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Chapter 1 - Accounting in Action Chapter 2 - The Recording Process Chapter 3 - Adjusting the Accounts Question - Answers
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Geotechnical Engineering Handbook
Editor: Ulrich Smoltczyk
•
rnst&Sohn A Wiley Company
Editor: Professor Dr.-lng. Ulrich Smoltczyk AdlerstraSe 63 D-71032 BOblingen
Cover: Campo Valle Maggia, Tessin Instrumentation for investigating an extensive 250 m deep sliding mass, Solexperts AG
This book contains 616 figures and 82 tables Die Deutsche Bibliothek - GIP-Cataloguing-in-Publication-Data A catalogue record for thFs publication is available from Die Deutschen Bibliothek ISBN 3-433-01449-3
It was in the early 1950s that a German consultant in Berlin can1e to the conclusion that structural engineers needed much more guidance on the special prohlerns which they faced on a daily basis due to geotechnical difficulties associated with designing structures_ He discussed this with his professional friends in civil engineering companies, administration and science and with a puhlisher who becan1e quite interested in editing an appropriate "pocket book" about geotechnical matters. This was the birth of the German "GrundhauTaschenhuch" (ground engineering pocket hook). The first edition had already been quite a success for the puhlisher but son1e professionals thought it could he improved_ The editor at that time was assisting his professor of soil mechanics and foundation engineering at
the Technical University in Berlin, who was also a member of the editorial board. He asked me to consider the concerns that had been raised, and as a result of I was given the job of criticizing authors who were n1uch more experienced and prominent than myself. I hope, however, that those authors who are still alive, will forgive the 'youngster' for some of the things he wrote. In suhsequent editions we added material that we thought might provide additional professional help. This, however, made the "pocketbook" expand until today it con1prises three volumes of a handhook that was published at the beginning of the 21st century in its sixth edition. There is a general topic to each volun1e: the first one deals with the fundamentals, the second with geotechnical procedures and the third one with foundation elements and structures. Potential subscribers asked me why I thought they n1ight be persuaded to spend money on a sixth edition when they already had the fifth one? I was glad to point out the fact that firstly, we have been lucky enough to obtain new and famous authors to bring a fresh viewpoint to many of the prohlems, and secondly that the significant harmonisation of design rules in Europe has produced new types of verification procedures due to limit state design which will be new to some practitioners. Recently, globalization has also hecome an essential topic, both in the field of puhlishing and in international civil engineering activities. Ernst & Sohn, Berlin, a publisher of technical literature for more than 150 years, became associated with Wiley of New York, and the question was asked as to why such a handbook on geotechnics was not availahle in English. You are now holding the result of this discussion but we should confess that it has not been an easy job. It was realized that for many of the chapters, a one-by-one translation would not have been appropriate. The authors of the various chapters were therefore asked to review their texts on behalf of the readers outside Gern1an speaking countries and to consider the international state of the art to that extent that would, at the very least, allow further concise guidance to be given by appropriate references. As a book devoted to daily practice of experts, it also had to take account of the considerahle hulk of technical rules already in place, the contents of which should not be repeated simply to fill pages but should be commented on, controversially if necessary.
VI
Preface to Volurnes 1 to 3
Volume 1 starts with an overview of the state of international geotechnical harmonisation, which has been achieved by the civil engineering Eurocode programme in which design is now based on the concept of limit state analysis and the establishment of characteristic values for actions and resistances. Since the editor for more than the last two decades participated in this work, he became well aware of the difficulties raised by the need to find the relationship between conventionally applied geotechnical parameters and characteristic values. Chapter 1.2 is therefore devoted to finding the characteristic values for geotechnical parameters_ The next two chapters deal with field and laboratory testing whilst emphasising the state of knowledge documented in the pre-standard versions of F,urocode 7 - Parts 2 and 3. Chapters 1.5 to 1.9 describe the scientific background and calculation models to be used in geotechnical design, whilst Chapter 1.10 explains how these numerical tools can be used nowadays in design practice.
As surveying has always been a most important method of controlling the performance of geotechnical structures during construction and thereafter - especially when observational methods are used ~the state of modern geodetic know-how, including satellite positioning is covered in Chapter 1.11. To supplement field-testing, Chapter 1.12 gives details of the recent developments in measurement equipment. The special issues associated with defining the actions caused by ice and ice flows are described in Chapter 1.14. Finally, Chapters 1.13 and 1.15 focus on the engineering geology problems of mass movements and rock mechanics problems of slope stability. Volume 2 collects together 14 chapters dealing with the various procedures available for ground improvement (Chapter 2.1), grouting (Chapter 2.2), underpinning (Chapter 2.3), freezing techniques (Chapter 2.4 ), anchoring (Chapter 2.5), drilling (Chapter2.6), driving and pulling (Chapter 2.7), offshore processing (Chapter 2.8), ground dewatering (Chapter 2.9), rock excavation (Chapter 2.10), tube jacking (Chapter 2.11), earth works (Chapter 2.12), application of geotextiles (Chapter 2.13), and engineering biology (Chapter 2.14 ).
Each of these chapters has been produced by authors who are experts in their specific professional field. They outline the most recent developments that have occurred and provide the information necessary for geotechnical designers to select the proper method to achieve their foundation proposals. The broad variety of techniques used required a very concise treatment of the information, often leaving the technical details to those who are especially familiar with these. Volume 3 is concerned with the geotechnical design of structures, starting with spread foundations (Chapter 3.1), pile foundations (Chapter 3.2), and caissons (Chapter 3.3). The application of the new limit state concept is illustrated by examples. This also applies to Chapter 3.4 on the stability of excavations, in which German and British practice are compared. Chapters 3.5 and 3.6 are concerned with excavation pits protected by trench retaining walls or sheet pile walls, and in Chapter 3.7 a general outline of gravity walls is presented. The special aspects of machine foundations and foundations in areas of subsidence are dealt with in Chapters 3.8 and 3.9 and finally the waterproofing of structures is discussed in Chapter 3.10.
Hopefully, the three volumes will enable the practicing engineer to interpret test results in a more meaningful way, to judge the likely limitations of any chosen method with more confidence and to therefore find the most appropriate solution to the foundation problems that he is faced with solving in his daily practice. The object of this handbook is also to close the credibility gap between geotechnical science and practice that is often seen in either type of congress and symposium.
Preface to Volumes 1 to 3
VII
The editor gratefully acknowledges the involvement of the authors who have spent a considerable amount of extra time producing the chapters, over and above their daily professional duties - especially as not all of them are sufficiently familiar with the English language. Where such difficulties arose, the authors were asked to focus on providing the correct translation of their technical terms. The linguistic improvement, was then provided by Robert W. East, of Aylesbury, UK, whose help reviewing the papers is much appreciated. October 2002
Ulrich Smoltczyk
List of contributors
Ulf Bergdahl
Prof. Dr.-lng. Dr. h.c. Gerd Gudehus Tnstitut filr Bodenmechanik und Grundbau lJ niversitiit Karlsruhe POB 6980 76128 Karlsruhe Germany (Chap. 1.5 Constitutive laws for soils froni a physical viet,.t,.•point, Chap. 1.9 Farth pressure determination)
Chief Engineer Swedish Geotechnical Institute 58193 Link6ping S\veden (Chap. 1.3 Geotechnical field investigations)
Prof. Dr.-lng. Peter Gussmann Am Baechle 3 74629 Untersteinbach Germany (Chap. I.JO Numerical methods)
Dipl.-Ing. Christophe Bauduin N.V. BESJX S.A. Avenue des Communautes 100 1200 Bruxelles Belgium (Chap. 1.1 International agreernents, Chap. 1.2 Determination of characteristic values)
Dr. Jan Bohac Department of Engineering Geology Charles lJniversity Albertov 6 12843 Praha 2 Czech Republic (Chap. 1.4 Properties of soils and rocks and their laboratory deternzination) Dr.-lng. Clans Erichsen WBI - Beratende Ingenieure filr Grundbau und Felsbau GmbH HenricistraBe 50 52072 Aachen Germany (Chap. 1.15 Stability of rock slopes)
Prof. Dr.-Ing. Martin Hager Merler Allee 99 53125 Bonn Germany (Chap. 1.14 Ice loading actions) Prof. Dr.-Ing. Gtinter Klein Ostfeldstralle 64a 30559 Hannover Germany (Chap. I.Ii Soil dynamics and earthquakes) Prof. Dr. Edmund Krauter geo-international Mombacher Stra!le 49-53 55122 Mainz Germany (Chap. 1.13 Phenomenology of natural slopes and their rnass movement)
xx
Lisi of conlribulors
Prof. Dr.-lng. Dr. sc. techn. h.c. Klaus Linkwitz Obcrc TanncnbcrgstraBc 24 71229 Leonberg Germany (Chap. 1.11 Metrological monitoring of slopes, embankments and retaining lvalls)
Prof. PhD Ian M. Smith Simon Engineering clo U nivcrsity of Manchester Bruns\vick Street Manchester M13 9PL Great Britain (Chap. 1.10 Numerical methods)
Prof. Dr.-Ing. habil. Dr.-Ing. E.h. Ulrich Smoltczyk Adlerstra!le 63 71032 Biiblingen Germany (l~hap. I.I International agree1nents)
Prof. Dr. Roberto Nova Politecnico di Milano Piazza Leonardo da Vinci. 32 20133 Milano Italy (Chap. 1. 7 Treatrnent of geotechnical ultimate ti1nit states by the theor_v of plasticity)
Dipl.-Ing. Paul von Soos Reullweg 30 81247 Munchen Germany ((~hap. 1.4 Properties of soils and rocks and their laboratory deterrnination)
Prof. PhD DSc (Eng.) Harry Poulos PO Box 125 North Ryde Ne\v South Wales Australia 2113 (C'hap. 1.6 Calculation of stress and settlernent in soil rnasses) Priv.-Doz. Dr.-Ing. Herrmann Schad Reinsburgstra!le lllb 70197 Stuttgart Germany (Chap. 1.10 Numerical methods) Prof. Dr.-lng. Willfricd Sch\varz Am Appelgraben 50 59425 Weimar-Tau bach Germany (Chap. 1.11 Metrological monitoring of slopes, embankments and retaining lvalts)
Dr.-Ing. Frank Sperling Spinoza\vej 12 2202 AV Nordwijk The Netherlands (Chap. I.R Soil dyna1nics and earthquakes) Dr. Arno Thut Solexperts AG POB230 8603 Schwerzenbach Switzerland ( C'hap. I. I2 Cieotechnica{ measurement procedures)
Prof. Dr.-Ing. Walter Wittke \VBI - Beratende lngenieure fi.ir Grundbau und Felsbau GmbH HenricistraHe 50 52072 Aachen Germany (Chap. 1.15 Stability of rock slopes)
Dr.-lng. Fokke Saathoff BBG Bauberatung Geokunststoffe GmbH Alter Bahndamm 12 49448 Lemforde Germany (Chap. 2.13 Geosynthetics in geotechnical and hydraulic engineering)
Prof. h. c. Dr.-lng. Hugo M. Schiechtf' (Chap. 2.14 Slope protection by bioengineering techniques) Prof. Dr.-Ing. Hons-Henning Schmidt Smoltczyk & Portner GmbH Untere Waldplatze 14 70569 Stuttgart Germany
(Chap. 2.12 Earthworks)
Prof. Dr.-lng. Stephan Semprich Prof. Dr.-Ing. Hans-I ,udwig Jessberger"I
(Chap. 3.5 Bored pile walls, diaphragm walls, cut-off walls) o. Prof. Dr.-Ing. Anton Weissenbach Am Geholz 14 22844 Norderstedt Germany (Chap. 3.4 Stahility of excavations)
Contents of Volume 1: Fundamentals Smvltczyk/Bauduin, International agreements Bauduin, Deter1nination of characteristic values Melzer/Bergdahl, Geotechnical field investigations von Soos/BohdC. Properties of soils and rocks and their laboratory determination Gudehus, Constitutive laws for soils from a physical viewpoint Poulos, Calculation of stress and settlement in soil masses Nova, Treatment of geotechnical ultimate limit states by the theory of plasticity Klein/Sperling, Soil dynamics and earthquakes Gudehus, Earth pressure determination Gussmann!Schad/Sn1ith, Numerical methods Linf..r.,·itz/SchH/arz, Metro logical monitoring of slopes, embankments and retaining walls Thul, Geotechnical measurement procedures Krauter, Phenomenology of natural slopes and their mass movement Hager, Ice loading actions Wittke/Erichsen, Stability of rock slopes
Contents of Volume 2: Procedures Kirsch/Sunderniann, Ground improvement Semprich/Stadler, Grouting in geotechnical engineering Witt!Smvltczyk, Underpinning, undercutting Jessberger/Jagow-Klaft!Braun, Ground freezing (Jstermayer/Barley, Ground anchors Ulrich, Dri11ing technology Vt:1n Weele, Driving and extraction de Gijt, Foundations in open water Sn1oltczyk, Ground dewatering Toepj'er, Construction methods for cuttings and slopes in rock Toepfer, Microtunnelling Schmidt/Rumpelt, Earthworks Saathojj; Geosynthetics in geotechnical and hydraulic engineering Schiechtl, Slope protection by bioengineering techniques
Preview
Contents of Volume 2: Procedures KirscWSondermann, Ground improvement Semprich/Stadler, Grouting Witt/Sn1oltczyk, lJndcrpinning and undercutting Jessberger!Jagow-Klaff!Brown, Ground freezing Usternzayer/Barley, Ground anchors Ulrich, Drilling technology van l-Veele, Driving and cxtractlon de Gift, Foundations in open water Riej31Kordonis, Ground \\later fkn.\l and drainage techniques Toepfer, Construction methods for cuttings and slopes in rock Toepfer, :\1icrotunneling Schmidt!Rumpelt, Earthworks Saathoff~ Application of geotextiles Schiechtl, Slope protection by bioengineering techniques
Contents of Volume 3: Structures Smoltczyk/Netzel/Kany, Spread foundations Kempfert/Eigenbrod/Smoltczyk, Pile foundations Lingenfelser, Caissons lVei.j3enbach!Hettler/Simpson, Stability of excavations Stocker/Walz, Trench walls Mazurkiewicz, Sheet-pile walls for harbours and waterways flaack/E1nig, Waterproofing of buildings and structures Klein/Klein. Machine foundations Placzek, Foundations in mining regions Brandl, Slope protection and retainment
Contents of Volume 1: Fundamentals Smoltczyk./Bauduin, International agreements Bauduin, Determination of characteristic values Melzer/Bergdahl, Geotechnical field investigations von Soos/BohtiC, Properties of soils and rocks and their laboratory determination Gudehus, Constitutive laws for soils from a physical viewpoint Poulos Calculation of stress and settlement in soil masses Nova, Treatment of geotechnical ultimate limit states by the theory of plasticity Klein/Sperling, Soil dynamics and earthquakes Gudehus, Earth pressure determination Gussrnann/Schad/Sn1ith, Numerical methods Linkwitz/Schtt1arz, Metrological monitoring of slopes, embankments and retaining walls Thut, Geotechnical measurement procedures Krauter, Phenomenology of natural slopes and their mass movement fl ager, Ice loading actions Wittke/Erichsen, Stability of rock slopes
Contents of Volume 3: Structures Sn1oltczyk./Netzel/Kany, Spread foundations Kempfert/Eigenbrod!Smoltczyk, Pile foundations Lingenfelser, Caissons Weij]enbach/Hettler!Sirnpson, Stability of excavations Stocker/Walz, Trench walls Mazurkiewicz, Sheet pile walls for harbours and waterways Smoltczyk, Gravity retaining walls Klein/Klein, Machine foundations Placzek, Foundations in mining regions Haack/Emig, Waterproofing of buildings and structures
Determination of characteristic values C~hristophe
1 2 2.1 2.2 2.3
2.4 3 3.1 3.2 3.3 3.4 3.5 4
1 3 4 6 6 6 7 7 7 8 12 13 14 16
Rauduin
Introduction ............................................. . From derived value to design value ............................. . Sequential steps ........................................... . Points of view when analyzing test results ........................ . Points of view when determining characteristic values of ground parameters (EN 1997-1, 2.4.5) ................................. . Use of statistical methods .................................... . Examples ............................................... . Local sampling ............................................ . Local sampling with V well-known ............................. . Soil property increasing linearily with depth ...................... . Analysis of shear tests ...................................... . Example: Boulder clay ...................................... . References .............................................. .
