation of the debris flow source above the deposit, Figure 57. First introduced by Heim (1932) and used later by Scheidegger (1973), the ratio can be predicted by considering the energy balance for a dry mass sliding down a slope. Figure 58 shows the volume of material involved and run-out ratio of various submarine and sub-aerial slides. From the compiled data, it appears that submarine slides can be much larger than sub-aerial landslides (due to the concentration of each group of dots), and they also tend to exhibit larger run-out distances for the same volume of sliding material. This indicates that water plays a particular role in the mobility of the sliding mass. 60


A submarine slide involves an initiation event, usually a shallow or deep-seated slope failure, often with retrogressive slumping, followed by mass gravity flow involving laminar viscoplastic debris flow, and loose suspension turbidity currents. The slope failure and mass gravity flow must be analysed as part of a geohazard assessment (Randolph et al., 2011).

Figure 57 – Definition of slide mobility

Figure 58 – Comparison of volume and run-out distance of submarine and sub-aerial slides (Randolph et al. 2011 after Scheidegger 1973, Edgers and Karlsrud 1982, Hampton et al. 1976, Dade and Huppert 1998, De Blasion et al. 2006)

61


62


6. CASE STUDY: GEOTECHNICAL CONSIDERATIONS FOR ULTRA DEEP OIL FIELDS OFF SÃO TOME E PRINCIPE 6.1. PROPOSED DEVELOPMENT In the past decade São Tomé e Princípe (STP) has been reported as future potential oil supplier to the world, yet little progress has been made, largely because potential exploitation sites have water columns ranging from 1800 to 3000 meters. With advances in deep sea production technology and given that the most readily exploited production fields worldwide are now being exploited, STP has once again come under greater attention. As a member of the Community of Portuguese Speaking Countries (CPLP), STP has been used to provide a focus for the development of this thesis, and this work provides a preliminary review on an ultra-deep water design proposal for its offshore sites. The foundation solution that is presented takes into account the economic issues, technical novelty and site conditions specific to STP. The foundation proposed an anchor type system to which a floating platform is moored. Possible anchor solutions might include suction caissons, torpedo anchors, and/or plate anchors; specifically, solutions using a Torpedo Anchor and a SEPLA will be evaluated.

6.2. GEOTECHNICAL SITE CONDITIONS São Tomé & Principe is a group of islands situated on the Gulf of Guinea, the island of Principe is the nearest to the site where possible oil exploration is more likely. Figure 59 shows the location of STP and the surrounding geology, it is also possible to see two red lines which refer to the schematic cross sections presented in Figure 60 and Figure 61; the cross sections extend from Principe Island to Nigeria, and Equatorial Guinea respectively. Based on the information obtained from Figure 60 and Figure 61, the seabed soil is composed of shales with a lithogenous nature, since they come from turbidity currents along the continental slope of Equatorial Guinea and Nigeria. In the direction of Nigeria, operating water depths range from 2600 to 3000 meters while towards Equatorial Guinea, they are 1800 to 2200 metres. Unfortunately there is not a precise characterization of the ultra-deep seabed soil offshore of Principe, though some investigation has been made in other parts of the Gulf of Guinea (GoG). Starting from these studies and extrapolating some results, a pre-design characterization can be carried out. Figure 62 illustrates the limits of GoG and the nations that border it.

63


Figure 59 – São Tomé & Principe location and surrounding geology, Courtesy of Agencia Nacional do Petroleo of STP from: http://www.stp-eez.com/DownLoads/Posters/3_STP_RegionalGeol.pdf .

Figure 60 – Cross section from Nigeria cost to Principe Island, Courtesy of Agencia Nacional do Petroleo of STP from: http://www.stp-eez.com/DownLoads/Posters/3_STP_RegionalGeol.pdf .

64


Figure 61 – Cross section from Principe Island to Equatorial Guinea Cost, Courtesy of Agencia Nacional

do

Petroleo

of

STP

from:

eez.com/DownLoads/Posters/3_STP_RegionalGeol.pdf .

Figure 62 – The red circle limits the extent of Gulf of Guinea.

65

http://www.stp-


6.2.1. I NDEX P ROPERTIES OF GOG S EDIMENTS The sediments present in GoG deep-waters are characterised by very high water contents, values which can be between 150% and 250% at the seabed and gradually decrease with depth. Figure 63 shows three different soil properties (water content, wet unit weight and submerged unit weight) from 10 sites in the GoG. Figure 63(b) suggests soil unit weights starting from 12-13 kN/m3 at seabed and increasing to 13-15 kN/m 3 below 6-8m, Puech (2004). Water content and wet unit weight results are also supported by the studies carried out by the International Ocean Discovery Program (IODP) in the GoG approximately 200 km off of the Ivory Coast, Figure 64 and Figure 65. However, the submerged unit weight graph suggests higher values (approx. 1.5 to 2 times) than Puech (2004) proposes for the first 20 meters. The locations of the IODP boreholes are identified in Figure 66. The soils present in the Gulf of Guinea have plasticity indexes (PI) ranging from 70% to 120%, and as high as 150% near the seabed, Figure 67. Liquid limits (wL) range between 125% and 175%, and when plotted in the Casagrande diagram (with its associated PI), Figure 68; reveals that the soils are classified as highly plastic clays (CH) and highly plastic silts (MH). Using the above liquid limit values and Figure 69, it is possible to determine that the coefficient of consolidation (cv) of these soils is likely to be somewhere between 10-8 and 0.5 10-8. When thoroughly analysed using X-Ray diffraction, the soil is found to have a high percentage of clay (60% to 80%) (Puech, 2004). The majority of the clay is Kaolinite (more than 50%) but an important proportion of smectites is also found (15 % to 20%), which explains the high plasticity of the soils (Puech, 2004).

Figure 63 – Deep-water sediments physical properties on Gulf of Guinea (Puech, 2004).

66


Figure 64 – Deep-water sediments physical properties, data from International Ocean Discovery Program testing results, after Mascle et al. (1998). Sites are located offshore of Ivory Coast and with depths ranging from 2090 to 4637 meters, Figure 66.

Site 959A Site 959B Site 960C Site 962B Site 959A Site 959B Site 960C Site 962B

Figure 65 – Comparison of the IODP testing results (offshore of Ivory Coast) and Puech (2004) sediment physical properties in GoG.

67


Figure 66 – Map showing the drilling locations of IODP in Gulf of Guinea, from: http://wwwodp.tamu.edu/publications/159_IR/images/map.jpg .

Figure 67 - Plasticity index of GoG sediments in depth (Puech, 2004).

68


Figure 68 – Plasticity chart of GoG deep-water sediments, Casagrande diagram (Puech, 2004).

Figure 69 – cv versus wL chart (Kulhawy & Mayne, 1990)

6.2.2. I N S ITU S TRESSES AND S TRESS H ISTORY On the continental slope of the GoG, the upper part of the sediment column is a consequence of sedimentation processes. When sedimentation progresses slowly, normally consolidated soils, where the submerged weight of the particles is entirely supported by the soil skeleton (no excess pore pressure), are more likely to be formed. An over-consolidated soil is one that the effective stress curr ently applied (p’0) is lower than the maximum effective stress applied previously (p’c). The ratio between these two (p’ c/p’0)

69


is called the over-consolidation ratio (OCR). The OCR is commonly obtained from standard oedometer test procedures. In situ vertical effective stress profiles from the GoG, as well as from Gulf of Mexico are presented in Figure 70. Whereas in the first few meters it is common to find OCR values around 2 in the GoG, the OCR decreases to values around 1.5 at depths of 15 to 20 meters. Below that depth, the OCR decreases further, reaching values around 1.2 and 1.3. This profile differs a lot from Gulf of Mexico, since in the GoM it is common to have an OCR of about 1 until 15 m depth, and a small increment with depth afterwards (Puech, 2004). Whereas, the difference between p’c and p’0 is almost constant all the way down in the GoG, in the GoM this difference gets greater with penetration.

Figure 70 – Typical

in situ

stress profiles for Gulf of Guinea deep-water soils and the comparison

with Gulf of Mexico (Puech, 2004).

The latest work that has been done in this field suggests that using the term yield stress ratio (YSR) rather than OCR to define the stress history of the Gulf of Guinea soils is more appropriate, since no implicit assumption regarding the past over-burden pressure is introduced 70


(J.L. Colliat & H. Dendani et al., 2011). The YSR is determined from the ratio between vertical yield stress σ’vy to the actual overburden pressure σ’ v0 (YSR= σ’vy/ σ’v0).

J.L. Colliat & H. Dendani et al. (2011) derived the vertical yield pressures from a number of standard oedometer tests performed on samples from several typical deep-water sites. They plotted the results versus depth, Figure 71, and used them to compute the YSR as well. Once again, the difference between σ’vy and σ’v0 is quasi constant with depth. Furthermore, they report this

difference to be typically between 15 kPa and 40 kPa for most sediments. The corresponding YSR values start at about 3 in the first couple of meters, and tend t o reduce to between 1 and 2, below after about 10 m depth. Even though the YSR exceeds 1, it does not imply over-consolidation in the geological sense (i.e. no past overloading of the material). De Gennaro et al. (2005) have shown that these soils exhibit a significant structuration effect, and that the difference between σ’vy and σ’v0 is a quantitative measurement of the “extra-strength” due to soil structure.