17 17 17 19 25 28 37 37 38 39 41 45 49
X
1.3
Contents
Geotechnical field investigations Klaus-Jurgen Melzer and Ulf Bergdahl
Organic content ........................................... . Carbonate content ......................................... . Characteristics and properties of soil aggregates ................... . Fabric of soils ............................................. . Porosity and voids ratio ..................................... . Density ................................................. . Relative density ........................................... . Water content ............................................ . Limits of consistency- Atterberg limits .......................... . Water adsorption .......................................... . Compaction; moisture- density relations ........................ . Size of voids; filters ........................................ . Capillarity ............................................... . Water permeability ........................................ . Air permeability .......................................... . Stress-strain behaviour ...................................... . General considerations ...................................... . One-dimensional compression and consolidation (oedometer) test ...... . Triaxial compression test .................................... . lJnconfined compression test ................................. . Tests with the general state of stress - true triaxial test and biaxial test ... . Ivleasurement of time dependent deformation ..................... . Determination of shear strength parameters ...................... . General aspects of strength testing ............................. . Triaxial compression test .................................... . Determination of unconfined compressive strength and sensitivity ...... . Shear box test ............................................ . Determination of tensile strength .............................. . Determination of slake durability of rock ........................ . Correlations ............................................. . Proctor density and optimum \\'ater content of fine-grained soils ....... . Water permeability ........................................ . Stress-strain relations for soils ................................ . Parameters of shear strength ................................. . Classification ............................................. . Soil classification .......................................... . Rock classification ......................................... . References .............................................. .
133 134 134 134 135 138 138 140
140 144 145
146 147 150 152 153 153 157 164
168 168 169 171 171 176 179 180
182 183 183 183
184 185 187
189 189 197 200
Constitutive laws for soils from a physical viewpoint Gerd Gudehus
1 I.I 1.2 2 2.1 2.2
Introduction . . . . . . . . . . . . Motive and objective . . . . . . Contents . . . . . . . . . . . . . . . States and changes of state . . States . . . . . . . . . . . . . . . . . Changes of state . . . . . . . . . .
Introduction Ohjectives of geotechnical measurements Measured parameters .. Para1neters in the foundation soil Parameters during construction .. Paran1eters in the supporting structure Parameters at adjacent structures ..... Parameters for permanent structures .. Para1neters for the rehahilitation of huildings Measuring instruments. installation and costs ..................... Geodetical measurements ................................... Geotechnical measurements .................................. Execution of the measurements, reporting ........................ Manual measurements ...................................... Auto1natic measuring systems ................................. Data visualisation software ................................... Case histories ............................................ Deep excavations, adjacent structures ........................... Test embankment load, observational method ..................... Adler Tunnel - readjustment of a structure .. Monitoring of unstable slopes .. Test loading of supporting structure, pile tests, displacement measurements in pile foundation .................... References ..............................................
561 561 563 563 564 564
565 565 566 . . . . . . . . . .
. .
566 566 567 587
589 589 590 590 590 601 603 607 611
615
1.13 Phenomenology of natural slopes and their mass movement J::dmund Krauter
Preliminary remarks ........................................ Types of ice loads and ice-structure interactions .................... Properties of ice ........................................... Mass density of ice ......................................... Elasticity of ice ........................................... Thermal expansion of ice .................................... Strength of ice ............................................ Definitive values of the ice strength for calculation ................. 'fhickness of ice ........................................... Calculation of the ice loads ................................... Ice loads on \\1ide structures .................................. Ice loads on narro\\' slender structures .......................... ·Thermal ice pressure loads ................................... Additional vertical ice loads .................................. Ice loads on groups of structures ............................... Ice loads under special clin1atic and ice conditions .................. References ..............................................
. . . . . . . . . . . . . . . . .
669 669 670 670 671 671 672
674 676 677 677
678 682 683 683
684 685
1.15 Sta!Jility of rock slopes Walter Wittke and Claus Erichsen 2 3 4
4.1 4.2 4.3 4.4 4.5 5
6
6.1 6.2 6.3 6.4
6.5 6.6 6.7 6.8 7
7.1 7.2
Introduction ............................................. . Structural models of rock mass ................................ . Mechanisms of failure of rock slopes ............................ . Model for the stress-strain behaviour of rock ...................... . General ................................................. . Intact rock ............................................... . Discontinuities ............................................ . Rock mass ............................................... . Model for the mechanical behaviour of a rock nlass \\'ith respect to a refined stress displacement behaviour of persistent discontinuities with no fillings . Model for seepage flow through a rock mass ...................... . Stability investigations according to the finite element nlethod ......... . General ................................................. . Computation of stresses and displacen1ents ....................... . Computation of a seepage fto\\' ................................ . Presentation and interpretation of the computed results ............. . Influence of shear parameters of discontinuities on the stability of a slope . Support of a slope \\'ith prestressed anchors ....................... . Influence of high horizontal in-situ stresses ....................... . Stability investigations on the wall of a construction pit using a refined conceptual model of the mechanical behaviour of a rock mass ......... . Stability analysis on the basis of rigid-body mechanics ............... . General ................................................. . Possibilities of translation and rotation of rock nlass wedges ........... .
687 688 693 696
696 698 698
701 705 707 712 712 712 716 718 720 723 725
731 735 735 735
XVIII
Contents
7.3 7.4
Stability analysis of planar rock mass wedges.. . . . . . . . . . . . . . . . . . . . . . Stability analysis of three-di1nensional rock mass wedges supported
Interaction of suction and recharge wells .. References . . . . ..... .
398 398
2.10 Construction methods for cuttings and slopes in rock Axel C Toepfer
I 2 2.1 2.2 3
3.1 3.2
4
Introduction ............... . Cuttings in rock ........... . Mechanical loosening by ripping . Loosening by blasting methods ... Construction method for rock slopes. Mechanical construction method for the production of rock slopes ...... . Smooth blasting methods ............. . References ....... .
399 400 400 403 417 418 418 427
2.11 Microtunnelling Axel C Toepfer
I 2 21 2.2 2.3 2.4 2.5 2.6 2.7 J
Introduction .. 'Ibe microtunnelling construction method for non-man-sized entry pipes 'Il1e components of the construction method Description of soil and rock .. Pipe material .............. . Microtunnelling system ..... . Driving and reception shaft .. Construction sequence .. Further development References
429 430 430 431 431 432 437 438 440
440
2.12 Earthworks Hans-Henning Schmidt and Thomas Runzpt!lt I 2
3 4 4.1
4.2 5 5.1 5.2
5.3 5.4 5.5
5.6 6 6.1 6.2
6.3 6.4
Introduction ...... . Standards. environmental legislation ... Terms and definitions . Construction materials, classifications and characteristic values Gcrneral introduction ..... _ ....... _ . Characteristic parameters ........ _ ....... . Design of earthwork structures . Site investitgation .......... . Design calculations ... . Standardised slope angles ........ _ ..... __ Assessment of the stability of slopes _ . Drainage measures for earthworks ...... . Landscape planning ....... . Earthwork processes/earthworks equipment .. . Machines for digging, transporting and placing Loading with hydraulic excavators Ilauling equipment Equipment for placing and spreading
1.1 International agreements Ulrich Smo/tczyk and Christophe Bauduin
1
Classification of geotechnical literature
The International Society of Soil Mechanics and Foundation Engineering (ISSMFE) have since 1981 used a classification scheme of geotechnical literature (IGC) as in Table 1.
There has also hccn a joint agreement \\.'ith the S1;vedish Gcotcchnical Institute (SGI) since 1998 1,vhich enables access to their Information Retrieval System for Geotechnical Literature (IRS-Geo) available 1,vorld\\.·ide via the Internet. Their continuously updated database provides about 54,000 literature references from 1976 onwards. Each reference is dcscrihcd by JGC key \\.'ords and classification codes (for combining key 1,vords, use and, or and not). Subscription is available via the SGI 1,vebsite at http:/h.\1\VW.S\\'cdgco.sc/ indcx-c.html. The references are 69 o/o English, 16 °/o S\\·edish, 6 % German, 4 °/o French and 5 °,6 with other tongues. Table 1. International geotechnical classification system, published in March 198"1 A General
B
A A A
B
A A A A A
A A
1 Geotechnical Engineering-Scope 2 Historical Aspects 3 lnfonnation Services, and I ,iterature Classification 4 Textbooks, J--Jandbooks, and Periodicals 5 Terminology 6 Companies, Institutes, and I ,aboratories 7 Societies, f\1eetings, and International Cooporation 8 Professional Ethics, Legal Requirernents., Codes of Practice, and Standardization 9 Education 10 Research Activities
B Geological and Environn1ental Aspects (Basic Geology, see Principal Group T) B Il B D B
0 1 2 3 4
B B
General Formation of Soil and Rocks Hydrogeological Aspects Mass Movements and I.and Subsidence Seisn1ic Activity and Crustal Movements 5 Clin1atic Conditions 6 Submarine Geological Aspects
B
7 -
Il B B
8
Extraterrestrial Soil and H. ock Conditions 9 Geomorphologic Aspects and Terrain Classification 10 Mineralogical Aspects 11 Description of H. egional Soil and Rock Conditions 12 Other En,,.·ironrnental Aspects
C Site Investigations Equipment and Techniques of Exploration, Prospection, Sampling, and Field Testing of Soils and !locks (excl. detennination of engineering properties), Presentation of Result C
C C C C C C
0 I 2 3 4 5
C
6 7
C
8
C
9 10
C
General Airphoto Sur,,.·eys and llernote Sensing Geophysical Surveys Probings (Soundings) Visual Exploration Techniques Boring Techniques and Equipment (cf. CIO) Sampling Measurement of Pield Conditions (incl. Post-Construction Monitoring) Pield Testing (excl. tests for engineering properties, see Groups D and F) Presentation of Results, Data Base Underwater Site Investigations
2
Ulrich Srnoltczyk and Christophe Bauduin
D Soil Properties: Laboratory and In-Situ Determinations (incl. Rockfill, Artificial Soils, Waste Materials) Concepts, Theories, Methods of Detern1ination, Equipn1ent, and Results D D D D
0 I 2 3
D D D
4
D D D D
5 6 7
8
9 10
General Classification and Description of Soils Physico-Che1nical Properties Composition, Structure, Density, and Water Contents Hydraulic Properties Compressibility and S\\.'eliing Shear-Deformation and Strength Properties Dynamic Properties Thern1al Properties Co1npactibility Properties of Soil-Additive Mixtures
E Analysis of Soil-Engineering Problems Theoretical, En1pirical, and Practical Methods of Analysis E E E E E E E E E E E E E E E
0 General I Stress Analysis 2 Deforn1ation and Settlement Problen1s 3 Bearing Capacity of Shatto\\' Foundations 4 Bearing Capacity of Piles and olher Deep 1-0undations, Anchors 5 Earth Pressure Problen1s 6 Stability of Slopes and Excavations 7 Seepage and other Hydraulic Problems 8 Dynan1ic Problems 9 Frost Action and Heat-Transfer Problems 10 Analysis of Layered Systen1s and Pavements Behaviour 11 Soil~Vehicle and Soil-Tool lnteraclion 12 Soil-Structures Interaction 13 Mathematical Methods, Con1puler Analysis 14 Model Test Analysis
F Rock Properties: Laboratory and In-Situ Detenninations Concepts, Theories, Methods of Detenninalion, Equipn1ent, and Results F
F F F F
General Classification and Description of Rocks and Rock Masses 2 Physico-Chemical Properties 3 Composilion, Density, and Slruclural Features 4 Hydraulic Properties 0 I
F F F F
5 6
Con1pressibility and Swelling Shear-Deformation and Strength Properties 7 Dynamic Properties 8 Thennal Properlies
G Analysis ofRock-Enginee.-ing Problems 1beoretical, Empirical, and Practical Melhods of Analysis G (i
G G G G G G G (; (J
G G G G
0 I 2
General Slress Analysis Deformation and Displacement Proble1ns 3 Bearing Capacity of Rock Masses 4 5 Rock Pressure on Tunnels and lJnderground Openings 6 Stability of Rock Slopes and Open Excavations 7 Seepage and other Hydraulic Problems 8 Dynan1ic Problems 9 Frost Action and Heat-Transfer Problems 10 II 12 Rock-Structure-Interaction 13 Malhen1atical Methods, Computer Analysis 14 Model Test Analysis
H Design, Construction, and Rehal'iour of Engineering Works Description and Case Records of Engineering Works H H H H H H H H H H H H
0 General 1 Foundalions of Structures (other than dams) 2 Relaining Structures and Cut-off Walls 3 Offshore Slruclures 4 Dams and Reservoirs, E1nbankments s Tunnels and Underground Openings 6 Roads, Railroads and Airfields 7 Harbours, Canals, and Coastal Engineering Works 8 Conduits and Culverts 9 Slopes and Unsupported Excavalions 10 Land Use 11 Waste Depositories
K Construction Methods and Equipment K K K K
0 1 2 3
General Drainage Melhods Sealing and Grouting Processes Preloading and Soil Replacement
3
1.l International agreernents K
4
K K K K
5
K
9
K
10
K K K K
~I
M M M M M M M M M
6 7 8
11
12 13
14
Earthworks and Rock Excavation, Processing and Transportation Co111paction Processes Soil Stabilization and Erosion Control Piles and Pile Driving, incl. Sheet Piles Construction of Caissons and Deep Piers Construction Methods for Shallow Foundations Slurry-Assisted Construction of Foundations and Cut-off Walls Support of Soil and Rock, Anchoring Offshore Construction Protection Measures against Frost Measures for hnproving Deforn1ation and Stability Conditions. Reconstructi on of Foundations
Materials of Construdion* 0
General
1 Steel 2 Wood 3 4
Bituniinous Materials Plastics and Similar Materials 5 Cen1ent and Chernicals 6 Concrete 7 Paints and Coatings 8 Construction Elernents
2
S Snow and Ice Mechanics and Engineering
s s s s
0
General
1 Snow and Ice Cover 2 Properties of Snow and Tee 3
Snow and Tee Engineering
T Related Disciplines* T T T T T T T T T T T T T T T T
0 General I Pure Sciences 2 Geosciences 3 Agriculture and Pedology 4 Meteorology and Climatology 5 Biosciences 6 Civil Engineering 7 Mining Engineering and Ore
Prospecting 8 Mechanical Engineering 9 Electrical Engineering 10 Ocean Engineering 11 Military and Naval Engineering 12 Instrumentation and Measuring Techniques 13 Library Science 14 Environment Problerns and Nature Conservation 15 Oil Prespecting
*·The principal groups Mand Tare not to be used with "Geotechnical Abstracts"
Symbols
Existing international agreements on symhols are not always consistent. For civil engineering purposes an international standard, ISO 3898 Basis fur design uf structures Notations - General symbols, was puhlishcd in 1997 and was accepted by the European Commission of Standards (CEN). lhe standard for SI units, ISO 1000: 1992(E), was also accepted hy CF,N. For geotechnical literature. the International Society agreed in 1977 to apply the following symbols: A B Ca C.:
c,
C" D E Eo~d
F G
area breadth of foundation rate of secondary consolidation compression index swelling index uniformity coefficient depth of foundation beneath ground inodulus of linear deforn1ation oedometric modulus factor of safety modulus of shear deforn1ation
le
IL Io Ip K Ko K" K,
N N,
consistency index liquidity index density index plasticity index rnodulus of con1pressibility coefficient of earth pressure at rest active earth pressure coefficient passive earth pressure coefficient blo\v count bearing capacity factor as a function of c
4
l,'Jrich Srnoltczyk and Christophe Bauduin
Nq
N..,.
o, Q,
R
s,
s,
T, l}
v a a c' c'R Co
c,
c, d
ct e e g
em ax emin f, h le, lq. ly
j
k
k, m n
q q,
q, qp q, q, s u
3
bearing capacity factor as a function
llw
of e111bedn1ent depth hearing capacity factor as a function of weight density y point resistance total shaft resistance residual factor degree of saturation
Ua
sensitivity time factor degree of consolidation volurne wall adhesion acceleration effective cohesion intercept residual cohesion intercept apparent cohesion intercept remoulded undrained shear strength coefficient of consolidation drainage path grain diarneter eccentricity void ratio acceleration due to gravity void ratio in loosest state void ratio in densest state local side friction hydraulic head or potential hydraulic gradient inclination factors seepage force coetficient of permeability modulus of suhgrade reaction mass porosity rate of discharge static point resistance (C~PT) dynamic point resistance point resistance pressure unit shaft resistance limit pressure settle1nent pore pressure
6
v
pore water pressure pore air pres.sure discharge velocity
w
water content
WL
liquid limit plastic limit shrinkage limit angle of slope to horizontal inclination of load angle of wall friction linear strain principal strains linear strain rate effective angle of internal friction apparent angle of internal friction residual angle of internal friction shear strain shear strain rate weight density weight density of submerged soil weight density of dry soil \\'·eight density of solid particles weight density of \\'·ater coefficient of viscosity coefficient of friction Poisson's ratio rnass density of soil mass density of submerged soil rnass density of dry soil mass density of solid particles rnass density of water total normal stress effective normal stress principal stress octahedral nonnal stress effective overburden pressure preconsolidation pressure shear stress octahedral shear stress shear strength residual shear strength average shear strength rnobilized along sliding surface
Wp \\'S
~
0
c fl,2,'3
'cp' or qi' rru or tf>u
lf.'R or (j)R y y
y y' Yd
Ys Yw I]
µ v p
µ' Pd
µ, Pw a
o' 01,2.3 Oocl
0 'vo
a~
'
,,toct 'R T
International rules for foundation engineering
The only internationally established codification of rules for foundation engineering purposes that existed until now \vas produced by the European C'.ommission of Standards (C'EN), the aim being the introduction of a consistent system of common technical rules for the design and construction of structures in the field of civil engineering. Eventually, this will replace the differing national standards that exist in the various member states of the European (~ommunity. By removing the problems caused by the national standardiza-
5
1.1 International agreements
ti on of technical rules, a common market of products and services will be established. 'This will improve the competltional capability of European contractors and their consultants when working in countries off the European Community. Structural Eurocodes generally include codified assessments for the design of structures_ Construction and supervision are only covered as required for quality assurance checks of the designer's assumptions_ At present, the following documents are published or in preparation: Technical Committee 250: EN 1990 Basis of design for structural Eurocodes EN 1991
Actions
EN 1992
Concrete structures. Part 1: General rules, ... Part 3: Concrete foundations
EN 1993
Steel structures. Part 1: General rules .... Part 5: Steel piles
EN 1994
Mixed steel and concrete structures
EN 1995
Timber structures
EN 1996
Masonry structures
EN 1997
Geotechnical design_ Part 1: General rules, Part 2: Laboratory and field testing
E;..J 1998
Seismic actions. Part 1: General rules, ... Part 5: Foundations, retaining structures and geotechnical aspects (in addition to EN 1997-1)
E~
1999
Design of aluminium alloy structures
Technical committee 288: EN 1536 Execution of special geotechnical work: Bored piles
EN 1537
Execution of special geotechnical work: Anchors
EN 1538
Execution of special geotechnical Vi.'Ork: Diaphragm walls
EN 12063 Execution of special geotechnical Vi. ork: Sheet piles 1
E;..J 12699 Execution of special geotechnical work: Displacement piles E~
12715 Execution of special geotechnical work: Grouting
E~
12716 Execution of special geotechnical work: Jet-grouting
E~
.. .
EN .. .
Execution of special geotechnical work: Micro-piles Execution of special geotechnical work: Reinforced soil
Technical committee 182:
By introducing existing ISO documents into the CEN system: ISO 14688 Identification and classification of soils ISO 14689 Identification and classification of rock Technical committee 189: Several standards for testing geotextiles have been published, see Chapter 2.15.
Ulrich Smoltc:1.yk and Cluistophe Ilauduin
Note: Developing European Codes fron1 a first draft into an officially accepted document generally takes several years time. It is recommended therefore that the national standardization bodies or the appropriate CEN secretariat are asked to confir1n the latest situation for each code (as to TC 250, contact NNI, POB 5059, NL-2600 GB Delft, Fax +31152690190. As to TC 288, contact AFNOR, F 92049 Paris La Defense Ccdcx).
Basic terms by EN 1990 and EN 1997
4 4.1
Classification of assessments in Eurocodes (EN 1990, 1.4; EN 1997-1, 1.3)
Depending on the character of the individual assess1nents, distinction is 1nade in EN 1990, 1.4, between Principles and Application Rules. The codes include a main text and annexes: l'rinciples are rules to which no exetnption or alternative is pern1itted. Application Rules are internationally acknowledged rules that arc recommended for application. In this case alternatives are possible, if they provide the san1e level of reliability and are consistent to the relevant principles. They can include the use of national specifications. Annexes contain additional information, including numerical values of partial safety factors. The values suggested in an annex may either be accepted or modified by national determination as the level of responsibility for the reliability of the works remains with the national authorities in charge of civil engineering control.
4.2
Limit states (EN 1990)
According to EN 1990, 6.4.1, in EN 1997-1, 2.4, the following limit states are defined:
Ultimate limit states (2.4.7.1): • Loss ofequilibriu1n of the structure or the ground, considered as a rir;id body, in lvhich the strengths of the structural materials and the ground provide insignificant resistance (EQU). • Internal failure or excessive deforTnation of the structure or structural elenients, including footings, piles, basement ivalls etc,. in which the strength of structural nzaterials is significant in providing resistance (STR). • F'ailure or excessive deforniation of the ground, in which the strenxth of soil or rock is significant in providing resistance (GF.0). • T.oss of equlibritan of the structure or the ground due to uplift by )-1,.'ater pressure (UPL). • Hydraulic heave, internal erosion and piping in the ground caused by h _vdraulic gradients (HYD).
Serviceability limit states (2.4.8): States when defor1nations, displace1nents or any non-structural damage affect the intended function of a structure in terms of comfort and appearance, where "appearance" is concerned with, for exan1ple, extensive cracking rather than with aesthetics.
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Anticipated deformations and settlements arc the main consideration here. For these. Annex Hof EN 1997-1 gives some guidance (see also Chapter 3.1 of Volume J of the Hand hook). EN 1990, 2.3, also requires that the function of the structure shall be warranted during its design \\'Orking life. For example, in norn1al buildings Table 2.1 of EN 1990 indicates a design life time of 50 years.
4.3
Design situations (EN 1990, 3.5)
In relation to a specified time interval the following design situations are identified:
• persistent design situations which comply with the normal use of a structure; • transient situations \\'hi ch refer to te1nporary conditions as for exa1nple during execution or repair; • accidental design situations by fire, explosion, in1pact or localised failure etc; • seismic design situations. Note: A 'design situation' constitutes a cotnplete scenario, cotnprising a number of various load cases and load com hi nations.
4.4
Geotechnical categories (EN 1997-1, 2.1)
To describe the niinin1un1 require1nents for the extent and content of ground investigations, design analyses and site supervision and the risks to property and life, three different geotechnical categories 1nay be applied as follo\\'S: GC 1: Small and relatively simple structures for which hasic stahility and performance require1nents can be fulfilled hy experience and qualitative ground investigation and for which risks are negligihle. Examples arc sin1ple one- or two-storey buildings, storage sheds, garages.
This classification assumes that ground conditions arc known hy experience. The design of the building should therefore be routine and straightforward. Rules "·hich help to classify a structure into GCl arc determined nationally. GC 2: Conventional structures and foundations that can be designed by routine geotechnical procedures. These should normally include quantitative analyses to verify that the fundatnental design require1nents are satisfied. Some guidance on the extent of ground investigations is given in EN 1997. 3.2. The decision, ho\\'ever, should be supported by local or regional experience. GC 3: All circumstances other than those covered hy GC 1 and 2.
Note: All parts of a project need not to be classified in the same category.
4.5
Observational method (EN 1997, 2.7)
It is often difficult to make a reliable prediction about the performance of a structure and in such cases an observational 1nethod 1nay be appropriate, in which the design data and
8
Ulrich Sn1oltczyk and ('.hristophe Bauduin
assumptions can he controlled during the execution and adapted if necessary.
This method makes the following assumptions: • the allowahlc values of the performance parameters shall he assessed prior to design;
• the potential variance of performance shall be deter1nined such thnt the real performance remains within the predicted limits by a sufficient level of reliability; • the type and extent of monitoring shall be pre-determined; • remedial measures shall be included in design and contract to allow for situations when allowable limit values become surpassed.
Smoltczyk [21 commented on the prohlcms of this procedure and emphasized that a reaction in due time requires the consideration of the follo\.ving: Activating remedies needs some time which should he considered during design. The method is not suitable for situations where reaction will most probably not be possible in time even though alarm signals have hccn recorded. Tt should not he used therefore where brittle failure phenomena may occur. The method should only be applied in situations where assurance can be given that a proper reaction can be implemented prior to an ultimate li111it state occuring. The method should not he ahuscd as a cheap protection against catastrophes. It should be understood as a means to provide economic alternatives within serviceability scenaria. The monitoring schc111c should he redundant. Rcdundancc, however, should he provided by applying independant methods rather than by multiple records. Tt is not only economy hut also the practicality using large numhcrs of recorded data which should make the designer think carefully about the necessary amount of gauges and readings. It is the sudden change of a recorded parameter or its rate rather than the ahsolutc magnitude which gives the warning of a deviating pertormance. Parameters taken for monitoring arc mainly displacements, strains, pore pressures and water levels (Peck ll]).
4.6
Portiul sufety fador method
According to EN 1990, section 6, the verification of limit states shall be achieved by applying partial factors of safety as follows: • the representative values of actions, Frep (see 4.6.1), multiplied by a partial safety factor yf; • the characteristic values of material strength parameters, Xk, divided by a partial safety factor Ym; • the characteristic value of a resisting force, Rk, which is determined directly, divided hy a partial safety factor YR· Thus the design values (index "d") arc derived by Fd = Frep · Yf and Xd = Xk/Ym or Rct ~ RkiYR· The design values for the effects of actions in structural clements (moments, shear forces, normal forces), Ed. are then ohtaincd hy either a statical calculation • based on Fct and Xct or R<.1. or • based on Frep and R1, multiplied hy the partial safety factor Yf·
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lJncertainty about the calculation model of actions may he allowed for hy an additional model !actor, Ysd, which in the first case (see EN 1990, 6,3,2, eq, 6,2) effects of actions are multiplied hy. In the second case (see EN 1990, 6.3.2, eq. 6.2a) this uncertainty is covered by the partial safety factor YF = YSd · Yf· A similar rule with a model factor YRd is included for resisting forces (e.g. earth resistance) such that Yfl.t = YRd ·Ym· This model factor however, is rarely applied in geotechnics because nor1nally Ym will be assessed to cover this uncertainty, too. For the satne reason, in EN 1997-1 no special conversion factor 11 (see EN 1990, 6.3.3) is required to consider effects such as load duration, 1nodel scale, te1nperature etc. Unless required hy national deter1nination, partial factors for accidental situations and for limit states of serviceability shall equal 1.0. The design values of geo1netrical data (a) are generally represented hy a notninal value
In situations where possible or actual deviations of geometrical data would cause a significantly adverse effect on the reliahlity of the predicted limit states, this shall he taken into account by adding a safety margin:
4.6.1
Representative value of an action
According to EN 1990, 6.3.1, the representative value of an action is generally given hy
where \j.! = 1 for - the characteristic value of permanent unfnvourable actions: Gk; - the characteristic value of a variable unfavourable action, Q1, which shall be analysed as heing the governing variable action to he taken at its full magnitude. For persistent and transient situations the other variable unfavourable actions, Qi (i > l) shall be reduced hy a cotnhination factor 1p = 1~0 < 1, to allow for the fact that all variable actions will not occur simultaneously by their maximum value. The representative value of an action for an accidental situation is also combined, but hy applying special reduction factors 1Jl11 (frequent value) for Q1 and 1J-'2i (quasi-perrnanent value) for Qi > !,associated with a time interval (EN 1990, 6.4.3). In addition, the nominal value of an accidental action, Ak, is to he taken into account (e.g.: collision force). For earthquake situations, the representative value consists of Gk, the variahle unfavourable actions Qkh reduced hy ·41zi and an earthquake action AE. The nu1nerical values of these reduction 1p-factors are separately tahled in the Annexes of EN 1990, for buildings and traffic structures, For example, 'i'u ~ 0,6 applies to wind as an accompanying action with traffic structures. If it is not obvious which variable action is the governing one, then each relevant variable action shall be analysed in turn as Qi.
l0
lJlrich Sn1oltczyk and C:hristophe Dauduin
4.6.2
Load cases (combinations of actions)
The combinations of actions concerning the various verifications of ultimate limit states arc collected in EN 1990, 6.4.3, and in 6.5.3 those for serviceability lin1it states. These combinations are based on the representative values of actions, both from the structure and the ground. In foundation engineering, where actions fro1n the structure normally act in conjunction with actions from the ground, the following equations apply: 1. Ultimate limit states for permanent and transient design situations LYG:j. Gk;j j::-:_l
+
YQ;1 .