Figure 71 – Vertical yield pressure and YSR versus penetration depth for typical GoG sites (J.L. Colliat & H. Dendani et al., 2011).

Within the first 2 m below the seabed, higher values of YSR can be found. This occurrence is called “crust”. The origin of this “crust” is still undefined and open to debate. It is known from

geological evidence that these sediments have never seen overburden stresses in excess of the present state, meaning that the relatively high shear strengths of the “crust” are not the result of 71


mechanical over-consolidation. Based on a study by Brausse (2001), there are three families of mechanisms that might be a possible cause for this phenomena: 

A strong inter-particle physiochemical bonding and ion exchange which may be combined with effects of cementation and cause higher attraction between particles (through ionic and van der Waals forces). This proposition is supported by Sultan et al. (2001) who propose an interpretation of the apparent over-consolidation using a microscopic model based on the theory of the diffuse double layer, which relates the mechanical behavior of the soil to the variations in ionic concentration.



An inorganic cementation involving cementing agents such as silicates, carbonates, iron oxides, or alumina.



Bioturbation (biological activity leading to reworking of soils and sediments) and organic cementation. Ehlers et al. (2005) have presented X-ray radiographs of surficial cores from offshore Nigeria showing a correlation between the most intensively burrowed sections and the highest soil shear strengths, thus favouring the role of biological activity.

Currently, there is research work being done on this issue, however further effort is needed to better understand the origin of the “crust”, the conditions for its development and the reasons

for its absence in some areas. 6.2.3. S HEAR S TRENGTH P ROFILES Shear strength profiles are a determinant part of the final design solution, as well as when choosing which structure it is more appropriate to adopt. Usually these profiles are established from in situ CPTU and VST tests (uncorrected) in the precise location where the foundation will be positioned, but in the STP case they have not yet been done. Thus, information was gathered from multiple sites across the GoG (data collected from the sites in Figure 66by the International Ocean Discovery Program), along with data from other similar places, such as: Brazil, Gulf of Mexico and Mississippi Delta. IODP performed several soil tests in four different sites, from the location map Figure 66 & Figure 73 below, and where it is possible to see that the sites were located in different geotechnical environments. This implies different undrained shear strength profiles; for instance, Figure 72 suggests that site 960 has very low undrained shear strength in the first 30 meters, while sites 959 and 962 have a constant growth throughout their depth. T his difference is justified by the sedimentary activity in the continental shelf; while sites 959 and 962 are in a relatively stable location, site 960 is at the edge of a slope failure where the soil has been disturbed recently and has not reconsolidated yet. This reveals how important is to choose the location before

72


installing any foundation, since, despite the proximity between the sites (note 10’ is approx. 20 km), the resistance achieved would be very different.

Undrained Shear Strength Su (kPa) -

50

100

150

200

250

0 20 40 60

Site 962B Site 961A Site 960A Site 960C Site 959A Site 959B

) 80 m ( h t 100 p e D120

140 160 180 200

Figure 72 - Undrained shear strength variation with depth in Gulf of Guinea, determined by International Ocean Discovery Program using Vane shear and penetrometer tests, and test locations.

In his analysis, Puech (2004) recognised that GoG soils have: 

Typical gradient in undrained shear strength that is often close to 1.5 kPa/m.



Su/p’o ratios are high in comparison with onshore clays, where the values are about 0.25, here they exceed 1 in the first meters of penetration and decrease to about 0.4-0.5 below 10 m.



GoG clays are typically medium sensitive with sensitivity values ranging from 3 to 4, but can reach higher values. Sensitivity is the ratio between the undisturbed and remoulded undrained shear strength of the soils, and varies from about 1 for heavily overconsolidated clays to values of over 100 for the so-called extra-sensitive or “quick” clays.



Steel-soil interface adhesion factor (α) in these soft plastic clays is considered to be between 0.8 and 1 (Tomlinson, 1957). Puech (2004) also observed that the first 2 m of penetration can have two distinct types of profiles (Figure 73): 73




A “no peak” profile where shear strength starts from 1-2 kPa near the seabed and increases linearly with depth (magenta line)



A “peak” profile which is charac terised by a sharp increase of s u, reaching 8.5-14.3 kPa at 0.5m penetration, then decreases to progressively get back to the deep gradient at about 2 m penetration (red line). This initial “peak” is what is called as “crust”. This profile is more common in greater water depths, i.e. greater than 1000 m. Figure 74 illustrates four different profiles of (undrained) shear strenght, τ: two normally

consolidated clays from the Gulf of Mexico (GoM) (Quiros et al., 2003), a Mississippi Delta clay (Quiros et al., 2003), and the GoG 1.5 kPa/m profile proposed by Puech (2004). The various symbols represent the results obtained by the IODP in field tests executed in the Brazil Basin, it is reasonable to say that these test results have an acceptable fit to the 1 kPa/m profile of GoM. From Figure 74 is comprehensible that the shear strength gradient in the Gulf of Mexico has a range of about +/- 50-60% from the characteristic 1.25 kPa/m (red line, Figure 74). This gradient obviously depends on the location; the closer it is to the Mississippi Delta, the lower the gradient will be because of the soft under-consolidated clay layers. When the location is far from the influence of the river the layers are much stiffer and the shear strength gradient is about 2kPa/m. In the Mississippi Delta profile, below 75 m depth the gradient is similar to the stronger GoM strength profiles (dashed line).

Figure 73 – Typical cone resistance profile for GoG deep water sediments (WD – water depth), Puech (2004).

74


GoG, Puech 1.5 kPa/m Miss Delta, Quiros (2003) GoM NC, Queiros 1.25 kPa/m (2003) GoM NC, Quieros (2003) Brazil Basin, DSDP (1977)

Figure 74 – Undrained shear strength profiles of Gulf of Mexico, Brazil and GoG proposed by Puech (2004).

Figure 75 reveals that the majority of the shear strengths measured near the Ivory Coast (sites 959, 960, 961 and 962) lie below the 1.5 kPa/m gradient line. However, this does not mean that this gradient is not typical for the GoG. This soil, like those in the GoM may have a range of about 50-60% in its gradient and these measured values may be in the lower range. Assuming this similarity to the GoM, the gradient of undrained shear strength in the GoG may range between 0.75 kPa/m and 2.25 kPa/m.

75


Site 962B Site 961A Site 960A Site 960C Site 959A

Figure 75 – Shear strength profiles of diverse marine sites

Sultan et al. (2007) reported some studies from the continental slope of Nigeria. Even though the water depth where this study was executed ranged between only 1100 m and 1250 m, the study area was actually very close to Principe Island as shown in Figure 76. The shear strength profile based on laboratory geotechnical tests is represented in Figure 77. This profile has a similar development to the Puech (2004) proposal of a gradient of 1.5 kPa/m, therefore in future calculations this is the gradient that will be used. However, the shear strength profile does not start from the zero, therefore it is assumed to be 5 kPa in the first 3 m, and after that assumes the proposed 1.5 kPa/m gradient (blue line Figure 77).

Príncipe Island Figure 76 –Nigerian continental slope study area, Sultan et al. (2007).

76


0

GoG, Puech 1.5 kPa/m GoM NC, Queiros

2

GoM NC, Quieros

4 ) m ( h t p e 6 D

8

10

12

Figure 77 – Shear strength profile of Nigerian continental slope obtained in laboratory geotechnical tests and its comparison with shear strength profiles proposed by others.

6.3. DESIGN OF ANCHOR SOLUTIONS Since the water depths in the zone of proposed offshore development near São Tomé & Príncipe range from 1800 m to 3000 m, future installed facilities must be floating platforms and hence the foundation systems in the seabed will be resisting tensile forces instead of compression. Therefore, the only types of foundation solution suitable for this region are anchoring systems. Anchoring systems that are selected for evaluation in this case study are the ones for which fewer studies and investigation has been made, but on the other hand are likely to be the most economic systems. 6.3.1. T ORPEDO ANCHORS The design process for Torpedo anchors, Figure 78, is rather complex due to the difficulties of predicting the anchor embedment and set-up after installation. These two factors have a direct effect on the anchor capacity and they depend on the geometry and characteristics of the anchor, as well as the soil properties such as undrained shear strength and coefficient of consolidation (horizontal and vertical). 77


Anchor geometry: -Diameter -Length - nºof flukes -Fluke configuration

Soil Characteristics: Impact Velocity: -Free-fall height

-Weight

-Undrained shear strength (su)

Determination of Tip Embedment Depth

- Coefficients of consolidation (Cv and Ch)

Soil-structure interaction and soil consolidation

Pull-out capacity

Figure 78 – Design process for torpedo anchors.

6.3.1.1. Anchor Geometry

The design of torpedo anchor structures is based on four parameters: 

Diameter, which ranges between 0.76 m and 1.07 m



Length, which may vary between 12 m and 17 m



Mass which depends on the filling and may be between 24.6 and 98 tons



The number of flukes; usually 4 but it can also be 3 or none, and the flukes may have a width from 0.45 m to 0.9 m, and be 9 m to 10 m long. The main problem with torpedo anchors is the lack of field experience, especially outside

Brazil. Therefore, the geometry suggested in this text will be based on that used in the Albacora Leste Field (FPSO P-50), a FPSO unit in water depth of 1400 m with required capacity of 7500 kN (Araujo et al., 2004). The torpedoes used for the FPSO P-50 were type T-98, which are illustrated in Figure 79 and Figure 80, and had the following characteristics (Brandão et al., 2006): 

Total mass of 98 tons



Diameter of 1.07 meters



Length of 17 meters



Four stiffener wings (flukes): 0.9 m wide x 10m long.

Figure 79- Schematic longitudinal section drawing of the T-98, Brandão et al. (2006).

78


Figure 80 – Photos of the T-98 body sections welding and its final adjustments, Brandão et al. (2006).

The ballast does not full-fill the torpedo interior, Figure 79; the reason is to lower the gravity centre of the structure closer to the nose. This greater proximity to the nose creates more stability while free-falling before reaching the seabed. Apart from oceanic currents and the misalignment of fins, stability is the principal factor in the torpedo trajectory. It has been recommended that, the centre of gravity should be 10% of the total anchor length below the hydrodynamic (geometric) centre (Hickerson et al., 1988; Raie & Tassoulas, 2009). 6.3.1.2.

Impact Velocity

Before defining the velocity reached by the torpedo when impacting the seabed, the equation for terminal velocity will be demonstrated. Normally, the impact velocity will less than the terminal velocity, since to achieve this it must have a higher fall than those recommended for torpedo anchor installation. These recommendations ensure torpedo stability during free fall. Higher drops may result in the torpedo trajectory crossing ocean currents, and greater velocities may induce stronger turbulence putting in risk the torpedo verticality and impact location. To determine the terminal velocity, Newton’s second law is considered. The forces acting on the penetrator during the fall are: drag force, anchor weight and buoyancy force (Raie, 2009). Therefore:

   =  .   = 12 ....  = . 

Where 

Wsub is the submerged weight of the anchor in the water 79

(1) (2) (3)




Fd is the drag force



m is the torpedo mass



ma is the added mass of ballast



V is the instant anchor velocity



t is the time







is the water density

Ap is the anchor frontal area



CD is the drag coefficient



g is the acceleration of gravity

After release, the velocity increases until the drag force and anchor weight have the same intensity. At this point, the incremental velocity change and thus the acceleration is zero, and the torpedo has attained the so-called terminal velocity. Considering the previous equations (Eqs. 1 to 3), the terminal velocity (VT) can be calculated:

 1 =  . = 2 ....  = √ 12 ....

(4) (5) (6)

The difficulty on this expression is in the determination of C D. Although, the drag coefficient value should change throughout the torpedo’s fall, a study from Hasanloo (2011) showed a good agreement between the measured velocities and the calculated results when using a constant drag coefficient . Table 5 summarises the CD values proposed for different penetrometer types that have been investigated. Drag coeff., CD 0.70

Penetrometer type Cylindrical penetrometers with a pointed nose

Reference True (1976)

0.15 to 0.18

European Standard Penetrators

Freeman et al. (1984)

0.03+0.0085L/D

European Standard Penetrators

Freeman and Hollister (1988)

0.63

Four fluke DPAs

Øye (2000)

0.33

Torpedo anchors

Fernandes et al. (2005)

Table 5 – Drag Coefficients according to different penetrometer shape and reference, in the Freeman and Hollister (1988) formula, L is the penetrometer length and D is the diameter.

80


Given that the company with most experience of this kind of anchor system is the Brazilian petroleum company Petrobras, the drag coefficient purposed by Fernandes et al. (2005) will be considered for the torpedo anchor assessment in this study. That value is CD = 0.33, so the terminal velocity for the T-98 is:

980009×9. 8  = √ 0.5 × 0.33 ×999. 72× 1.07/4 = 80 /

(7)

Nevertheless, it is advised, as previously noted in Section 4.6, to use drop heights above the seabed between 30 m and 150 m, which usually result in impact velocities from 0.5 to 0.33 times the terminal velocity (Medeiros, 2002). Hence, for this study, an impact velocity of 40m/s will be considered. Using equations 1 to 3 it is possible to define the acceleration when the defined velocity is reached (in this case 7.4 m/s-2), and consequently, using the equations for conservation of mechanical energy, determine the height needed to achieve the chosen impact velocity as the anchor reaches the seabed:

6.3.1.3.

 = 12 .  .  = 0.5 ×98000 ×40  = ..ℎ = 98000 ×7.4 ×ℎ  =  ⟹  − = . 

(8) (9) (10)

Soil Characteristics

As previously mentioned, accurate information about the constitution of the seabed of the São Tomé & Príncipe (STP) Economic Exclusive Zone (EEZ) is currently very limited. So, in order to continue the design of the torpedo solution, the following table summarises the characteristics of the soils adopted in Principe Island surroundings that are needed for further analysis. Soil Characteristic Undrained shear strength gradient

Adopted Value

Reference

1.5 kPa/m

Puech (2004) Kulhawy & Mayne

Coefficient of consolidation

10-8 m2/s

Wet unit weight

14 kN/m3

Puech (2004)

Sensitivity

4

Puech (2004)

Adhesion factor (α)

0.8

Puech (2004)

(1990)

Table 6 – Adopted values for different soil characteristics and references from which they were based on.

81


6.3.1.4.

Tip Embedment Depth

Tip embedment depth is the penetration into the seabed achieved by the pointed tip of a torpedo anchor. The deeper a torpedo is installed, the higher the pull-out capacity will be for a given shear strength profile. Therefore, it is absolutely crucial to accurately predict this penetration so that the resistance is well defined. The penetration of a torpedo strongly depends in three aspects: geometry of the penetrator (L/D), impact velocity of the penetrator and strength of the soil. The prediction of anchor embedment has been evolving over the years, not only because of the use of dynamic penetration models based on empirical results, e.g. True Model, and Computational Fluid Dynamics (CFD) (Raie & Tassoulas, 2009; O’Loughlin et al., 2013) ; but from the implementation of numerical procedures for coupled finite element analysis of dynamic problems in geomechanics as well (Aguiar et al., 2011; Carter & Nazem, 2013; Sabetamal et al., 2014). Each of these approaches have proved to be reasonably accurate. Petrobras, which developed the torpedo technology, perform a conductor drive analysis

to evaluate the penetration of the torpedo base system using a computer program based on the model proposed by True (1976). This model relies on soil parameters whose values are assumed as known, fixed, and deterministic. Yet, it is common sense that these soil parameters have a significant degree of variability, which may affect negatively the accuracy of the response given by the simulation method (Kunitaki et al., 2008). On the other hand, a more accurate approach is to predict the penetration based on empirical model trials in the field or in lab tests. Though, first carrying out a CFD or FEM prediction and confirming it with field trials or lab tests is the recommended procedure. Applying numerical procedures is out of the scope on this work, but would be recommended for future studies. So, making a prediction for the STP offshore area requires a review of studies made in this field. Richardson et al. (2009) carried out several laboratory tests and compared the results to empirically based predictions. These laboratory tests involved a series of centrifuge model tests of 1:200 reduced scale model anchors in normally consolidated clay (Kaolin). Table 7 summarises some relevant characteristics of the tests that were performed. These tests studied the influence of different soil properties in the embedment of model projectiles, which were designated as either quasi-static or dynamic installation, with impact velocities up to 30 m/s. The influence of soil properties such as shaft adhesion and undrained shear strength gradient in the resulting penetration are illustrated in the graphics of Figure 81 and Figure 82 respectively.

82


Dimension

Model

Prototype

Anchor Length

75 mm

15 m

Anchor Diameter

6 mm

1.2 m

Mass

6.2-12.7 g

49.6-101.6 tons

0.21 kPa/mm

1.05 kPa/m

Average Undrained Shear Strength Gradient

Table 7 – Model and Prototype anchor dimensions and soil property.

Using the studies by Richardson et al. (2008) of the sensitivity of embedment depth predictions to shaft adhesion factor (Figure 81), it was possible to interpolate for an adhesion factor of 0.8 and extrapolate an impact velocity of 40 m/s to predict a penetration of 40.4 m for the T-98 torpedo being considered. Nevertheless, this value is not accurate since the undrained shear strength gradient in this case is lower than the 1.5 kPa/m, presumed value for STP. However, on examination Figure 82, contains a curve for an undrained shear strength gradient of 1.5 kPa/m, and it indicates that the penetration would be greater than 40.4 m. This happens because the adhesion factor used for this trial was 0.4, much lower than the value of 0.8 assumed for the T-98 torpedo. As a result, this prediction of 40.4 m is an upper bound limit for the final value.

0

40

α…

Figure 81 – Sensitivity of embedment depth predictions to shaft adhesion factor according to Richardson et al. (2008) and interpolation for adhesion factor of 0.8 and impact velocity of 40m/s.

83


Figure 82 – Sensitivity of embedment depth predictions to undrained shear strength gradient, Richardson et al. (2008).

O’Loughlin et al. (2013) after gathering penetration data from worldwide field tests and

comparing them to centrifuge tests of equivalent prototype scale models, were able to propose a relationship that predicted penetration depth with reasonable accuracy for this very large dataset that encompassed a wide range of anchor masses, geometries and impact velocities, Figure 83.