Qk;l
··+"
,LYo:i · 1+1o;i · Ok;i i::-:_l
where
''+" means in combination l-1.'ith; Gk - characteristic value of a pern1anent action from the structure and/or fron1 the ground such as earth pressure or \\ ater pressure. According to EN 1997-1, for the 1
weight density of the soil YG:j = l applies (see Table A.2.2 of the Code);
Ok - characteristic value of a variable action from the structure and/or from the ground allowing for the representative values n1entioned in 4.6.1. Applying these combination factors, the different load combinations are obtained by substituting Ok for 0 1 to detern1ine the governing one; YG - partial safety factor of G with distinct numbers for unfavourable (YG:sup) and favourable (YG:u>f) actions. It should be noted that the factor YG = 1.35(EN1990, Annex Al.3.1 and EN 1997-1, "!able A.2.1) with a minor factor of about 1.1 covers possible uncertainties of weight. The larger factor takes account of uncertainties resulting from load re-distributions during construction and thereafter; YQ - partial safety factor for unfavourable variable actions 0 with a value 1.50 for any ultimate limit state (EN 1990, Annex Al.3.1, and EN 1997-1, Table A.2.1). J-<'avourable variable actions shall not be considered.
No special reduced values for transient design situations during construction arc given in EN 1997-1. The introduction of such values is permitted by national standards because reductions of this kind will depend significantly on distinct regional construction methods. 2. Ultimate limit state for an accidental design situation LGk:i "+"Ad"+" (1Jl1;1 oq1z,1) · Qk:I "+" L'l'2;i · Q,,, j>l
i>l
where Ad
- the design value for the accidental action that shall be assessed as a nominal value
by contract. or lt)2 ; 1 shall be determined according to the type of accident or its consequences (EN 1990, 6.4.3.3). This combination also applies to the situation after an accident when A = 0.
1t1i: 1
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l.1 International agreements
3. Ultimate limit state for an earthquake design situation
L Gk;j ,,+" AE;d ,,+" L ~'2;i. Qk;j j:.::_l
i:.::_l
where AE;d-
the design value of the action caused by an earthquake that shall be determined by EN 1998 (see also chapter 1.8).
4. Irreversible limit state of serviceability (characteristic combination) LGk;j ,,+"Ok;!,,+" LlJlO;i. Qk;i j>l
5. Frequent limit stale of serviceability (see 4.6.1) LGk;j ,,+·· 'l'l;l. Ok;l ,,+" L>Jl2;i. Qk;i j~l
i>l
6. Quasi-permanent limit state of serviceabilit)' (see 4.6.1)
L Gk;j ,,+" L >Jl2:i. Qk;i j:.::_l
4.6.3
i2::1
Geotechnical verification of ultimate limit states
Verification n1ethods defined in EN 1997-1 for ultimate limit states STR and GEO differ depending on where the partial safety factors arc applied in the course of analysis. There are principally two possibilities to introduce partial factors: either on the input data of the calculation models or on their output data. In the first case this applies to the material parameters (shear strength, concrete strength, yield strength of steel etc. material factor approach, MFA). Jn the second case this applies to the output from the model used to calculate a ground resistance (pile load capacity, bearing capacity, earth resistance etc. - resistance factor approach, RFA).
As calculation models in geotechnics depend either linearily on the shear strength (sliding, skin friction, slope stability of undrained cohesive ground) or non-lincarily (earth pressure, bearing capacity, slope stability of drained ground), different design results are obtained when a foundation clement is sized by each procedure. For this reason, the choice of one of the verification procedures indicated in EN 1990, A.1.3.1, with appropriate partial factors given in EN 1997-1, Annex A, remains with the national standardization bodies. Three alternatives are offered for the STR and GEO ultin1ate limit states: • Design approach 1: Two verification types are required; (1) based on factored actions and non-factored shear strength paran1eters and (2) factored shear strength paran1eters and non-factored permanent actions with variable actions factored by YO = 1.3. The sizing and positioning of the foundation elen1ents depends on both types, with the more adverse to be adopted. The factors are given in Annex A of EN 1997-1, Tables A.2.1 and A.2.2. (1) provides safety against adverse deviations of the actions from their characteristic values and (2) provides safety against adverse deviations of the shear parameters.
12
lJlrich Smoltczyk and Christophe Bauduin
• Design approach 2: Uses only one type of analysis, based on characteristic values. Actions are then factored as in approach 1(1) and resistances are divided hy partial factors given in Annex A, Tables 23, of EN 1997-L • Design approach 3: Again uses one type of analysis, based on design values of structural actions, applying the same partial factors given by approach 1(1). On ground actions and resistances partial factors are applied as in approach 1(2).
For ultimate limit states EQU, UPL, HYD straighforward procedures (EN 1997-1, 2.4.7.2, 2.4.7.4 and 2.4.7.5) are required similar to the traditional global safety concept Equal partial safety factors for the three approaches are recommended in EN 1997-1, Annex A (Tables A.I, A.3 and A.4) but depend, of course, on national requirements and standards. 4.6.4
Verification of the limit state of serviceability (SLS)
The limit state of serviceability can be checked by proving that a limit value, Cct,s, of a chosen quality of the structure does not occur therehy confirming the normal use of the structure (EN 1997-1, 2.4.8). The analysis may be based on checking limit values of action effects, settlements, displacements, tilting angles, accelerations etc. For the definition of 1novement 1nodes see EN 1997-1, 2.4.9, and Annex H which give some limit values (for further details, see chapter 3.1 of volume 3 of the Handhook). Design values Normally, design values for the SLS are equal to the characteristic values (see 4.6.2). However, partial factors> 1.0 may be appropriate if a deterioration of soil qualities or a change of boundary conditions cannot be excluded during the lifetime of the structure, see EN 1997-1, 2.4.8(3). If the check on the limiting values of deformation or movement is not required, it may be sufficient in si1nple cases and hased on coin parable experience, to prove that the level of mobilized shear strength in the ground is sufficiently low (EN 1997-1, 2.4.8(4)).
5
Geotechnical report
A geotechnical design report and a ground investigation report are detailed in EN 1997-1 for the documentation of soil investigation results and the appropriate conclusions.The infor1nation required for these reports is summarized in clauses 2.8 and 3.4. Tbe following notes should be read as co1n1nents with acco1npanying re1narks by the second author (see [J]). The aim of a geotechnical report generally, is the expert desciption of geotechnical conditions, pre1nises and assumptions for the design and construction of a structure classified in category 2 or 3_ Expertise on a category 1 structure will normally only be made when this classification looks dubious or if it is required by court proceedings for example. The statements in the geotechnical report may a) he prepared for a project during design, b) acco1npany a project with site consulting,
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c) provide guidance during construction in terms of supervision,forexample hysuggested monitoring etc, especially when the observational method is applied. At the stage of contract negotiations (a) normally is considered, whilst (b) often becomes necessary during construction. It is recommended that the layout of the report is retained in a way that allow later supplements to be added easily in a clear and logic order. With very extensive projects such as traffic lots, the official request of public consensus only needs a rather general description of the ground conditions. In these cases, distinction can reasonably be made between preliminary investigations (see EN 1997-1, 3.2.2) and design investigations for individual structures and problems (see EN 1997-1, 3.2.3). This may even be done by different consultants.
5.1
Ground investigation report (EN 1997-1, 3.4)
The ground investigation report should start by explaining the reasons for the investigation. This is especially important in cases where facts have become ohvious during investigation which were not realized at the time when it was planned and contracted. To manage these situations, a list of unit prices for tests and services should he in the contract. Following EN 1997-1, 3.4(2) this report should contain • a summarizing documentation of the investigation results and the methods applied with reference to EN 1997-2, • a critical commentary on the results and the parameters derived from them. It is recommended that the expert who does the test supervision is also involved in writing the report to ensure a consistent description is produced. When the general situation is explained, it should be said whether the investigations had to be done in an area complety unknown until now or if reference was possible to previous geological findings and earlier investigations. Furthermore, valuahle general information may have been obtained from people who are familiar with the environment (for example: "a hunderd years ago, there was a lake here", "this area was used for dumping for a long time'', or "the ground water level was pumped down" etc). Next. the report should contain the soil succession, the spatial geometry of the soil layers and the surface, the ground and free water levels and the ilow rates and directions and the existing structures and their recognizahle sensitivity to excavations or any other change of topography. Special risks such as creeping slopes, geological faults, changing water levels, erosion phenomena should be included wherever possible. All of these more general remarks should be checked to determine whether or not they are relevant for inclusion in the design report. Arguments should be given for risks that can he neglected, although the public is frightened about a risk-bearing phenomenon. The description of the soil situation to a large degree is already an expert interpretation because it provides a suggested spatial coordination hetween successions investigated at singular locations. It is therefore always possible (especially from a scientific point of view) to question this interpretation. This does not mean however that it should not be tried, since an expert description should provide overall information of the ground character for contractual purposes - it is reasonable to clarify in the report what is fact and what is supposition. This may be indicated hy introducing classes of reliability:
14
lJlrich Smoltczyk and Christophe Bauduin
Class 1: Areas where interpolation between investigation points is easy and, therefore, possible to a very high degree of probability. Class 2: Areas with re1naining uncertainty although the available facts appear to allow interpolation. Class 3: Areas where the results allow distinct interpretation which necessitates assutnptions based on facts derived from additional sources of information (for example: evaluation of geological statements). Class 4: Areas where additional investigations are necessary, as available information allows only limited assutnptions to be 1nade. Records which contain the results of the visual inspection of bored cores by an expert soil engineer or geologist and the details of sounding or test pit inspections should he included in an annex to the report. This nlakes the nlain text clearer where not all details need to he considered. The report should rather clarify which details listed in the annex arc of 1najor in1portance in understanding the total situation. As an exa1nple. a thin cohesive interlayer, detected in but a few borings, would be insignificant when mainly normal stresses are nlobilized for the bearing capacity of a foundation en1bedded in plane terrain, but would have to he taken into account in sloping ground where shear strength plays the do1ninant role. Consideration should also be given to the fact that the results of gcotcchnical calculations often have an integral character. For example, it would not make much sense to produce an earth pressure calculation when the soil succession is detailed into layers of 30 cm each. On the other hand, thin less pcrmcahlc inter layers will have great significance for the fl.ow of groundwater or for consolidation time predictions. Where organic matter is found in the boreholes. the borehole readings should be supplemented hy indicating the distrihution of this matter in a sufficiently long section (e.g. 1 m) and its degree of deterioration. For instance, it would be 1nisleading if a layer is described as "highly organic clayey silt layer'' when within a clayey silt layer a peat interlayer has been found. In EN 1997-1, 3.4.2, it is mandatory to explain the reasons for defective or incomplete investigation results.The report shall also specify where additonal or special investigations arc still missing.
5.2
Ground design report (EN 1997-1, 2.8)
The ground design report contains the conclusions drawn by the geotechnical expert from the investigation report as well as calculations which he has made to verify lin1it states. The extent of work connected with this should only he given in a preliminary way: like the service of a doctor or a lawyer. the real an1ount of consulting often beco1nes obvious only as the project proceeds. The consultant should he charged hy estahlishcd experience and nlutual confidence in his capacity, rather than on a co1nparison of prices. The client should understand that the cheaper such work is offered, the more general and undetailcd the report will be with a tendency to shift hidden risks to the client - often without hin1 recognizing it. In EN 1997-1. 2.8, it is recommended that the report should also contain statements on the suitabli(v of a site with respect to the proposed construction and the lei-·el ofacceptable risks.
1. 1 International agreements
15
There will, however, be little freedom for alternatives when a site is located in an area which is already densely populated. ln such situations, the guidance shou1d focus on finding a suitable type of foundation in terms of its feasibility, economy, ground water conditions, compatihilty with the ahove ground parts of the structure etc. Suitahle alternatives may therefore be apt for discussion where traffic routes are to be built. The ground design report, if not contracted otherwise, will he limited to findings in terms of geotechnical verifications and duration against adverse environmental effects. This marks the boundary between geotechnical and structural engineering works. The "external" sizing of foundation elements like footings or retaining walls depends on geotechnical points of view, the sizing of concrete or steel sections and the reinforcements is truly a structural task. Even with such delimitation, the work of a geotechnical engineer contains a significant amount of quantifying engineering hy analyses. General phenomenological considerations will not yield the kind of information that is of real value to a client. Although the details of a structure may not he known at the time when the report is delivered, the ground design report should provide examples of typical calculations - especially when the design shall he hased on the partial safety factor method. For legal reasons it should be clearly indicated to the client for each number in a design report whether it is the result of an investigated fact which can be proved, if queried by anyone, or an expert assumption. The expert is obliged to state his assumptions where sufficient facts or established rules are missing. When established application rules are not used, arguments shall he provided in such a way that they can he understood and realized by a user or supervisor who may not have scientific geotechnical training but a good general knowledge of civil engineering problems. Regional terminology should not be used to avoid misunderstanding by users from outside the region. Recommendations for foundation procedures should contain all the possible alternatives to avoid a restricted competition. Methods that are acknowledged by the profession but can be implemented by only a fe\V contractors must not be recommended exclusively hut they may he considered hy allowing for a special additional tender. In EN 1997-1, 2.8(4) it is recommended that the geotechnical design report should also deal with items which require checking during construction or which require maintenance after construction from geotechnical point of view. lt would be reasonable to provide concise "directions of use" guidance note to the as-built documents, delivered to the user when the work is finished, giving for example the recommended settlement checks to be carried out at later time intervals. As the design report follows the completed site investigation, it may hecome apparent during negotiations with contractors- especially when special tenders are evaluated-that the choice of a particular method of construction would need additional soil investigations. In such situations the consultant that compiled the design report should be asked for his advice and possibly to provide to necessary supplements to his report. However, it often happens (maybe even by intention, maybe just forgotten) that these additional soil investigations do not become an item in the tender contract. If at a later stage an unacceptable construction yields a legal case, it will be important that the geotechnical consultant can prove that he had pointed to the need for additional investigations and when. On the other hand, excessive additonal investigations should not be asked for as an alibi to avoid straightforward decisions. They should always remain in reasonable relation to the value of the project. In this way design reports will always have a degree
16
lJlrich Smoltczyk and Christophe Bauduin
of compromise: documenting the art of realizing the works in relation to ground risk avoidance.
Large geotechnical design reports should have a summary containing the most important technical statements in a concise manner \Vith examples referring to the annexes enclosed in the documents published for tenders.
6
References
fl] Peck, R.B.: Advantages and limitations of the observational method in applied soil n1echanics. 9th Rankine Lecture: Geotechnique 9 (1969), 171-187. 12] Stnoltczyk, [l: Beobachten-aber methodischrichtig. Vortri:ige Christian Veder-Symposium, Graz
1999.1-11. [3] Sn1oltczyk, U:: Internationale Vereinbarungen. In: Grundbau-Taschenbuch Teil 1, 5th edition. Ernst & Sohn, Berlin 1996, 1-23.