Figure 83 – Comparison of centrifuge and field test embedment data, O’Loughlin et al. (2013).

84


From Figure 83 it is possible to see the good agreement between the formulated curve and the dataset. So, the prediction for STP using the O’Loughlin et al . (2013) proposed relationship, would be:

 ≈ . /   =      = 0,5..  .  .   .  .   0, 5 .  .   + ≈  1.5 × 1.17  ×1.17 = 39.24 

(11) (12)

(13)

Where m’ is the effective mass of the anchor (anchor mass minus mass of soil displaced) and g is the Earth’s gravitational acceleration. Penetration had to be calculated using an iterative process in Equation 13. It is important to notice that the ratio Z e/deff =33.54, according to Figure 83, is not as accurate as to lower Z e/deff ratios (0 to 20), and when examining Figure 83, it is apparent that Equation 13 provides a conservative prediction of Z e/deff ratio for a given situation or installation energy value. The two grey lines traced in the graph bound almost every result, additionally the grey line below the original equation line intersects the field test result from T98 (Brandão et al., 2006). Therefore, this ratio may vary 13% from the real depth, so the penetration achieved is expected to be between 36 m and 43 m. Freeman et al. (1984) reported penetrations of 30 meters for penetrators with an aspect ratio of 10, weight of 1.8 tons, and an impact velocity of 50 m/s when installed off the West coast of Africa, and Freeman and Burdett (1986) report penetrators reaching 62 m/s with an associated burial depth of up to 35 m at the NAP site (offshore of the Bahamas). 6.3.1.5.

Effects of installation in the resistance of the soil

The effect of installation cannot be ignored, as it generates an initial perturbation of the soil during installation. A reduction in the undrained shear strength in the vicinity of the anchor occurs not only because of the change in the initial stress state, but also because of the shearing of the soil caused by the penetration and the soil deformation with a constant volume (Aguiar, 2011). As the pile penetrates into the soil, the soil moves around the pile in the vertical direction (Komurka, 2003). Randolph et al. (1979) comment that the penetration of a cylindrical pile into a clay soil affects a region of up to 20 times the diameter of the penetrating pile. The disturbance in the soil around the pile produces an excess of pore-water pressure, which reduces the effective stress in the vicinity of the pile. As the volume of mobilised soil tends 85


to be equal to the volume of the pile, this issue is more significant when installing piles with larger diameters. The process of reconsolidation of the soil is described as the dissipation of the high pore-pressure generated during the pile installation, while this occurs the effective pressure increases, thus increasing the lateral resistance of the pile (Richardson et al., 2009). Therefore, this reconsolidation process is a very important aspect in the torpedo design, and must be carefully predicted. Several studies have been made in order to better understand this phenomenon. Richardson et al. (2009) carried out a series of experimental tests. Comparing the results with theoretical models, they indicated that 50% of the resistance acquired by anchors with dynamic installation was achieved within 35 to 350 days after the pile installation. In the same studies, the authors comment that 90% of the capacity is achieved within 2.4 to 40 years after the pile installation. It is important to notice that these values are highly dependent on the soil characteristics where the pile is installed, and the authors recommend the use of this kind of anchor where reconsolidation occurs rapidly, i.e. locations where the pile achieves 50% of the resistance capacity in a period within 30 to 90 days. It is also important to mention that the applied loads are much lower than the total capacity of the pile, for the reason that safety factors used in this type of anchor is between 1.5 and 2.0 (Eltaher et al., 2003). Therefore, because the design load conditions are usually associated with a long period, this means that the reconsolidation period is sufficient to achieve the resistance capacity taking into account the safety factor used. 6.3.1.6.

Pull-Out Capacity

The pull-out capacity depends on the characteristics of the soil (i.e. undrained shear strength), the geometry of penetrator (e.g. weight, number and geometry of the flukes, shaft area of the pile) and the angle between the applied load and the plane of the flukes. Therefore, it is important to predict the time the soil takes to reconsolidate, thus, estimating the short and long-term pull-out capacities. Despite the importance of soil reconsolidation, this will not be approached in this work, and only the means for prediction of the long-term pull-out capacity will be examined and evaluated for the case being considered. Normally, pull-out capacity is predicted using FEM programs and laboratory model tests. The estimation using the finite-element method has some limitations, since the pile is usually considered to be vertically oriented in the soil with an initial state of stress in the soil equal to the state prior to the anchor installation and the initial excess pore-water pressure equal zero, ignoring the installation effect (Medeiros, 2002; Lieng et al., 2000).

86


The holding capacity behaviour of the FPSO P-50 in Campos Basin (Brazil) was studied using a 3-D finite element program, called AEEPEC3D, Brandão et al. (2006). During the torpedo pile design the following aspects were considered: 

The wings of the torpedo pile were considered to be located in the alignment that produces the minimum load capacity (i.e. 45O off the alignment of the load – Plane 1 in Figure 84).



The submerged self-weight of the torpedo pile was considered in the load capacity evaluation.

Figure 84 – Cross section of a torpedo anchor with four flukes (Aguiar el al., 2009).

In Figure 85, the graph shows the results for loads applied to a vertically installed pile at three different angles (0, 43.6 and 90 degrees to the horizontal) and for three piles installed 10, 20 and 30 degrees from the vertical (tilt) with the load applied 43.6 degrees from the horizontal (Brandão et al., 2006).

Figure 85 – T-98 Holding Capacity for 4 tilts and 3 different load directions in Campos Basin, Brandão et al. (2006).

87


The minimum ultimate capacity is associated with the pull-out load being applied along the axis of the pile (pink line in Figure 85). Even if, the pile suffers a deviation in its penetration, the pull-out capacity will be the lowest when the load acts in the same direction as the pile is installed. This tendency is proven by the decline of capacity when the tilt on the pile tends to be in the direction of the load (lines green, blue and red). Therefore, determining the vertical pile resistance is a good approach in terms of the evaluation of a safe design load. This resistance can easily be determined using some simple calculations; the vertical pile resistance is given by the sum of three components:

 =       =  × ̅×  =  × ×

Where,

(14)

    ̅   

NC

(15) (16)

friction resistance obtained from the torpedo shaft plus the flukes weight of the torpedo pile (980 kN) resistance obtained from the soil that has to move so the pile can be pulled out (reverse end bearing resistance) shaft area plus the fluke area that contributes to friction resistance and it is 129 m2 average shaft undrained shear strength along the pile adhesion factor transversal area of the torpedo pile unit weight of the soil end bearing capacity factor (12) Offshore Site

Campos Basin, Brazil

São Tomé & Príncipe

Tip embedment depth (m)

33

40

Average undrained shear strength gradient (kPa)

5+2z

1.5z

Average undrained shear strength on shaft (kPa)

54

47.25

Adhesion factor (α)

0.6

0.7

15

14

Rs (kN)

4180

4880

Rb (kN)

2590

3470



(kN/m3)

88


Vertical pile resistance (kN)

7750

8720

Table 8 – Soil properties, penetration and vertical resistance of Campos Basin in Brazil and STP offshore. Average undrained shear strength gradient was proposed by Medeiros (2001).

Table 8 shows the properties of the soil and its interaction with the pile, as well as the installation specifications of the pile, and finally, the vertical pile resistance determined using Equations 14 to 16. The friction resistance and reverse end bearing resistance are the two principal sources of resistance; however, both of them are very sensitive to an empirical factor. On one hand, the reverse end bearing factor has a huge effect on the resistance. Unfortunately, at present, there is no agreement on which N c value should be used. Based on experimental results, Watson et al . (2000) suggested that the tension bearing resistance is similar to the compression bearing resistance in terms of magnitude, therefore, the reverse end bearing value may be calculated in a similar way to that for compression end bearing. On the other hand, shaft friction resistance is highly dependent on the adhesion factor, this factor changes a lot within the consolidation of the soil around the pile. Although, adopted value of 0.8 was adopted for the adhesion factor, as proposed by Puech (2004), there are some divergent opinions from others authors. The α factor is often estimated as 1/St , where St is the sensitivity,

though, this would suggest quite adhesion low factors in some clays which are not borne out by testing, i.e. in this case, a sensitivity of 4 would imply α = 0.25, which is quite low (Houlsby & Byrne, 2005). Dendani & Jaeck (2007) reported that the axial resistances recorded during model tests of pipelines in clay recovered from the GoG were best predicted by taking α=0.7 and α=0.35 for the peak resistance and residual resistance, respectively. Thus, and assuming that the general structure and undrained shear strength of the clay in the tests was the same as that in situ, the shaft resistance for the torpedo anchor considered here could be estimated as 2130 kN in the short-term (with the soil remoulded, i.e. residual strength applies) and 4270 kN for the peak resistance, i.e. after the soil has reconsolidated following installation and recovers the natural structure present prior to installation. Note that α = 0.7 is lower than the 0.8 value proposed by Puech (2004), which may indicate that it is not expected that the soil recovers all its initial structure.

89


6.3.2. SEPLA The development of suction embedded plate anchors (SEPLA) emerged from the need for a reliable alternative to driven piles, suction embedded piles and drag-embedded plate anchors (VLAs). With this in mind, SEPLA was conceived to combine the advantages of suction piles and VLAs without their drawbacks: Anchors

Advantages

type 

Known penetration



Geographical location

Suction pile

VLA



Drawbacks 

Large



Costly



Difficulty to handle



High loads required to achieve initial penetration and final capacity.

Geotechnical efficiency



May not be suitable where precise positioning specified.

Table 9 - Advantages and drawbacks of suction piles and drag embedded plate anchors (VLAs), Wilde et al. (2001)

The application of this type of anchor is not entirely new in the GoG. SEPLA anchors were selected by ExxonMobil in 2003 for mobile offshore drilling unit (MODU) moorings between the Xicomba and the Kizomba A fields offshore from Angola (Dahlberg et al., 2004). The functional requirement of SEPLA is to resist the specified maximum factored mooring line load, while avoiding significant displacements, both in the direction of the applied load or vertically. SEPLA holding capacity is related primarily to three basic aspects, which must be defined for the design of the solution, those aspects are: 

Anchor plate area



Undrained shear strength



Penetration depth

6.3.2.1. Anchor Geometry and Installation

The SEPLA consists of a solid rectangular steel fluke (approx. 20 cm thick), a shank connecting the fluke to a pad-eye (i.e. the loading point) and a full-length keying flap hinged to the top edge of the fluke, Wang et al. (2012), Figure 86.

90


Figure 86 – Typical SEPLA with keying flap, Wang et al. (2012)

Typically, for mobile offshore drilling units (MODU), the fluke is a solid steel plate 2.5 m t o 3.0 m in width and 6 m to 7.3 m in length. For permanent installations, the fluke is usually 4.5 m wide and 10 m long (Wilde et al., 2001). According to information published by Technip, for two example applications of these solutions for MODU and a permanent installation called Red Hawk , the suction embedded followers had a diameter of 3 m and 5.5 m, and the installations depths were 21.5 m and 24 m respectively. Courtesy of Technip, http://events.energetics.com /deepwater/pdfs/presentations/session5/gengshenliu.pdf . Based on the above, the geometry and the characteristics proposed for SEPLA applied offshore of STP are: 

Plate area: 10 m length by 4.5 m width



Shank: 2.5 m high (padeye eccentricity) and at an angle of 60ᵒ with the fluke.



Anchor thickness: 0.20 m



Anchor weight: 50 tons



Installation penetration: 24 m

During installation, the SEPLA uses a suction pile (referred to as a "follower") for embedment to the design penetration. The SEPLA is inserted in slots at the bottom of the follower and retained by the mooring line and recovery bridle. The suction follower, with the SEPLA slotted into its base, is lowered to the seafloor, allowed to self-penetrate and then suction embedded in a manner similar to a suction pile. Once the SEPLA has reached its design penetration depth, the mooring line and retrieval bridle that hold the SEPLA secure in the bottom of the follower are released by the installation ROV 91


(remote operated vehicle). The pump flow direction is then reversed and water is pumped back into the follower; the follower moves upward, leaving the SEPLA in place. The SEPLA fluke will be in a vertical orientation at this time. For a recoverable application, a small submersible buoy will support the recovery bridle. At this time, the mooring line is tensioned by the installation vessel in the direction the SEPLA is to be loaded. This "keying” tension will: • Pull the initially vertical mooring line through the s oil so that it forms an inverse

catenary shape from the mudline to the anchor shackle. • Start rotation of the SEPLA fluke to an orientation perpendicular to the direction of the

mooring line at the anchor end. • Set the keying flap to prevent fur ther loss of penetration beyond that necessary to set

the keying flap. At this time, the SEPLA is ready to develop its full pull-out capacity. Regardless of the initial orientation of the fluke to the mooring line, the SEPLA with its long, rigid shank will rotate to present the maximum projected area to the direction of pull. This ensures that the ultimate pullout capacity, based on the anchor's penetration depth and soil properties, is achieved (Wilde et al., 2001). However, keying also represents the major issue of the SEPLA, since the plate moves vertically and horizontally in addition to rotating, this leads to several uncertainties associated with the final embedment depth during operation. 6.3.2.2.

Keying

The anchor keying process promotes two negative effects in the plate holding capacity. First, it induces an upward movement on the anchor during the plate rotation, hence reducing the embedment depth; and second, the soil in the immediate vicinity of the plate anchor is remoulded, therefore reducing the soil strength (Randolph et al., 2005). Even though, this latter effect may be recovered as the soil reconsolidates, the loss of embedment is permanent. As clay deposits in Gulf of Guinea are typically characterized by an increasing strength profile with depth, any loss in embedment will correspond to a non-recoverable loss in potential anchor capacity. The U.S. Naval Civil Engineering Laboratory (NCEL) has produced some guidelines (NCEL, 1985) about the loss of embedment during anchor keying; recommending the design takes into account a loss of twice the anchor breadth (2B) in cohesive soils. However, NCEL (1985) 92


recognizes that the loss of embedment is also a function of anchor geometry, soil type, soil sensitivity, and duration of time between penetration and keying, thus these aspects may reduce or enlarge the embedment loss. More recently, Wilde et al. (2001) reported in situ full scale and reduced scale onshore and offshore test results for SEPLAs in clay. Soil sensitivity was in the range of 1.8 to 4.0 for the different test sites and the loss of embedment during keying was 0.5 to 1.7 times the anchor breadth, with lower embedment losses corresponding to higher soil s ensitivities. None of the studies previous to O’Loughlin et al. (2006) studied the effect of loading

eccentricity on the keying process. The loading eccentricity is the distance between the padeye, where the keying forces are applied, and the centre of rotation of the plate anchor; Figure 87 illustrates the plate anchor setup, where the loading eccentricity represented by the letter “e”. O’Loughlin et al . (2006) showed a strong dependence between the loss in embedment and loading

eccentricity (e). At a large eccentricity ratio ( e/B≥1), the loss in anchor embedment was no greater than 0.1B. When the eccentricity ratio was less than 1, the loss in anchor embedment increased linearly with the reduction of the eccentricity ratio, as is shown in Figure 88. As the plate displaces and rotates, F may be expressed in terms of forces perpendicular (F n) and parallel (Fs) to the plate and a moment (M) at the mid-point of the plate, Figure 87 . The larger the eccentricity is, the larger will be the moment on the plate anchor caused by the keying forces, the more rapidly the plate anchor will rotate and the less it will translate. Figure 89 illustrates the effect of high eccentricity and low eccentricity loading paths, low eccentricity requires a much higher Fs to achieve the same Fn. This greater parallel force makes the plate translate upwards before rotating.

Figure 87 – Plate anchor setup before keying process.

93


Figure 88 – Graph showing the results achieved by O’Loughlin et al. (2006) for eccentricity ratios between 0.17and 1.0.

Figure 89 – Combined loading paths for high and low eccentricity plate anchors, O’Loughlin et al. (2006).

94


Gaudin et al. (2006) investigated the influence of suction installation on anchor keying and anchor capacity. They found that the loss in embedment after a suction installation is lower than after a jacked-in installation, which is when the plate is pushed into the soil. For a square anchor with e/B=0.66 and a 45ᵒ pull-out angle, the loss of embedment was reduced from the range of 1.3 to 1.5 times the anchor breadth for a jacked-in plate to a range of 0.9 to 1.3 times the anchor breadth for a suction installed plate anchor. Gaudin et al. (2006) concluded that greater soil disturbance during suction installation was responsible for the lower loss in embedment during anchor keying. As the disturbance drops the t he shear strength of the soil, the t he resistance that the soil gives to the rotation of the plate is lower, thus the rotation is faster and the loss in embedment is lower. They also observed a strong correlation between the loading angle and the loss in anchor embedment. Taking into account the issues discussed in these previous studies, Song et al. (2009) ran several centrifuge tests and developed large deformation finite-element (FE) analyses. From the results of those studies, they found that t hat the loss in anchor embedment during anchor keying may be expressed in terms of a non-dimensional anchor geometry factor, which is a function of the eccentricity of the padeye, angle of loading, and the net moment applied to the anchor at the st age where the applied load balances the anchor weight. The procedure to estimate the loss in anchor embedment to any given pull-out p ull-out angle during keying is summarized as follows: 1.

Calculate the loss in anchor embedment (Δz e/B) during vertical pull-out (θ=90ᵒ ), using

eq. (17) taking into account all the anchor geometry measurements and anchor submerged unit weight. The initial moment M0 corresponding to zero net vertical load on the anchor is given by eq. (18).

∆ =   .0.2     .  =         ′

(17) (18)

2. Determine the constant C θ, using the gradient kθ=0.005 (per degree of pull-out angle) and the loss in anchor embe dment (Δze/B) calculated in the first step for a pull-out angle θ=90ᵒ.

 = ∆   ∙ 

(19)

3. Finally the loss in anchor embedment under any given pull-out angle is obtained from eq. (20), substituting the constant Cθ found in the second step. st ep.