1.2 Determination of characteristic values C'hristophe Bauduin
1
Introduction
One of the most difficult tasks in geotechnical engineering is the selection of the "charac-
teristic value" of a geotechnical property, this is the value of the property that governs the behaviour of the geotechnical structure at the limit state considered. The difficulty results from the variability of a property in the soil, from differences between test conditions and real conditions and from the small number of tests usually available. Therefore, all relevant information complementary to the test results should be considered. This chapter illustrates the various aspects leading from test results to characteristic and design values of material properties and gives some statistical methods to assess characteristic values from a sample of test results. A clear insight can only be obtained by dividing the process into sequential, well defined steps and by kno\.ving exactly what has been done, what information is requested, and \Vhat output is obtained by each individual step. The first part of the chapter therefore identifies these steps and indicates the type of uncertainty covered by the calculation procedure of the step being considered. When soil or rock tests are evaluated, the limit state on \Vhich the test results shall be applied should be taken into account. Once all the tests are analysed. a set of (derived) values of the ground parameter are obtained giving stochastic variations from \Vhich the characteristic value for the limit state considered has to be assessed (Section 2.2). Statistical methods may be useful to do this. In Section 2.3 a fto\v chart is given, as a schedule of factors affecting the choice of the characteristic value and a description of application rules in practice. Finally, in Section 3, the different approaches are exemplified for some typical soils.
2 2.1
From derived value to design value Sequential steps
The design value of a soil parameter is obtained through a sequence of steps, starting with some site measurements and a subdivision of the soil succession into "homogeneous" layers. This subdivision is based on previous soil investigations and if available, engineering geology judgement. Homogenity means that the parameters of a soil layer show only stochastic variance, i.e. variations that are at random around the true value or follow a well established trend such as a linear increase \Vith depth.
18
Christophe Bauduin
It should he noted that homogenity need not he an ahsolute category: a soil layer may he
homogeneous with respect to a particular property (e.g. shear strength expressed through angle q/) hut not to another one. Homogenity is also related to scaling; considering a soil mass at a small scale may lead to the conclusion that it is not homogeneous but it may be assumed to he homogeneous when considering a soil mass as a whole.
The sequence from test result to design value comprises the following main steps. In some situations they may he performed simultaneously or in a different order: Step 1: Measuring of the !J'oil behaviour on !J'ite or in the laboratory An example of this step is carrying out a CPT or a SP1~ where the resistance opposed by the soil to the (static or dynamic) penetrntion of a measuring device is measured as a function of depth. Another example is the use of an oedometer test where the axial strain is registered as a function of the applied stress (and also sometimes as a function of time). Step 2: From mea.Jurement to mea!iured !J'oil property Examples of this are the assessment of the undrained shear strength of the soil, from the measured cone resistance or blow count, or the assessment of the compression modulus and consolidation coefficient from the oedometer test results. Step 3: From nzeasured soil property to in situ property In this step, the measured soil property is adjusted/corrected to fit to the real conditions in the soil for the problem under consideration. This step therefore needs a good insight in the expected behaviour of the soil and the geotechnical structure. Examples are the adjustment of the undrained shear strength, obtained in previous steps, to a very different stress level in the real problem, or the estnblishment of the value of the compression modulus for the expected stress increase. Step 4: From in !J'itu property to characteri!J'tic (representative) value In this step, all individual values of the in situ property are brought together to select a representative value for the soil layer. This step includes some conservatism. This requires that not only the values obtained from devoted tests be considered, but also all other information which might increase the knowledge of the soil should he included. For example, water content and grain size distribution may also contain some information to he considered when selecting the characteristic value of the undrained shear strength. Previous knowledge is extremely valuahle and should he included wherever possible. Also, previous knowledge may be test results in the same soil, or the observed behaviour of existing structures. Another important point to consider at this stage is the variability of the soil parameter, so that not only the expected (mean) value of the parameter is estahlished, hut also the magnitude of its variation. Of course the larger the number of tests and other relevant information, the higher the confidence one can have of the representativity of the value chosen and the less conservative the choice needs to he.
1.2 Determination of characteristic values
19
Step 5: /<'rom characteristic value to a design value for use in a calculation model, applying a partial safety factor Ym : Xd = Xk/Ym
When considering the above described process of determination of geotechnical parameters, one realizes that it involves a lot of uncertainties: • Step 1 contains rather few uncertainties, especially as testing material and procedures are more and more standardized. Any measurement corrections should be performed at that step; • Step 2 contains more uncertainty, as some interpretation is involved. For example, triaxial tests, even when well performed, may include some difficulties in trying to establish the values of c1 and q/ by drawing a straight line tangent to the failure circles. 'i\'hen deriving a parameter value through an empirical relationship (c. g. undrained shear strength from cone resistance) the uncertainty and validity of the relationship used arises. So this step contains more uncertainty than is usually admitted; • 1he uncertainty related to step 3 is often wrongly neglected. For example, the soil is much stiffer at small strains than at large strains; the stiffness of the soil is often strongly influenced by the stress path. 1he adjustment from the measured soil property to the in situ property requires good soil mechanics judgement • Step 4 contains often the largest contibution to the uncertainty. A difficulty in this step is that the engineer has either very little information, or has to deal with a lot of diverse, sometimes contradicting information. Statistical techniques are helpful to complement sound judgement in this crucial step. • The partial factor, introduced in step 5, aimed to cover unfavourable deviations of the real value of the soil property from the characteristic value, also includes some uncertainty Table 1 summarizes the aspects of proceeding.
2.2
Points of view when analyzing test results
Properties of soil and rock and rock masses arc quantiiicd by geotechnical parameters which are used in design calculations. EN 1997-1 Section 3 requires ground properties to be derived from the results of field and laboratory tests and other relevant data. The test results should be presented in such a way that the determination of gcotcchnical parameter values is understandable. The values of geotechnical parameters are either directly obtained from test results or as derived values based on tests. ''"Derived values" are defined in Eurocode 7 part 3 as: "a value of a geotechnical parameter obtained by theory, correlation or empiricalism from test results". Examples of soil parameters that may be obtained directly, are the results of triaxial tests (shear strength: peak value, at critical state, at a given strain; E-modulus at a given stress level, e.g. Eso) and oedometer tests (compression modulus at a given stress interval; consolidation coefficient at a given stress interval) etc. Derived values are mainly obtained from field tests. Examples of this arc undrained shear strength or the angle of internal friction deduced from cone resistance (CPT) or blow count (SPT). Part 3 of Eurocode 7 gives correlations for Iinding derived values for gcotcchnical parameters from most standard field tests (see Table 2 for some examples).
Validated value of 2-3 parameter or relationship between parameters
Test errors
Geutechnical Evaluation of test parameter results wlue
Each single test
- Calculation model intended to be used for analysis of the limit state considered - Assumed behaviour at the LS; range of stress or strains if relevant
Value of the geotechnical paran1eter for the LS considered
Adequacy bet•,1,1een test result and calculation model for LS considered
Each single test result, after idealisation if relevant
- Calculation model intended to be used for analysis of the limit state considered - Assumed behaviour at the LS; range of stresses or strains if relevant
Derived value out 2-3 of each test result. \\''hich can be: - geotechnical parameter value - coefficient or resistance in indirect method
Derived value
Derivation of parameter(s) as needed in the calculation model, converting and - if necessary correcting the test result into soil parameter(s) assumed to modelise the soil behaviour
Applying on test results: - Empirical relations - Semi-empirical relations - Theoretical formulas - Calibration factors
Output
Part of EC7
2-3
Asepct of uncertainty COill'e["ed
-Adequacy between test result and calculation model for LS considered - Calibration of model
Table 1 (continued) Acth,ity
Characteristic value
Method
Applied on
Requested info
Output
Part of
EC7
Asepct of uncertainty convercd
Spatial stochastic varia-
Selection of a cau- Statistical methods tious estimate of the value governing - Engineering judgement the occurrence of the limit stcitc under consideration, taking into account soil variability \Vithin the homogeneous layer, and incl. all relevcint soil parameter knowledge from previous experience
All avail- Limit state considable releered vcint deriv- - Stcitistical t:ype of data (local ed values and comregional- previous plernentary knowledge) information - Ability of soil and/ (ci priori or structure to knowledge redistribute previous experience)
The chcin1c1eristic value(s) for the limit state considered
1
Dividing value by to be used in the ver- mciterial ractor YM ificcition of the LS under consideration
Characteristic value
Design value
1
Design ''aloe Deriving the value
(2.4.3(P)) tions of the soil para-
meters ciround its value or established trend in homogeneous soil layer
Scifety mcirgin: - Unfcivourable devicitions from char. values - Inciccur acy in conversion factor - Uncertainties geometry cind calc. model
c
22
Christophe Bauduin
Table 2. Examples of gcotcchnical parameters derived from field tests -
Field test
Test results
CPT
q,
Derived value
B/R11
Reference to EC 7~3, Annex
Em
B B
B.l B.3
CPT
q,
pile resistance
R
B.4
PMT
Prm
bearing capacity
R
C.1
PMT
p
settlement
PMT
Prm
pile resistance
R
C.3
SPT
(N 1)00
In
B
B
D.2 D.4 E.l E.2
-
C.2
~
DP
Nw qi'
Io for bearing capacity calculations
DP
Nio
Eoed
B
E.3
WST
half turns/ 0.2m
Em
B B
F F
Cfv
Cfu
R
G
FVT
11 B: parameter to be used in analytical method R: resistance lF/L2 J or coefficient in semi-empirical methods
Note that derived values can he a "basic'' parameter (strength parameter c' or tp'; Cu, Cc etc.) or a soil resistance (hearing capacity). An example of this is the hearing capacity of
a shallow foundation or a pile deduced from the results of pressuremeter tests. As examples of lahoratory tests, derived values are only applied when shear strength parameters are derived from the results of classification tests (e.g. gradation, water content, Atterherg limits, density). A very important point when dealing with derived values, is the kind of correlation used. EN 1997-1, Section 2, further requires • Values obtained from test results, derived values and other data shall be interpreted appropriately for the limit state considered (2.4.3 (2)). • The characteristic value ofageotechnical parameter shall be selected as a cautious estimate of the value affecting the occurrence of the limit state (2.4.5.2(2)). In the process from measurement to characteristic values therefore the adequacy between the value of the geotechnical parameter and the limit state considered should be introduced. To obtain reliable data, the following points shall he taken into account (2.4.3 (4)): • many geotechnical parameters are not true constants but depend on stress level and the mode of deformation; • soil and rock structure (fissures, laminations, large particle!>~ etc.) that may play a different role in the test and in the geotechnical structure;
1.2 Determination of characteristic values
• • • • • • •
23
time effects; the softening effect of percolating water on soil and rock strength; the softening effect of dynamic actions; the brittleness or ductility of the soil and rock tested; the method of installation of the geotechnical structure; the influence of workmanship on artificially placed or improved ground; the effect of construction activities on the properties of the ground.
In subclause (5), the need to consider relevant published information and any kind of useful experience is also emphasized. A calibration factor shall be applied where necessary to get laboratory and field test results to represent the behaviour of the soil and rock in the ground (2.4.3 ( 6) ). An example of such a calibration factor is the factor 1.1 to convert a q:: 1 value from triaxial tests to plane strain conditions. Preferably calibration factors should be applied to each individual test result before summing them up and selecting the characteristic value. ll11L"itration of' the points mentioned above
1. Lffect of stress and strain dependency: strain compatibility As hard and soft layers differ significantly in their stress-strain performance, the shear strength of a soil succession is influenced such that the hard layers yield their peak strength at small strain, when the soft layers are still far from mobilizing their maximum strength. When ultimate limit states are to be verified therefore, it is not appropriate to apply peak strength values of the different layers involved without careful consideration. In the same way, deformation moduli of the soil should be chosen in accordance with the stress level. 2. Effect of differences in the soil structure between the test and the real structure A typical example of the different effect of a fissured soil structure on the test results and on the real structure is the size effect on CPT results in stiff fissured tertiary clays. For these cases, the bearing capacity for piles based on the CIT method (according to [ l]) has to be reduced by a factor depending on the ratio of the pile diameter to the cone diameter. Any corrections related to the specific test or test conditions should be performed at this early stage of the process. Some considerations about the use of correlation to obtain derived values
One of the difficulties often encountered in applying derived values is the lack of knowledge about the variance and safety contained in a correlation. A correlation might be • • • •
a "mean" correlation, a "conservative estimate"correlation, provide a "characteristic" value, provide a "design" value.
If a correlation is a "conservative estimate", how conservative is it? (e.g. a cautious guess of a mean value, a lower bound correlation?).
24
Christophe Bauduin
Some existing national codes provide "characteristic values" of geotechnical parame~ ters derived from the results of field tests or identification tests (see e.g. NEN 6740 or DIN 4014). The characteristic values in this case are often given for "classes" of measured values. Other national codes give "design values" to he derived from (field) tests that are often denoted as "allowable values", mostly related to a deterministic design approach (i.e. unfactored loads). Here also the design values are given for ''classes" of measured values. In these latter two cases homogeneous conditions are usually assumed without specifying some criteria to check if homogenity is fulfilled or not. The test results available to establish derived values are often results given as a function of depth for different test locations. To manage this information, some ''idealization" of the obtained profiles is usually made. However, with natural soils, variations of measured properties may indicate significant variations in site conditions. It is therefore very important that weak zones are identified. Summarizing test results from several test locations at too early stage, by using statistics, may mask the variability of the geotechnical parameter value and/or the presence of weaker zones. Therefore, both the idealization of the soil succession and the determination of derived values should be done for each test location separately. When correlations are applied to derive a shear parameter or deformation moduli from measured values, a diagram as illustrated by Fig. 1 is produced (e.g. taking the horizontal axis for measured water contents, the vertical axis for derived shear parameter values).
•
•• • ••• • • • • •• • •
• test location 1 • test location 2 a test location 3
. . ..
~--·· •
Fig. 1. Example for a cautious choice of derived values frotn scattering measured values
measured value
As it is not evident if the correlation already considers the variance of values, it seems logical to choose a lower hound curve in terms of a "cautious estimate" when characteristic input data for verifying calculations are to be developed taking account of the spatial distribution in the next step. The reason for this is the fact that the applied correlations are normally based on wide-spread regional experiences and need not necessarily reflect local conditions. If however, it is possible to base a correlation (for example between Atterberg limits and cu) on a sufficiently large number of site samples, a "mean" correlation will be appropriate.
Derived values used in semi-empirical models When using a semi-empirical model, a test result (usually from field tests) is transformed into a resistance value. The transformation rule is often the result of the comparison between the test result and the behaviour of the geotechnical structure that has been
1.2 Determination of characteristic values
25
nleasured experin1entally: e.g. pile shaft unit skin friction deduced by cone resistance, is based on the comparison of the skin friction obtained by pile tests and CPT results. A calibration factor may be introduced to adapt the value of the test result or derived value to the limit state considered. Such factors also aim to increase the reliability of the prediction by covering its scatter. Examples are: • The correction factor on the field test result to obtain the value of undrained shear strength for slope stability design (see EN 1997-3, Annex G). The factors given are based on local experience and back calculations of slope failures. • The factor ~ to derive the ultimate compressive pile resistance from semi-empirical methods (in situ tests, EN 1997-1, 7.6.2.3(6)P and Annex A). If for example CPTs are applied, the factor should be such as to have a probability of less than 5 o/o from a deviating individual result. Remarks:
a) The value of the calibration factor depends on the level of reliability chosen. This in turn depends on the total safety required for the type of geotechnical structure considered and the load and material factors applied. b) The calibration factor is linked with the calculation method used.
2.3
Points of view when determining characteristic values of ground parameters (EN 1997-1, 2.4.5)
Once a set of values for the geotechnical parameters has been determined, the characteristic value for the problen1 being analyzed has to be evaluated from it. The selection of the characteristic values of soil and rock properties shall take account of the following (EN 1997-1, 2.4.5.2(4)P): • geological and other background information, such as data from previous projects; • the variability of the measured properly values and other relevant information, e.g. from existing knolvledge; • the extent of the field and laboratory investi{iation; • the type and number of samples; • the extent of the zone of ground governing the behaviour of the geotechnical structure al the limit stale being considered; • the ability of the geotechnical structure to transfer loads fro1n }veak lo ,j·trong zones in the ground. Complementary sources of relevant information may include databases, known variation coefficients of the geotechnical parameter considered and test results from nearby locations. 1. Effect of soil volume and/or ability of the strut.1ure to transfer loads
The local (point) values of a geotechnical paran1eter fluctuate in the soil around its mean value (or trend). It is always necessary to identify the specific volume of ground which governs the behaviour of the structure in the limit state considered.
26
Christophe Bauduin
As the characteristic value is required to be a cautious estimate of the value aff"ecting the occurrence of the limit state (EN 1997-1, 2.4.5.2(2)P), the first question to be answered is: what magnitude of the soil parameter governs the limit state in relation with the extent of the soil influence zone? A value close to the mean value may govern the limit state when (see 2.4.5.2(7) and (9))
• a large soil volume within the homogeneous layers is involved, allowing for co1npensation of (stochastically occurring) weaker hy stronger areas; • the structure carried by the soil allows a transfer of forces from weaker to stronger foundation points_ A value close to the (stochastically occurring) lowest value of the soil para1neter may govern the limit slate when
• a small soil volume is involved that does not allow for compensation: a kinematically possible failure surface may develop mainly within the weak soil volume; • the structure may reach its ultimate limit state before the transfer of forces from weak to strong parts can occur. The extent of the influence zone depends on several factors: • The type and size of the structure have the most significant effect on the influenced soiL As an example, in the case of a raft resting on a sand soil containing some loam lenses, the behaviour is governed by the sand, while in the case of a foundation on individual pads, the behaviour can he governed by a weak spot occurring under a single pad. • The stiffness of the structure: a stiff structure will allow load transfer from weak to stronger zones, allowing it to reduce the etnphasis on the local weak spots. As an example, in the case of a stiff structure founded by strip footings on a sand soil containing so1ne loatn lenses, the behaviour is governed by the sand, while in the case of a weak structure foundatlon on individual pads, the behaviour can be governed by a weak spot occurring under a single pad. • The loads involved: for example, the shape of the failure surface under a strip footing depends on the ratio between vertical and horizontal components of the load. It should be etnphasized here that "large" and "small" are related to the distance over which the stochastlc fluctuation of the property occurs around its mean value (autocorrelationlength): the smaller this distance (i.e. quick variations around themean value), the easier a "mean" value can be obtained. For a long auto-correlation length (i.e. slow variations of the soil property around the 1nean value) a significant part of the influence zone can be located within the part of the soil having the low point values. Remark: The influeneed volume of soil should not be confused with the volume of soil which needs to be investigated by borings and in situ tests, as part of the soil investigation, prior to the design. 2. Processes governed by extreme values of soil properties When carrying out the design analysis, it should be noted that processes in the soil 1nay be governed by "extreme'' values of the soil parameters, despite the fact that a large soil volume is involved or that the stiffness of the structure allows transfer of loads.
1.2 Determination of characteristic values
27
Examples of this are deposits forn1ed hy thin, parallel layers (sand\vich formation) that are often modelled as homogeneous layers, although for special calculations they behave in a non-homogeneous \Vay. Examples are: • groundwater-fto\v perpendicular to layered deposit, governed by lowest vertical permeability; • groundwater-fto\v parallel to layered deposit, governed by highest horizontal permeability; • compression of a layered deposit, governed by compressihility of \Veakest component; • reactivation of slip movements along previously developed slip surfaces; • a kinen1atically adn1issible slip surface through a "chain" of weak ground section. When it is recognized that a process for a limit state is governed by the extreme values of a soil property within the influence zone of soil, the characteristic value should he selected as a cautious estimate of the governing extreme value, and not as a mean or fractile value of the property. For instance: in a layered deposit, the mean value of the compressibility of the most compressible strata should be considered rather than the point value of the tests including compressihle and non-compressihle strata. When soil or rock nlasses sho\v a brittle (or strongly expressed softening ) behaviour, one should he extremely careful when considering "mean values" or "redistrihution from strong to weak areas"; brittle materials do sho\\' a brutal rupture, together with a loss of all strength compared to ductile materials, which more or less nlaintain their strength when strained continuously after reaching their full strength. Failure systems in ductile soils are nlore or less parallel systen1s, while they are close to series systen1s in hrittle soils. This is illustrated in Fig. 2 hy a stiff structure supported hy four foundation points, having ultimate bearing capacities of 125 kN and 200 kN. In the exan1ple, "brittle" corre.'iponds to a complete loss of strength after the peak value is reached, while ductile means that the maxin1un1 resistance is retained even after the nlaximun1 value is attained.
125
200
200
125
Fig. 2. Ex::implc of footing support
For ductile behaviour the ultimate hearing capacity of the foundation is 650kN (125 + 125 + 200 + 200) and is related to the mean value of the individual foundation points. For hrittle hehaviour, the ultimate hearing capacity is reached once the first foundation element reaches its nlaxin1al bearing capacity and is 500 (there is no redistribution possible as once 125 is reached at supports 1and4, further loading means loss of strength in these points as the two other supports are loaded to more than 250 kN, what they cannot carry). The strength in the latter case is determined by the lowest value.
3. Type of sampling, extent of investigation and complementary information Distinction should he made hetween "local" and "regional" sampling, taking account of the fact that the same formation may deviate in its performance. Local sampling means that the test are taken at (or close to) the site of the intended geotcchnical structure. In the case of regional san1pling, the results of tests taken over a wide area are hrought together (may he in a databank). Examples of this are road, railway or dyke projects. In these cases spatial variation has to be considered: the same soil layer nlay have rather
28
C:hristophe Bauduin
different property values at different locations along the project. Special account has to
be taken for this source of uncertainty. Clearly the more information available, the more reliable the prediction of the character-
istic value will be. The Eurocode insists on adding all the complementary information relevant to the test results of the local soil investigation because \\ ith local sampling only a few tests arc 1
normally available. The complementary information may be introduced through: • A good kno\\ lcdgc of the coefficient of variation of the property. Within a soil layer, the coefficient of variation docs not vary much, especially if compared \\ 1th the mean 1
1
value. \Vhen the coefficient is known, this notably reduces the uncertainty of assessing the characteristic value. The statistical formula used \\ ill therefore be different when both the mean value and the coefficient of variation have to be determined from a set of results, compared with one used when the coefficient of variation is already known and only the mean value has to be determined. 1
• Bayesian statistics, in which a prior kno\\ ledgeof values of the soil property (established by regional databases or previous comparable projects) is used as a starting point and in \\ hich the re-suits of samples from a given location are used to verify it, if the prior values are used or need to be updated. This method is especially useful \\'hen a good prior knowledge is available and it is not intended to perform many tests at a given construction site. However, if only local investigation results arc available and any kind of pre-information is lacking, then very conservative characteristic values should be assumed due to the small number of local tests. 1
1
2.4
Use of statistical methods
EN 1997-1 allows the use of statistical methods to select the characteristic value. \Vhen they arc used, in 2.4.5.2(11) it requires:
If statistical niethods are used, the characteristic value should be derived such that the calculated probability of a worse value governing the occurrence of the limit state under consideration is not greater than 5 o/o.
In this respect, a cautious estimate of the mean value is the selection of the mean value of the limited set of geotechnical parameter values, \\ ith a confidence level of 95 °/o; a cautious estimate of the low value is a 5 °/o fractile. The difference between these cases is illustrated in Fig. 3. The flo\\' charts in Figs. 4 and 5 sum up all the previously mentioned essentials leading to a mathematical formulation for the assessment of characteristic values. These essentials are summarized as follows: 1
a) b) c) d)
local or regional sampling; homogeneous soil or trend; large or small soil volume involved-load redistribution by structural stiffness; useful experience available, prior knowledge.
The flow charts do not give all of the possible combinations between items (a) to (d), but only those that are most relevant for daily practice. Bayesian analysis is not treated further here.
29
1.2 Determination of characteristic values mean value
n
:•
•,
• • s: •
,,'
/I
,
/
t
: s. '--,,,' ,, /I
•
~
I :
XK, m~an 1Xmean
Xk 5% fractile /
possible distribution (estimated for N tests)
•
.t1g. 3. Example of 95 o/o confidence required for a mean and of a 5 % fractile
yes mean value at95% confidence level, linear variation
no
mean value at 95% confidence level, Vknown
Fi.g. 4. Local sampling as starting point
no
yes mean value at95% confidence level, Vunknown
point value 5% fra.ctile, linear variety
point value 5% fractile, Vknown
point value 5% fractHe, Vunknown
30
C:hristophe Bauduin
geotechnica/ parameters or derived values regional experience; no significant trenrl
supplemented local
---~n.o
__ _
tests
yes
mean value for 95% confidence level
local vertical variance
go on as in fig. 4 from "no significant trend"
to be considered, supplemented by known regional
variance
Fig. 5, Regional sampling as starting poinl
Note: Statistical tnethods are inappropriate when very few test results are availaible, except if they can be used in conjunction with previous experience and/or Baysian analy-
sis. Very often the "previous experience" is su1n1narized in standard tables, giving characteristic values of soil properties, cautiously estimated on basis of region.al experience (see Section 2.2). 2.4.1
Statistical formu]as to assess the characteristic 1-'a]ue
A norn1al distribution is assumed. This assu1nption is discussed in 1nore detail in Section 2.4.2. Case 1 No significant trend; no experience available. ln this first case the characteristic value will be an estimate of the real mean value from n tests, with a confidence level of 95 o/o (50 % fractile with 95 % confidence level). ln other words the characteristic value with will be estimated such that there is a probability of 95 o/o that the "real" mean value is more reliable than the estimated one:
xk
= -x. [ 1
-
0.95 . t(n-1)
1 J v. Jn
where
X
- arithtnetical mean of n values Xi of the geotechnical parameter V = s/X - coefficient of variation 2
1
~
-2
s = - - · L)X; - X) n - 1 i=l t0.95 (n-1)
- Student factor with (n - 1) degrees of freedom at 95 °/o confidence
(1)
31
1.2 Determination of charactt:ristic values
Table 3 gives the values of ~I = Xk/X as a function of n and V. Table 3. Numt:rical values of ~ 1 V=
0.05
0.10
0.15
0.20
0.25
n:::3
0.92 0.94 0.95 0.% 0.97 0.97 0.98 0.99
0.83 0.88 0.90 0.92 0.93 0.94 0.96 0.97
0.75 0.82 0.86 0.88 0.90 0.91 0.94 0.96
0.66 0.76 0.81 0.84 0.87 0.89 0.92 0.9.1
0.58 0.71 0.76 0.79 0.84 0.86 0.90 0.93
4 5 6 8 10 20 40
Case2
As Case 1, but calculating a lo\\' value where there is only a 5 o/o chance of finding a location in the soil having a \\'orse value:
xk
09' = -x. [ 1- tn~n)°
v
p1)] -x
v\1 + ~)
=
~2
(2)
It should be noted that there is a small difference between the statistical aspects of the formulas for the mean value and for the fractile value. The mean value (50 °/ci fractile) at 95 {Yo coniidence level considers the stochastic variation and the statistical uncertainty as t\\'O different sources of uncertainty that should be covered by the statistical formula. The 5 °/o fractile considers both uncertainties together. The mean value at 95 % has a more "predictive aspect'' while the fractile has an "observation from test result set" aspect. A "95 °/o reliable 5 °/o fractile" value leads to extremely low 1; values. Conversely, the~ values given for the 5 o/t> fractile in Table 4 are close to a 75 °/o reliability guess of a 5 °/c> fractile. Table 4. Numerical values of 1;2 -
--
-
V=
0.05
0.10
0.15
0.20
0.25
n=3 4 5 6
0.83 0.87 0.88 0.89 0.90 0.90 0.91 0.92
0.66 0.74 0.77 0.78 0.80 0.81 0.82 0.83
0.49 0.61 0.65 0.67 0.70 0.71 0.73 0.75
0.33 0.47 0.53 0 ..16 0.61 0.62 0.65 0.66
0.16 0.34 0.42 0.46 0.51 0.52 0.56 0.58
8 10 20 40
Casc3 Useful well~known experience available. \Vhen enough test results (including on site and off site information) are available to state that the variation coefficient V of the particular soil parameter is well-known and
32
Christophe Bauduin
that locally V does not much differ from that known value, advantage may be taken of well estahlished experience in terms of V hy replacing t~:~l) by 1.645 in the previous formulae. The decision to consider Vas heing well-known depends on the judgement of the responsible engineer. Such a statement normally requires a supporting database of acknowledged test results.
Case4 Experience available hut V not fully known. This case might he approached hy using Bayesian analysis ((3. 6, 7]). However, its usefulness for standard problems seems to be questionable, especially when comparing the two limit situations (V unknown - V known) mentioned ahove.
Remark: The following cases refer to situations where the value of a soil property increases (linearly) with depth or the shear resistance-t increases with the mean principal stress s.
Cases No useful experience available, characteristic mean value at a 95 o/o confidence for a linear trend. Statistical methods are availahle to calculate the confidence interval with a given prohability and the prediction interval of a linear regression curve. The lower bound of the confidence interval, when calculated with a confidence level of 95 %, gives the characteristic value of the real mean.
The best estimate, X(z), of the ground parameter at a depth z, follows the Student's t-distribution with (n - 2) degrees of freedom. The mean value equals to the true mean of the ground parameter at this depth, and the standard deviation, s1, as follows:
Thus, the characteristic value of X at depth z is:
x. "" x + b (z - z) -
t~~9:\ 1 · s1
x = x+ b(z - z) where
(4) (5)
1
x=-(x1+x2+ 11
I
+ z2 + ... +Zn)
Z = -(z1 n n
2..:(x, - X)(z; - z) h=
i_~_l_ _ _ _ __
n
2..:(z,-zJ 2 i=l
(6)
33
1.2 Detennination of characteristic values
It should be noted that the calculated characteristic values arc not a linear but a slightly hyperbolic function of depth due to the term tZ~~2 ) · s l · 'fhe distance between the linear regression and the characteristic value is s1nallest at the centre of gravity of the readings. "This shows the advantage of performing the tests hoth within the relevant problen1 interval and slightly outside. As 1nany calculation nlethods and co1nputer codes use linear relationships, the hyperbolic relation in the relevant (stress) interval has to be linearized, hut subsequent errors are usually insignificant. Remark: The significance of the presumed trend can easily be proven hy a statistical test.