95


∆ =    ∙ 

(20)

Where,

∆         

Loss in anchor embedment Loading eccentricity for pull-out force Anchor thickness Initial moment corresponding to 0 net vertical load Anchor area Gradient for estimating anchor loss in anchor embedment Shank resistance Overall submerge anchor weight Loading eccentricity for friction resistance Loading eccentricity for anchor weight

In Appendix II, the calculations made to determine the loss in anchor embedment according to Song et al . (2009) for this study are detailed. Below the results of the calculations using equations (17) to (20) are presented: p resented:

∆/   =

= 0.94 = 5080.71 kN.m 0.49

The equation that gives the loss of embedment under a given pull-out pull -out angle in the studied area is:

∆  = 0.0.49  0.0.005 ∙  ⟹ ∆ = 0.94 × 4.5 = 4.1919  ≈ 4

(21)

This prediction (0.94B), according to Song et al . (2009), has a good agreement with that expected according to Wilde et al . (2001), which gives a range between 0.5B and 1.7B, and also with Gaudin et al . (2006) that predicts displacement in the range of 0.9B to 1.3B for suction embedded plate anchors. From O’Loughlin et al. (2006), the expected loss of embedment is 0.85B for a ratio e/B=0.58,

however the Song et al . (2009) formulation is more conservative because the equation that fitted the results obtained in their studies with correlation coefficient of R2=0.95 was

96


∆ =   .0.15  .   

(22)

Equation 17 is an upper bound of the fitted line which provides a conservative design curve, Figure 90. Therefore, for this case study, it is plausible to say that the SEPLA embedment depth after installation is about 20 m.

Eq.17 Eq.22

Figure 90 – Loss in anchor embedment during keying versus anchor geometry factor, Song et al. (2009).

6.3.2.3.

Plate Holding Capacity

As previously stated, the functional requirement of SEPLA is to resist the applied loads without undergoing significant displacements. SEPLA holding capacity can be divided into short and long term capacities. The short term capacity represents the capacity the soils have when they mostly have an undrained behaviour. This situation suits best in typical storm conditions, and where soils have low permeability like highly plastic clays. Long term capacity is the capacity the SEPLA must have to resist long term quasi-static loads in permanent installations. This term is not considered critical for deeply embedded plate anchors in normally consolidated plastic clays, since the effective stress conditions around a p late anchor will gradually change as the excess pore pressure dissipates. Hence, the soil will have greater long term capacity value than the short term for normally consolidated soils.

97


There are several suggested procedures for evaluating the plate holding capacity in clay, and in this work this estimation will be made according to three different procedures, which are: 

Wilde et al. (2001)



Merifield et al. (2001)



DNV-RD-E302 (2002)

Wilde et al. (2001) proposal

Wilde et al. (2001) based their procedure on the general bearing capacity equation for square or rectangular foundations, which has the following formulation:

Where,

 =  ×  × , × 10.2 × 



bearing capacity of the anchor

Nc

bearing capacity factor

 su,avg

(23)

projected bearing area of the fluke and keying flap average undrained shear strength across the anchor plate after keying

B

width of the plate

L

length of the plate

This expression assumes full suction can develop behind the fluke. Therefore, the bearing capacity factor used is 12.5 (Brinch-Hansen, 1970). However, if somehow water is able to flow to the back of the plate anchor, and the suction formed on the back is r elieved, the value of Nc should be reduced to 7.5 for deep failures modes (i.e. more than 5 fluke widths deep), for shallow failure modes the author does not give any guidance (Wilde et al., 2001). Yet this expression can be refined, because during deployment and proof loading a certain amount of remoulding and soil disturbance near the SEPLA occurs due to the installation process (penetration plus keying). This results in a temporary loss in available bearing capacity (NCEL, 1977). To take this effect into account, an additional term α n has been incorporated. Thus, the revised equation is:

 =  ×  × , × 1 0.2 × ×

(24)

Based on US Naval Civil Engineering Laboratory (NCEL) experience and tests, the value for αn for clays is expected to range from 0.7 to 1.0, depending on clay type and sensitivity. For

98


clay sensitivities of 2 to 5, which are characteristic in the GoG, Wilde et al. (2001) recommend the αn take a value of 0.7.

Merifield et al. (2001) proposal

Merifield et al. (2001) applied numerical limit analysis to rigorously evaluate the stability of vertical and horizontal strip anchors in undrained clay. By using two numerical procedures that are based on finite element formulations of the upper and lower bound theorems of limit analysis, the ultimate pull-out capacity was obtained. Those formulations followed a standard procedure by assuming a rigid perfectly plastic clay model with a Tresca yield criterion. The results were compared with existing numerical and empirical solutions, and for most of the cases, the exact anchor capacity was proved to be predicted to within 5%. The equation used to quantify the ultimate anchor pull-out capacity in undrained clay is:

 =  ×  × ,  

(25)

Where Wa is the weight of the anchor.

It is apparent that this equation is quite similar to that proposed by Wilde et al. (2001), the main difference is found in the estimation of the bearing capacity factor (Nc). Once again, N c relies in the soil-anchor behaviour. The anchor behaviour can be divided into two distinct categories: immediate breakaway and no breakaway. The immediate breakaway case is when the soil-anchor interface cannot sustain tension, while the no breakaway case is assumed when there is adhesion or suction between the anchor and the soil. In reality, it is likely that the true breakaway state will fall somewhere between the extremities of the immediate breakaway and no breakaway case (Rowe & Davis, 1982). The suction force developed between the anchor and soil is a function of several variables, such as: embedment depth, soil permeability, undrained shear strength and loading rate. Therefore, with so many variables, Merifield et al. (2001) propose taking into account the immediate breakaway case only, resulting in a conservative estimation of the actual pull-out resistance with suction. Figure 91 shows the general layout out of the problem considered in the analyses.

99


Figure 91 – Problem notation

Merifield et al. (2001) studied the influence of the failure mode in the ultimate resistance of the anchor. The anchors are classified as shallow, if at collapse, the observed f ailure mechanism reaches the surface, Figure 92(a) and (b). In contrast, a deep anchor is one whose failure mode is characterized by localized shear around the anchor and is not affected by the location of the soil surface, Figure 92(c). For a given anchor geometry and soil type, there is a critical embedment depth, H cr, at which the failure mechanism no longer extends to the soil surface and becomes localized around the anchor. When this type of failure occurs, the ultimate capacity reaches a maximum limiting value. Physically, this transition arises because the undrained shear strength is assumed to be independent of the mean normal stress. From a practical point of view, this result is important as embedding the anchor beyond H cr will not lead to an appreciable increase in anchor capacity, Figure 92(d). Even though, the studies made by Merifield et al. (2001) cover most problems of practical interest, the situation defined in this work is out of this range of theoretical solutions. These are derived for problems where H/B ranges from 1 to 10 and ρ B/su,o varies from 0.1 to 1. In this work H/B=20/4.5=4.44, but ρ B/su,o = 1.5x4.5/5 = 1.35.

100


Figure 92 – Shallow and deep anchor behaviour, Merifield et al. (2001)

The suggested procedure for estimation of uplift capacity is as follows: 1. Determine representative values of the material parameters s u,o (undrained shear strength at the ground surface), γ (soil wet unit weight) and ρ (undrained shear

strength gradient). 2. Knowing the anchor size B (width) and embedment depth H, calculate the embedment ratio H/B and overburden ratio γH/s u,o.

3. Calculate the break-out factor Nco using Figure 93 depending on the anchor shape. 4. Adopt Nc*=11.16 (for horizontal anchors) 5. In soil with

 >0  =  1 0.383 , 2  1

, calculate the break-out factor N coρ using:

(26)

6. Calculate the break-out factor N c using:

 =   ,

7. Calculate the limiting value of the break-out factor

101

(27)

∗

using:


8. If

 ≥ ∗  ≤ ∗

(28)

then the anchor is a deep anchor. The ultimate pull-out capacity is

given by (25), where 9. If

∗ = ∗ 1 ,  = ∗  .

then the anchor is a deep anchor. The ultimate pull-out capacity is

given by (25), where

is the value obtained in Equation (27).

Figure 93 – Design chart for rectangular anchors in clay, allows determining anchor break-out factor Nco at various embedment ratios (Merifield et al., 2003).

DNV-RP-E302 (2002)

This design code is based on the results from the Joint Industry Projects “Reliability analysis of deep water plate anchors” and “Design procedures for deep water anchors” (Dahlberg et al., 2004). It applies to the geotechnical design and installation of plate anchors in NC clay, including the suction embedded anchors, for the several types of moorings systems. According to this code, the design of the anchor must account for installation effects on the plate anchor resistance. In particular, the design must specify a minimum anchor penetration, which should be verified during the anchor installation. The safety requirements are based on the limit state method of design, and the design criterion shall be satisfied by:

 ≥ 0

102

(29)


Where



  =  ∙  ∙  ∙,∙ ∙  ∙ 1⁄

is the design value for the anchor resistance and

is the design value of the

mooring line tension. The design anchor resistance is given by:

(30)

The parameters in eq. (30) are:

 = ,/, , ,   = 1 0.2 ∙ /      

Cyclic loading factor Cyclic shear strength Mean static undrained shear strength within soil volume influencing the anchor resistance Bearing capacity factor Shape factor Equivalent plate width Equivalent plate length Empirical reduction factor Plate area Installation penetration depth of plate Partial safety factor

Most anchors can utilise the beneficial effect of cyclic loading, which may increase the anchor resistance by 10% to 20% in normally consolidated clay when subjected to storm loading, because of the suction forces developed in the back of the plate anchor. For special environmental loading conditions, such as loop currents or wind-generated squall conditions, the beneficial effect of cyclic loading will be marginal. Therefore, in a conservative approach U cy is set to be equal to 1.0, since it normally is larger than 1.0. The bearing capacity factor, Nc, is valid for plane strain conditions (strip footing) and isotropic, incompressible material, and is adjusted for the plate geometry through the shape factor sc. If the anchor is installed at a depth equal5.14or superior to 4.5W F, then it is considered a deep anchor and Nc is set to 12.0. However, if the installation depth is less than 4.5WF, the penetration is considered shallow and the equation or graph in Figure 94 must be used to obtain the appropriate bearing capacity factor. 103


12

Figure 94 – Equation and curve describing the variation of N c value in shallow failure zone, Dahlberg et al. (2004).