Case 6 As Case 5 hut for point value 5
s~ =
[ [ ---=2 n
1 + -1 n
+ ,,(z -
''( L., Zj
Z)2 -) - Z 2
l
n [
x) - h(z; - z)
2
J
(7)
._, L-
i=!
Thus the local characteristic value
xk
=
-
x + b.
-
xk at depth z is
0.95
(z - z) - t(n-2) . s2
(8)
Case7 As Cases 5 and 6, but with pre-information. Available experience may be used. \\'hen very complicated statistical methods arc to he avoided, pragmatic approaches can be followed as hclow: a) to qualitatively check the assumption of an existing significant trend; b) to verify the b-value (semi-quantitative evaluation); c) to add con1ple1nentary 1neasure1nents to the local set. (a) and (b) evaluations should always be made. Usually, when a significant trend is established by local test results and checked according to (a) and (b ), there is not much to gain by (c). This last one should always be done with care as the value of (b) may have some regional consistency but might vary on site at z = 0 due to different values of overhurden.
34
Christophe Bauduin
Cases Regional soil sampling with no regional or local trend versus depth and no local soil test available. Mean value at a 95 o/o confidence level.
This case has to be calculated taking care to consider the variability at one point (along vertical axis) and the glohal variahility. A formulation according to [2] n1ay be: (9)
with the symbols are as before plus
Sf
- standard deviation at a specific test location (variations around the local n1can
Sxm -
value) assu1ned to be the same at all test locations, standard deviation of the local mean values around the overall mean value.
Table 5 gives nu1nerical values 1;3(u) = Xk/X for different values of V and n. Values given for small test nun1bcrs (c. g. n :S 8) should be disregarded: a regional data set based on so few results is very questionable. Remarks: • a = 0 means dominant stochastic variations over the area under consideration. The characteristic value is then the 5 o/o fractile. • u = l 1neans no stochastic variations over the area, only local variations. The characteristic value is the nican value at a 95 % confidence level. • A niajor difficulty is the choice of an appropriate u value. Dutch experience based on a small nun1hcr of nicasurcn1cnts indicates values hctwccn 05. and 0.7 rzJ. • The forn1ula was established for embank1nent stability analyses. It may be used where large soil volumes (e.g. embankments, long retainments etc) arc involved in the lin1it state under consideration. Where small volumes are involved (e.g. strip footings), it probably provides too optimistic values: a 5 °/o fractile from the regional san1plc would then prohahly he a hcttcr approach.
Case9 Local tests combined with regional sampling (sec Cases 3 and 4). Calculate the local nican value and the characteristic value as in Case 3, with V produced according to regional experience. Care should be taken if the local mean value differs significantly from the regional nican value, either much lower or 1nuch higher. Additional tests would then be required. Prior to using this nicthod it should he confirn1cd that there is no underlying regional trend (e.g. linear variation of the para1neter value in horizontal direction). This can only he established from large data hascs. Statistical analysis of such regional datahascs is beyond the scope of this chapter.
35
1.2 Determination of characteristic values Table 5. Numerical values of S3 V=
0.05
0.10
0.15
0.20
0.25
0.96 0.97
0.93
0.94
0.90 0.91
0.87 0.89
0.84 0.86
0.96 0.%
0.91 0.92
0.87 0.88
0.82 0.84
0.78 0.80
0.95 0.95
0.89 0.90
0.84 0.85
0.79 0.80
0.73 0.75
0.94 0.94
0.88 0.89
0.82 0.83
0.76 0.77
0.70 0.71
0.93 0.94
0.87 0.87
0.80 0.81
0.73 0.74
0.66 0.68
0.93 0.93
0.85 0.86
0.78 0.79
0.71 0.72
0.63 0.65
0.92 0.92
0.84 0.85
0.76 0.77
0.68 0.70
0.60 0.62
0.92 0.92
0.83 0.84
0.75 0.76
0.66 0.68
0.58 0.59
0.91 0.91
0.82 0.83
0.73 0.74
0.64 0.66
0.55 0.57
0.91 0.91
0.81 0.82
0.72 0.73
0.62 0.64
0.53 0.55
0.90 0.90
0.80 0.81
0.70 0.71
0.61 0.62
0.51 0.52
u = 1.0 n/t(n-ll = 8/l.860 1011.812 u = 0.9
811.860 1011.812 a= 0.8
811.860 1011.812 a= 0.7
811.860 1011.812 a= 0.6
811.860 1()11.812 a= 0.5
811.860 1011.812 u = 0.4
811.860 1011.812 a= 03
811.860 1011.812 a= 0.2
811.860 1011.812 u = 0.1
811.860 1011.812 a=O
811.860 1011.812
2.4.2 Choice of a distribution The formulae in Section 2.4.1 have been established assuming a normal distribution of the geotechnical parameters. Is this assumption plausible? A definite answer is not possible but some general thoughts might help rellex.ion on this matter.
36
Christophe Bauduin
• The log-normal distrihution (i.e. the logarithm of the parameter has a normal distrihution) may be used by transforming each value as follows: x;=logXi
x' = S2 =
1 nL....
1 nL....
- · °"X'1 = - · °"(logX;)
(10)
L(X -X'l2 I
n - 1
• As geotechnical parameter values are always positive, they are in fact not normaldistrihuted.
• Enough tests are very seldom available to make a definite choice of the best distribution. • For large values of V, the log-normal distrihution should be adopted_ • For small values of V, the difference hetween the results of both distributions is small. Note that for the processes governed hy extreme values, the data assemhly should be reduced to the extreme values before making an assumption about the distribution of the relevant extreme values What should he done \.Vhen only very few tests are availahle? lt is a common situation in geotechnical engineering that only 1 or 2 tests are made. It is impossihle to apply statistical methods to such small samples. Engineering judgement and complementary information become then of enormous importance. As a first step, one should try to increase the amount of local information hy using values obtained as derived values from other tests performed at the site. If there a large datahase is availahle, correlations can he cross-checked and a regional approach or Bayesian techniques applied. "Regional" information is often recorded in standard tables, indicating characteristic values (or ranges of values) of the usual soil properties as a function of some sample in-situ or laboratory measurements (qc, sounding tests, classification tests etc). Comparison of the fe\.v test results \.Vith this complementary information should lead to the selection of the characteristic value. It is, however, impossible to establish definite rules indicating the relative importance of either source of information, and the few test results. Taking the most conservative value for both will probably yield a safe design, although in many cases not the most economic one.
2.4.3
Characteristic values in the light of geotechnical categories
For geotechnical Category 1 prohlems there are usually only a fe\.Y field tests (soundings) or a bore-log with some classification tests and maybe some general geological information. Characteristic values may then he taken from standardized tahles hased on regional experience, with the results of the field or classification tests used as input. The values in such tahles are of course conservative estimates. The problem of choosing characteristic values is most complicated for geotechnical Category 2 problem~ at least theoretically. Indeed, more sophisticated calculation methods
37
1.2 Determination of characteristic values
are used requiring "accurate" input values, whilst usually few (or no) appropriate test results are available. It is clear that the standard charts mentioned for geotechnical Category 1 provide a first estimate. However, this usually leads to a conservative design. The use of a more detailed analysis hy comparing the few test results v..·ith those of (regional) databases allows for a better estimate of the characteristic value. Normally the extreme assumptions concerning the knowledge ofV (knov..·n or totally unknown, see Section 2.4.l) are sufficient to estimate parameters for calculations on Category 2 level prohlems, although sometimes Bayesian analysis should he used. The specialism however,
related to this kind of analysis usually prohibits the application to geotechnical Category 2 problems. For prohlems in geotechnical C~ategory 3, the extent of field and laboratory testing is usually such that local knowledge becomes more important than the regional information. C~omparison of hoth, however, is alv..·ays advisahle.
2.4.4
Conclusions: advantages and pitfalls of statistical methods in geotechnical engineering
The use of statistical methods v..·ithout sound judgement might lead to completely erroneous results despite the appearance of some accuracy due to the use of mathematical formulae. That is why statistics should never he applied without good understanding of the problem; statistics are not a method that yields automatically a correct result. They are only an ohjective tool to support sound reasoning. If the reasoning is false. then the statistical methods will not eliminate the faults and the results will remain false. On the other hand an engineer who is trained to justify the selection of a characteristic value based on all available information, will have no difficulty in choosing the most adequate statistical approach.
3
Examples
3.1
Local sampling
Assume the data set illustrated in Fig. 6 and calculate the characteristic mean value at a confidence level of 95 °/o. shear angle
30° 0
T
'
•
6
:g:
8
.§-
10
s
I
i '
•
•
2
' 4 '----'-
q:i
40°
35•
• -
-~
•
•
Fig. 6. Shear :::ingles from 10m deep investigation (example)
38
Chrislophe Bauduin
(a) All tests considered The analysis is performed on tan qi: tan q; = 0.6914; Oji= 34.6° s(tanqi) = 0.060; (s(
(b) Less tests considered To illustrate the effect of the number of tests, only some of the test results of Fig. 6 are considered: Tests
tan qi"
'P
1 to 5 1,2.].7 1.2.3 1,2,7,8
0.699 0.735 0.720 0.707
36.3' 35.T' 35.2'
35.0°
s 0.0677 0.0770 0.0867 0.0943
3.9' 4.4' 4.9.. 5.4'
tan <[lk
~k
0.635 0.644 0.574 0.596
32.4° 32.8°
29.9" 33.4°
With only a fe\v tests available, the characteristic mean value is rather low when the formula from Case 1 in Section 2.4.1 is used as a method of evaluation.
The reasons \vhy and a proposal to get much closer to the reality is given in Section 3.4 "Analysis of strength tests". There are also suggestions [9] that a simplified rule Xk ~ X ·(I - 0.5 V) might be applied which does not consider explicitly the number of tests. In this example this formula \vould yield values of (Pk between 33.4" and 34.0".
3.2
Local sampling with V well-known
The same test results in Fig. 6 are taken but Vis assumed to be known with a value of 0.087. This is the value of the sample. Of course, normally many more tests would be required than are shown here but for this example it allcJ\vs easy comparison of results. (a) T\vo tests are performed. From regional sampling tanip = 0.691 (CV = 34.6')) and V = 0.087 are known. Applying the equation t"nqik = tanq>[l - 1.645 · 0.087(1/,/2)] the calculation of the local characteristic value will now· be based on the various sets of measured values: 2 values of q:
tan qi"
tan
32° 38° 32° 32°
0.625 0.788 0.703 0.650
0.562 0.709 0.632 0.584
and 32" and 38.5' and 38° and 34° -
'fk
~k
29.3' 35.3' 32.3' 30.3' ' -
39
1.2 Determination of characteristic values
It is now possible to check that no errors have been made by introducing the knowledge of V to a set of only two measurements: • the characteristic value is about 10 °/o lower than the mean value; • except in the second example (38°; 38.5°) the calculated characteristic value is lower than the regional mean value of the layer (34.fr"). As both tests are close together and significantly higher than the mean value of the soil layer, the calculated characteristic value would be acceptable if other tests on the site (e.g. CPT, DPT,SPT) indicate that the soil is denser locally.
(b) Comparison of the benefit of knowing V by using the results of ex.ample (b) in Section 3.1: V known
It can be seen that the cpk values for a known V arc 5 o/o to 10 °/o higher than those for an unknown V. 111e difference increases with decreasing numbers of tests. The simplified formula by Schneider [9] mentioned in Section 3.1 yields a value close to the value of
3.3 Soil property increasing linearily with depth The next example illustrates Case 5, leading to the "characteristic mean value" and to the "point value" as a function of depth. lbe available data include n = 20 vane test results of Cu which clearly show (sec J<,ig. 7) a linear trend versus depth. 'lbe characteristic mean value and point value are calculated by applying the Case 5 equations and arc plotted in Fig. 7 together with the measured values.
As the fiuctuations of the measured values are low, the characteristic mean value is close to the regression line.
5 0 2
4
15
10
----r· ...• '",,. . . .
p-..~-i----
charact.--' mean values-+-~""-'"-·•- · regression line ..-;;+----"---l
6f---t-
. ',
s
'
charact.~·'"""+~.,,
c----t-fractile values :[ B
25
20
j
• ,
''1
.-t----;
-----f.:!,.~. -
'
',
•
~ 10 f---~----------~·~·~'~·,,,,
Fig. 7. Cu profile in a normally-consolidated clay
Another example, Fig. 8 shows measured values of the modulus of elasticity versus depth. The linear regression line, an engineering judgement of a cautious estimate of the mean value and a 95 °/o reliable mean value, as function of depth, are shown.
20
30
modulus of elasticity [MPa] 40 50 60 70 engineering judgement
5f--~f---"'1"""-+-~~~-t-~-t-~-<
charade[ mean values •
- - , regression line
10r--~+-~-+-~~-'c-~~,.--j~~+----j
•
"€' s ~
~
15
• • • 20f---~~~~~~~~~"'-'"'-'~~~
Fig. 8. Modulu~ ot elasticity depth profile (example)
The following applies in principle to all tests used to measure the shear resistance of a soil as a function of the effective normal stress_ For the sake of simplicity triaxial tests have been used to illustrate the analysis. 3.4.1
Cautious estimate of the mean value of shear resistance
Using the equations in Section 2.4.1, Case 1, often leads to characteristic values that are lower than the purely arithmetical mean values. This is especially true when only a few tests are available, which normally happens, and when Vis not very small, which seldom happens_ When three to five tests arc available, the characteristic mean value is frequently as low as the lowest test result. This does not support the adaption of a ''cautious estimate of the mean value"_ 1\vo main reasons basically cause this and understanding them will lead to a better procedure to cautiously estimate the mean shear resistance: • As few tests are available, high values oft~·~~ have to be applied (e.g.: 5 tests ::::} t/ Jn = 0.953). This means that the characteristic value is less than the arithmetical mean minus about one standard deviation_
42
Christophe Dauduin
• The procedure for determining effective shear strength parameters is applied to c' and tan tp' independently. However, it is well kno\vn that they are negatively correlated (low c' values often go together with high cp' values and conversely). This favourable effect is neglected when ck- and q:k- are calculated separately. Remembering that the soil hchaviour is in fact not governed by c1 and q:/ (in lJLS) but by the ultimate shear resistance. and that the shear resistance is a linear function of normal stress (in the stress ranges of the gcotcchnical problems). all points {'tult: ou11} of the triaxial test results should be treated together. to look for a characteristic mean shear resistance as function of the effective normal stress according to Section 2.4.1. Case 5.
For triaxial tests, the easiest method is to plot all the 3n test results in coordinates p' = (o'L + l"1z); q = (n1 - 02). see Fig. 9. The criterion for determining the relevant {p'; q} ¥.rill he based on physical considerations ahout the geotechnical prohlem heing considered, e.g. peak strength. residual strength or strength at a certain strain level to allow for strain compatibility in multi-layered systems. As most of the calculation models use c' and q:', the hyperbolic relationship has to be linearized eventually for the relevant stress interval. Remark: If the ahove equations yield results in the range p' < 0. these shall be omitted. q
____,, p'
Fig. 9. Scheme of plotting 1neasured (qm) and calculated characteristic values ( Qk)
3.4.2 Estimate of the 5 % fractile The same can be said concerning the cautious estimate of the mean value holds for the assessment of the point value of the shear resistance, if the equations of Cases 2 and 3 in Section 2.4.l result in non-realistic values of c' and tp'. especially \Vhen few tests are available and Vis unknown.