The factor η is an empirical reduction factor, set equal to 0.75. This factor accounts for: 

Progressive failure (strain-softening) of the clay within the soil volume involved as the plate anchor is loaded to failure



Strength anisotropy, to the extent the actual average undrained shear within the soil volume affected by the failure differs from the assumed shear strength



Load eccentricity if the anchor is not loaded normally to the plate

This factor may be higher or lower depending on laboratory soil strength test data. For example, clay with a high sensitivity, which exhibits a significant strain-softening, may require a lower η value, whereas one with insignificant strain- softening may use a higher η value. In addition, a greater load eccentricity may also need a lower η factor. The design criteria can be formulated in terms of two limit state equations: a) An ultimate limit state (ULS) to ensure that the individual mooring lines have adequate strength to withstand the load effects imposed by extreme environmental actions b) An accidental limit state (ALS) to ensure that the mooring system has adequate resistance to withstand the failure of one mooring line, failure of one thruster, or a failure in the thruster system for unknown reasons. There are also two different classes of consequences, which are considered for both ULS and ALS: 104


1. Class 1 – Failure is unlikely to lead to unacceptable consequences such as loss of life, collision with an adjacent platform, uncontrolled outflow of oil or gas, capsizing or sinking. 2. Class 2 – Failure may well lead to unacceptable consequences of the above types. Limit State: Consequence Class: Partial safety factor



ULS

ALS

1

2

1

2

1.40

1.40

1.00

1.30

Table 10 – Partial safety factor for anchor resistance, Dahlberg et al. (2004).

Predicted Plate Anchor Capacity

The predicted plate anchor capacity according to the three different methods is expressed in Table 11, the calculation steps made to achieve those results are presented in Appendix III. The Merifield et al. (2001) procedure returned an ultimate holding capacity 50% higher than the others, and this may be explained by the two facts. Firstly, one parameter of the evaluated situation is outside of the range of the theoretical solutions, the value of

 must be between 0.1 ,

and 1.0, but in this case is equal to 1.35. Secondly, Merifield et al. (2001) do not use any empirical reduction factor whereas the other two methods do; 0.70 is used by Wilde et al. (2001) and 0.75 in DNV-RP-E302 (2002). Wilde et al. (2001) and DNV-RP-E302 (2002) have similar results, and the major source of their difference is the usage of a partial safety factor by DNV-RP-E302 (2002), for that reason a partial factor of 1.4 is also applied in Wilde et al. (2001) method to obtain a design holding capacity. For the design holding capacity in Merifield et al. (2001) method if safety factors (of 1.4) are applied to both shear strength gradient (ρ) and initial shear strength (s u,0), the design holding

capacity will be about 3.5 MN higher than DNV-RP-E302 (2002). The difference is explained by the fact that Merifield et al. (2001) does not use and empirical reduction factor and they include de self-weight of the anchor in the plate resistance. If these were similarly taken into account for Merifield et al. (2001) the resistance would be about 9.4 MN, which is quite similar to the Wilde et al. (2001) and DNV-RP-E302 (2002) results.

Contrary to Merifield et al. (2001), Wilde et al. (2001) and DNV-RP-E302 (2002) procedures do not take into account the self-weight of the plate anchor in their resistance. This conservative decision means not taking into consideration 500 kN of resistance, which represents 4 to 5% of the anchors holding capacity. 105


As a result, is believed to be more reliable to use DNV-RP-E302 (2002) method, since it seems to have a better calibration of its empirical reduction and safety factors. Thus, plate anchor capacity adopted is 9.5 MN, determined after DNV-RP-E302 (2002) plus the anchor self-weight, 0.5 MN, therefore the predicted plate anchor capacity in this case study is predicted to be 10 MN. Procedure suggested

Ultimate Holding

Design Holding

by:

Capacity (MN)

Capacity (MN)

Wilde et al. (2001)

13

9.2


18

13

DNV-RP-E302 (2002)

13.3

9.5

Table 11 – Plate anchor holding capacities in the idealized STP offshore conditions according to three different methods.

6.4. DISCUSSION OF RESULTS Both of the considered anchor solutions are likely to be economically competitive alternatives to conventional offshore anchors, for application in STP. In Table 12 and Table 13 are enumerated, respectively, the advantages and disadvantages of both SEPLA and torp edo anchors. Torpedo type anchors are simple and economic to fabricate, and are able to be installed easily and quickly. On the other hand, the anchor element of the SEPLAs represents the lowest cost of all the deep-water anchors (including torpedoes). Moreover, it only needs one follower (suction caisson) to install any required number of anchors and due to their small size; multiple SEPLAs might be carried on a single installation vessel in a single voyage (Ehlers et al., 2004). For SEPLA, all these aspects make the system significantly less costly than e.g. compared to suction anchor systems; the cost is about half to one-third (http://events.energetics.com/deepwater/pdfs/presentations/ session5/gengshenliu.pdf ). Both of the torpedo and the SEPLA solutions have a disadvantage in common which is their patented installation method. The installation methods for torpedo anchors and SEPLAs are very distinct. The latter uses a proven installation methodology based on suction caisson technology that has been in use for the last 30 years, this technique provides an accurate measure of embedment and position of the anchor, before keying. The torpedo anchor installation is simply achieved by dropping the torpedo from a certain height above the seafloor and letting it drive freely into the soil. Despite the simplicity of the torpedo method, it has the disadvantage of the final orientation (inclination) of

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the torpedo pile being unknown, which has been shown, can lead to the load resistance being lower than expected. In the economic aspect of the installation, torpedo anchors have a small advantage over SEPLA, in that while they both need an Anchor Handling Vessel, the torpedo has no need of a remote operating vehicle, while SEPLA does. The geotechnical design of these anchor systems is very different as well. The design of Torpedo anchors is patent protected by Petrobras, which means there are very few documented installation and the design methods are unavailable to the public. However, it is known that a computer program based on the True model (True, 1974) is used to evaluate the penetration of the torpedo, and finite element analysis is used to carry out a detailed non-linear analysis to predict the pull-out capacity of the pile. The estimation of the penetration depends on several factors such as the layering of the soil, the precise soil strength profile and the presence of stiff elements. These factors may be difficult or impossible to identify and they have the potential to stop the penetration of the anchor or to deviate the penetration path from the vertical. Furthermore, the pull-out capacity relies on the adhesion between the soil and the shaft of the pile, the adhesion increases with the reconsolidation of the soil in the vicinity of pile after installation. It is very difficult however to quantify this increment without field trials and long term experience. Since there is no experience outside Brazil, and the soils are not exactly similar to the soils present in Gulf of Guinea, it would be very important to carry out those full-scale tests before adopting this solution. Apart from the loss of embedment during keying, the position of the SEPLA is precisely known, because it is positioned wherever it is needed in a controlled manner. SEPLA design is based on classic limit equilibrium methods, for which the most difficult parameter to determine is the bearing capacity factor. However, intense studies have been made to understand the behaviour of plate anchor failure and to quantify the bearing capacity factor. Besides, the design of the SEPLA being more reliable than the torpedo, the SEPLA have already been put into use in the Gulf of Guinea, which can provide important information of the SEPLA behaviour in a similar situation. Finally, when comparing the resistance obtained by each foundation element, the SEPLA has a greater advantage. After applying safety factors in the proposed SEPLA, solution the predicted resistance is about 10 MN. For torpedo anchors the ultimate resistance is predicted to be 8.7 MN, however the use of a safety factor of 2 is advised (Eltaher et al., 2003), which reduces the design

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resistance to about 4.4 MN. Thus, double the number of torpedoes will be required to provide the same load resistance as a single plate anchor. In other words, the resistance obtained per unit weight of the anchor element is higher for SEPLA,

     ⁄ = 200 ; than for the torpedo anchor,       = 44 ⁄ .

Even though, both torpedo and suction embedded plate anchors have smaller unit costs in comparison with the other types of anchoring systems, one of them is more suitable than the other to be recommended for use in this case-study. The best solution is the one which has more unit resistance per foundation element, the design and installation method is more reliable, research has provided a more fundamental basis for understanding the method, and if possible, should already been used in the Gulf of Guinea or a similar geotechnical region. Taking into account the above criteria, the more appropriate solution for this case study is considered to be the Suction Embedded Plate Anchor. SEPLA advantages 



Torpedo anchor (DPA) advantages

Cost of anchor element is the lowest of all the deep-water



Simple and economic to fabricate

anchors.



Simple to design.

Uses proven suction caisson



installation methods. 





Accurate to position with no requirements for proof loading.

Provides an accurate measure of



Rapid installation

embedment and position of the



Robust and compact design makes

anchor.

handling and installation simple and

Design based on well-developed

economic with only one Anchor

procedures for plate anchors.

Handling Vessel (AHV) and no ROV.

Experience in the Gulf of Guinea Table 12 – SEPLA and Torpedo anchor advantages.

SEPLA disadvantages

Torpedo anchor (DPA) disadvantages



Patented installation method.



Patented installation method.



Installation time greater than for a



No experience outside Brazil.

caisson.



Lack of documented installation and



Requires keying and proof loading.

design methods with verification



Requires a ROV.

agencies.



Limited field load tests.



Unknown orientation once embedded.

Table 13 – SEPLA and Torpedo anchor disadvantages.

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7. CONCLUSION AND FURTHER RESEARCH 7.1 CONCLUSION This thesis has presented an outlined of existing platform and associated foundation technologies used in offshore developments, and the most important aspects involved in their design. Also, geotechnical characterization issues relating to the offshore environment such as topography, seabed composition and geohazards were discussed. In order to relate the foundation technologies discussed with a real scenario, a specific region was chosen to assess the viability of two different foundation solutions. Proposed offshore developments near São Tomé & Príncipe were chosen for the case study, because it is a region that has recently gathered attention from the oil & gas industry, and has not been subjected to any platform construction yet. Water depths in the zone of proposed offshore development near São Tomé & Príncipe range from 1800 m to 3000 m; thus, future installed facilities must be floating platforms and hence the foundation systems in the seabed will be resisting tensile forces instead of compression. Therefore, the only types of foundation solution suitable for this region are anchoring systems. Anchoring systems that were selected for evaluation in this case are those for which fewer studies and investigation have been made, but on the other hand are likely to be the most economic systems – in this case, torpedo anchors and suction embedded plate anchors were evaluated. Geotechnical characterization of the zones offshore from São Tomé & Príncipe was based on investigations performed in the Gulf of Guinea over more than 10 years. This large database on the behaviour of deep-water sediments made possible the assumption of several soil parameters. The soil in the Gulf of Guinea typically is a highly sensitive clay (St = 2 to 4), and one of the most important parameters is the undrained shear strength profile, it was apparent that it would have a positive gradient with depth of about 1.5 kPa/m. It also became clear that many sites in this region exhibit a greater resistance (up to about 15 kPa) in the first 2 m, t his phenomenon is called a “crust”, and no unique or convincing explanation has been proposed for its existence so far. Other important properties, which are not well defined yet, are the interface soil-steel friction resistance and the set-up effects. The selected torpedo anchor to be employed in this case study would be the same that was used in Albacora Lest Field in Brazil Basin by Petrobras, and it is the T-98 torpedo. The design of the torpedo anchor was based on the design of simple cylindrical pile. Its pull-out capacity comes from three different sources: shaft friction resistance, self-weight of the pile and reverse end

109


bearing capacity. The shaft friction resistance is the fraction that gives the greatest contribution, thus the correct assessment of the adhesion factor (α) plays an important role. The assessment of the torpedo anchor free penetration into the soil is also important to recognise the torpedo embedment, and therefore the soil shear strength along the torpedo shaft. Following O’Loughlin et al. (2013), after releasing the torpedo from 108 m above the seafloor, it should reach a velocity of about 40 m/s before impacting with the soil and the torpedo should penetrate about 40 m into the soil. An adhesion factor, α=0.7 was considered, which corresponds to full reconsolidation of the soil in the vicinity of the pile after being remoulded by the torpedo penetration, consequently the pull-out resistance is expected to be 8.7 MN. The SEPLA solution considered is a 4.5 m x 10 m plate anchor, which is proposed by Wilde et al. (2001) for permanent installations. Plate anchor keying induced loss of embedment was calculated according to Song et al. (2009) and is expected to be about 0.81B (B is the anchor breadth), i.e. approximately 4 m. The holding capacity of the SEPLA is provided by the end bearing resistance plus the selfweight of the anchor. This capacity was calculated according to three different design procedures: 

Wilde et al. (2001)






DNV-RP-E302 (2002)

The most conservative procedure proved to be from DNV-RP-E302 (2002), with a holding capacity of about 9.5 MN. However, this do not take into account the self-weight of the anchor and includes a partial resistance factor, therefore an additional 0.5 MN may be added to provide a total resistance of 10 MN. Based on a criteria that involved the resistance, installation process, experience, knowledge and reliability of both anchoring systems, is was considered that the use of suction embedment plate anchor systems (SEPLA 4 x 10 m2) would be more appropriate offshore from São Tome & Principe.

7.2. FURTHER RESEARCH In terms of geotechnical issues associated with this case study, there is still need for further studies, in particular: 

Site specific characterisation of the seabed in the São Tomé & Principe region, instead of the broad characterisation of the Gulf of Guinea made in this study and based on the amalgamation of various published data from the Gulf generally. 110




Understanding of the origin and cha racterisation of the near seabed “crust” particular to this region and the effect this may have on foundation installation processes.



The interface friction resistance between the soil and the steel elements in the t he short and long-term.

Since there is no experience of torpedo anchors in the Gulf of Guinea, it would be of great interest to develop in situ model scale tests to study the behaviour of the torpedo during penetration of the soil, and its resistance in the short and long-term as well. As this in situ tests are very expensive, it would be very interesting to evaluate the influence of the near seabed “crust”

on the torpedo penetration, using for that computational programs or laboratorial tests. A modulation of both SEPLA and torpedo solutions would be interesting to assess a more efficient evaluation of their potential resistance in STP. A study to evaluate the forces that the platforms will be subjected in the offshore of São Tomé & Principe would be of great interest, because depending on those forces the foundation solutions may be more or less economic, i.e. number and size of the foundation f oundation elements.

111


112


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Appendixes INDEX: Appendix I – Details of Shallow Foundation Studies ………………………………………………. Appendix II – Calculation to determine Loss in Anchor Embedment ……………………….. Appendix III – Calculations to determine the resistance of the SEPLA……………………..

121

136 138 140


122


APPENDIX I – DETAILS OF SHALLOW FOUNDATION STUDIES.

123


124


APPENDIX II – CALCULATION TO DETERMINE LOSS IN ANCHOR EMBEDMENT. INITIAL MOMENT M0 CALCULATION:

 =        ′  =    ×  = 500 9.58×10 = 404.2   = 4.5 × 10×0.2  4×2.2 9 ×0.05×2 = 9.58   =  ×  = 45×1.5 ×24 = 1620  ×   = 2.5/3×0.58 0.05   =     9 0.58  = 0.1   = 2.5 0.1 = 2.6   = 1620 404.22.6 1620 ×0.1 404.2 ×0.05 = 5080.71 . D , /B: ∆ =   .0. 2  .     = 24..65 = 0.58  = 04..25 = 0.04   = 5080.71 = 0.70  45×4. 5 ×24×1. 5 ∆ = 0.580.040.2.0.70. = 0.94 D C:  = ∆  ∙ = 0.94 0.005× 90 = 0.49 ETERMINATION OF THE LOSS IN ANCHOR EMBEDMENT ΔZE

ETERMINE THE CONSTANT

Θ

125


126


APPENDIX III – CALCULATIONS TO DETERMINE THE RESISTANCE OF THE SEPLA.

WILDE ET AL. (2001)

 =  ×  × , × 10.2 × ×  =  × = 10×4.5 = 45   = 12.5 , =  × = 1.5 ×20 = 30   = 0.7  = 1.4,  = 45×12.5 ×30×10.2 × 4.105×0.7 = 12875.6   = 13  ⟹ , =  = 1.134 = 9.2  , assuming full suction is developed

design safety factor

MERIFIELD ET AL. (2001)

 =  ×  × ,    =  × = 10×4.5 = 45   = 500  The ranges of theoretical solutions are:

1 <  < 10 ⟺ 1 < . < 10 ⟺ 1 < 4.44 < 10 0.1 < , < 1 ⟺ 0.1 < .×.  < 1 ⟺ 0.1 < 1.35 < 1

, verifies , does not verify

However, the calculations will proceed.

→      : 127


1.

2.

, = 5   = 14   = 1.5 /  = 20  = 4.5 H/B=4.44

 =56 ,

3. From Figure 93 L/B=2.22 H/B=4.44 4. 5. 6. 7. 8.

 = 8

∗ = 11.16    =  [10.383 ,   1] = 810.383 .×.  ×. 1 = 40.63  =   , = 40.63 × = 96.63 ∗ = ∗ [1 ,] = 11.161 .×  = 78.12  > ∗ ⟹  = 78.12  = 45×78.12×5500 = 18077  = 18  . =  = 2.180 = 9  , for horizontal anchors

If partial safety factors are applied, instead of global safety factor: 1.

2. 3. 4.

, = , = . = 3.6   =  = 11..54 = 1.07 /  = 810.383 ..×.  ×.  1 = 40.6  = 40.6  ×. = 118 ∗ = 11.161 ..×  = 78.1 , = 45×78.1 ×3.6 500 = 13150  = 13.15  128

Offshore Foundations - PSA

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