3.4.3
Shear parameters that are not negatiYel,y correlated
If the analysis of the test results shows that c' and q:;' are not negatively correlated, the above- procedure should only be used with extreme care or not applied at all.
3.4.4
Numerical examples
The procedure described above has heen applied to a set of 10 triaxial tests of clay specimens out of which one test was rejected as being completely unreliable. In the first stage. all nine relevant tests are analysed, comparing the "p'; q" approach with the analysis of independent shear parameters c1 and q/ (local sampling. V unknown).
43
1.2 Determination of characteristic values
In a second stage the procedure is repeated for the more usual situation that only four tests are made. The results of the ;;p'; q" approach are then compared with those from an analysis of independent shear parameters (local sampling, V knov.·n, taken from the value obtained with 9 tests). The data of classification of the 9 clay samples were: Sample
Bl/38
B5/41
B7/43
B7/48
B9/45
Bl0/46
Bll/49
B12/52
32.8 19.1
31.2 17.4
31.2 19.3
27.8 18.7
30.4 19.4
34.9 18.7
33.6 19.1
32.7 19.2
48.5 0.8 3.2
62.6 1.0 2.2
64.8 2.0 3.7
47.2 1.9 2.2
86.4 0.8 4.2
94.3 1.7 4.7
84.3 2.7 1.7
78.0 1.7 5.2
96.9 2.3 3.2
w [o/o] \\'eight density lkNim 3 ] Ip[%] Vgl [ 0/o] Vea [ 0/oj
B12/61 -
and the results of CU tests are: Sample c' [kN/m2 ] 0 q/ l J
Bl/38
BS/41
B7/43
B7/48
B9/45
BI 0/46
Bll/49
B12/52
B12/61
20 25
25 22
30 24
15 29
10
40 15
55 16
85
65
25
16
15
From these results the arithmetic mean values and the characteristic mean values (95 o/o confidence) for a separate evaluation of c' an
Stsndard deviation
- - - {p'; q}-mean value ····-·-··--·· {c'; (/.!')-mean va/u-e~-~
400
..200
...
.---
o~~---~~---~~-~-~~-~
0
3.5
200
400
600
800
1000 p'[kNlm']
Fig. 11. Results of 2nd stage of calculation
Example: Boulder clay
The German Federal Jnshtution of Hydraulic Works has a database to collect test results obtained from regionally typical soils [8]. As an example, data from a boulder clay from Northern Germany will be evaluated that has been obtained by direct shear tests. Figure 12 shows histograms of shear parameters c' and c:p'. 3.5.1 Regional sampling; no local tests available The procedure according Case 8 in Section 2.4.1 is based on established experience collected in the database. A cautious characteristic mean value will therefore be determined to be applied to the stability analysis of an embankment.
46
Christophe Bauduin
Fig. U. Histo grams of c' and rp' measured by direct shear tests of boulder clay [10] •C
= 131 tests one obtains the arithmetic mean values tan
From n
2
= 9kPa with V = 0.95.
Standard deviations are assumed as follows: - for each vertical section op(tan cp) = 0.07(4°); or(c) = 15 kPa; - for the variation over the whole area oc(tan q;:i,ean) = 0.04; oc(c;nean)
= 9 kPa.
Therefore a( tan cp) = 0.072 /(0.04 2
+ 0.072 ) = 0.75;
n(c)
=
152 /(15 2
+ 92 )
= 0.73 =>take 0.7.
Then by applying the equations in Section 2.4.1: tanq{
= 0.627 · [ 1 - 1.65 · 0.09
c~ = 9 · [1 -
1.65 · 9 ·
·Ifn +
Jl~l +
1 - 0.7]
l - 0.7]
= 0.546 and
= 0.9kN/ m2 ;
chose n
c~ =
2
t kN/ m
These characteristic values are valid for the whole region in locations without tests. If th e confirmation of stability fails using these values, local testing should be performed, as there will be a good chance that the local value of shear resistance will b e higher.
47
1.2 DeLerminaLion of characteristic values
3.5.2
Regional sampling; 4 local tests
The local soil investigation gave the following results: Sample
Clay
1 3
17°/u 11 °/o 8°/o
4
HJ%
2
I
Sild
i
i
37°/u 17°/o 39°/o 30%
I I
Sand
Gravel
45o/u
0 5°/o 1 °,{, 0
67%
52o/u 60%
Ir
fkN/m2 ]
Weight density [kN/m3 ]
qi I']
fkN/m 2]
600 550 300 300
20.7 19.7 21.3 21.9
23 27 29 33
55 15 16 13
Co
0.25
!
0.17 0.20 0.17
c'
Figure 13 shows the results of the shear tests and a failure line determined by engineering judgement.
-7
7;250f-----~---~---~--~ -
~ -
~
I
I
1i5 150
I
!
t
200 f-----+----+-' I I
V
;2/
100
,,,V·
!
I
q;
i
=2r I
I __: I
50f--,..L..~.!---+,-----1-------1,
4·= 15kNlm2
I
I
o~~--~---~---~---~
0
f
100
Fig. Ll. Mohr-Coulomb diagram wilh results of direct shear tests of a boulder clay
200 300 400 effectiVe normal stress {kN/m2]
(a) Statistical method, local sampling, V unknown
Assuming a statistically homogeneous soil, in the sense that the shear strength has a random variahility around a "constant" mean value and that the ULS considered is governed by the mean values of the shear parameters c' and 4/, the equations from Case 1 in Section 2.4.1 give the following characteristic values: tanq/ = 0.534 (28'"); standard deviation 0.068 (3.9°); V = 0.127 c' = 24.8 kPa; standard deviation 20.2 kPa; V = 0.81 For n
=
(11)
4 is t~;;i~l) = 1.176. Thus the characteristic value of <.p becomes
tantpk =0.534-1.176 · 0.068 = 0.451; --> 'Pk= 24.3'• ck= 24.8 - 1.176 · 20.2"' 0 The characteristic values are low, especially for the value of c~. This is due to the very large standard deviation originating from one sample with very high cohesion. As there is no reason to disregard this sample and the other three samples have c' values close together hut much larger than ck· the validity of a normal distrihution for this set of test results might well be questioned. Since in this example no other information is known about
48
Christophe Bauduin
the soil, the log-normal distribution should be used to find a better fit. For the cohesion intercept this gives the following: Test
c' [kNlm2]
loge'
1 2 3 4
13 15 16 55
1.114 1.176 1.204 1.740
Mean value of log c': 1.309; standard deviation of log c': 0.29 Characteristic value of loge': 0.968""' c~ = 9.28kPa (b) Statistical method; local sampling; V known Starting \Vith the san1e four tests, the clements of the database arc no\v introduced (sec Section 2.4.1, Case 3). ln Section 3.5.l the values were
(r' = 32.1 ~·
\Vith a standard deviation of 2.86''.) and V = 0.089
c' = 9.4 kPa with a standard deviation of 9.0 kPa and Y = 0.95 Using the mean values determined in (a) the characteristic values then become
tan
= 24.8 · [1
-1.645 · (v'0:25) · 0.95]
= 5.4kN/m 2
Note that the local coefficients of variation are close to the regional ones. (c) Statistical method using the o' /•·relationship The calculation is based on n = 3 · 4 = 12 tests: Test
TJe Reer, F., Lousberg, F., lVallay,,~ M., Carpentier, R., TJe Jaeger, J., Paqua)'~J.: Bearing capacity of displacement piles in stiff fissured clays. lRSlA-lWONL, Comptes Rendus de Recherches Verslagen over Navorsingen No. 39 ( 1fJ77), Brussels. Calle, F.0.F.: Probabilistic analysis of stability of earth slopes. Proc. 11th ICSMFE, II, 1985, p. 809-812. TJenver, H., TJetlevsen, 0., Tarp-Johansen, /\~J.: Groun
1.3
Geotechnical field investigations Klaus-Jlirgen Melzer and U/f Bergdah/
l 1.1
Basics Standards
Section 3 of Eurocode EN 1997 Part 1 (see also Chapter 1.1) covers gcotcchnical investigations. Section 3.1 contains the requirement that field investigations are to be carried out according to internationally recognised standards and reco1n1nendations. Regarding the requirements Jor equipment and test procedures for laboratory and field investigations, reference is made to Parts 2 and 3 (ENV 1997-2; ENV 1997-3); these documents also demonstrate possibilities and examples for deriving geotechnical parameters from the test results. Besides investigations related to soil and rock mechanics properties, the field investigations have to include explorations relating to the engineering hydrology and hydrogeology and also consider aspects relevant to the environment. The scope of the investigations should be adjusted to the geotechnical category (sec Chapter 1.8, Section 4.4). This has to be supplemented in case unforeseen conditions are encountered. The geotechnical investigation shall provide all data necessary for determining the groundstructure-systen1 dependent characteristic geotechnical para111eters and those relevant to the planning and design of a structure or to determining construction materials. Only general require111ents were given in EN 1997-1 regarding the nlost commonly used field tests. Thus, preparing Part 3 became difficult because only a very limited number of internationally acknowledged standards for equipn1ent and test procedure exist. For this reason, ENV 1997-3 was prepared not only to describe means of deriving values of geotechnical parameters from the results of field investigations (the original purpose of Part 3 of the code) but also to define essential requirements for the corresponding equipment, test procedures and evaluation (differences to the German status of standardisation arc reported in [1, 21). Existing gaps arc filled by complementary national standards in different countries. Tiris has been done in Germany by adjusting the existing established DIN Standards to the corresponding ENV 1997-3 procedures. The following German standards are relevant to this chapter: • DIN 4020 Geotechnische Untersuchungen ftir bautechnische Zwecke (Geotcchnical investigations for civil engineering purposes) • DIN 4021 Baugrund - AufschluB durch Schilrfe und Bohrungen sowie Entnahme von Proben (Ground - Exploration by excavation, boring and sampling) • DIN 4022 Baugrund und Grundwasser (3 Teile) (Subsoil and groundwater, 3 Parts)
52
Klaus~Jiirgen
!\tfel.1.er and Ulf Bergdahl
• DIN 4023 Baugrund- und Wasscrbohrungcn; Zcichncrischc Darstcllungdcr Ergcbnissc
(Subsoil and water borings; graphic presentation of the results) • DIN 4030 Beurteilung betonangreifender Wasser, Biidcn und Gase (2 Teile) (Assessment of water, soil and gases for their aggressiveness to concrete) In particular DIN 4094, Baugrund - Erkundung
• DIN 4094-1 Drucksondierungen (CPT) (Cone penetration tests) • DIN 4094-2 Bohrlochrammsondierung (BDP) (Borehole dynamic probing)
• DIN 4094-3 Rammsondierungen (DP) (Dynamic probing) • DIN 4094-4 Fliigclschcrvcrsuchc (FVT) (Field vane test) o DIN 4094-5 Bohrlochaufwcitungsvcrsuchc (PMT) (Borehole deformation tests) In the meantitne, the CEN Technical Committee 34J on Geotechnical Investigation and Testing has been established to develop European Method Standards.
1.2 Preliminary investigations Preliminary investigations are necessary to decide: • whether a proposed structure can be constructed at all, at the intended location and to an acceptable cost, with regard to the ground conditions; • which technical and economic requirements for the design of the foundation, the structure and the construction have to be considered. In Germany, such preliminary investigations are the basis of the legal procedures for development planning. This means that these preliminary investigations have to also sho\\' \\'hat influences there are on the vicinity of the construction site, what environmental effects have to be considered and to what extent the ground in the vicinity of the planned structure can be loaded (e.g. by anchors). The extent of the investigation depends on the existing inforn1ation available, which especially in densely populated areas, may consist of geological maps, ground maps, ground expert opinions in the vicinity, aerial photography (important to the assessment of war damage), hydrological and geotechnical assessments, historical knowledge (tilled cavities, quarries, underground air raid shelters., caverns, cavities. in lime stone formations, old slopes or creeping slopes, mining activities etc.) and so on. In all other cases, ground and groundwater conditions have to be determined at least in a coarse grid. Hydrological data should be available for a significant period, generally, at least a full year. The same is applicable for meteorological data if construction in open \\'aters is considered.
1.3 Geotechnical fteld investigations
53
Preliminary investigations of soil and rock for the purpose of obtaining construction materials should give information on whether, where and to what amount suitable material is available, considering the economic aspects.
1.3 Design investigations Design investigations are the topic of this chapter. They consist of: • carrying out excavations, drilling, penetration tests and other tests for the determination of geotechnical parameters (sec Chapter 1.2); • determining the ground strata and all relevant gcotcchnical properties of the soil and rock necessary for the design, the invitation to bid, the construction and for the geotechnical observation of the behaviour a structure or for deciding on the suitability of materials for construction purposes; • determining potential difficulties during construction of the chosen foundation; • recovering soil and water samples from excavations and drilling (especially special samples for laboratory tests for the determination of gcotcchnical parameters). Field investigations include, in a wider sense: • load tests of foundation elements, for example spread foundations (ENV 1997-3, 11), of piles (Chapter 3.2 and ENV 1997-1, 7) or of anchors (Chapter 2.5 and ENV 1997-1, 8). These tests arc not covered by this chapter; • measurements of settlements and deformations which are treated in Chapters 1.11 and 1.12.
Additionally, reference is also made to DIN 4020, Supplement l, and relevant references (3-5]. It is in the hands of the engineer with geotechnical experience to design the ground investigation program and to select the tests to be carried out in such a way that the selection of tests, equipment or an intelligent combination of different methods results in the best technical and economical solution for the intended purpose. It is not always the "best" test equipment that ensures the most appropriate solution for given boundary conditions and circumstances.
2 Ground investigation by excavation, drilling and sampling 2.1
General
Trial pits, including headings (horizontal or with slight inclination) and shafts (vertical or with steep inclination), drilling and so called small-scale drilling are direct investigation methods which allow an inspection of soil and rock, their sampling and their performance evaluation in the field. Table 1 gives an overview of the suitability of some of the direct investigation methods for soils and rocks. Trial pits give the best investigation results because details of the ground strata and the soil condition can be clearly identified and high quality sampling is possible. However, the investigation only reaches moderate depths and is in general only possible above the groundwater level. The cost rises considerably with increasing investigation depth because of the need to retain the ground or possible groundwater lowering.
54
Klaus-Ji.irgen Melzer an
Table 1. Suitability of some direct investigation methods (following DIN 4020, Supplement 1, Table 5) I
1
2
4
Suitability of direct investigation methods Investigation n1ethods
Type of soil/rock strata, density
Weathering, state of loosening, (in rock)
Discontinuities (strata, cleavage, joints)
1
Existing, inspectable explorations
++
+-
+
2
Excavation
++
"
down lo moderate depth and above groundwater level 3
Shaft
++
+' also \\'idth of joints, filling, roughness, s1noothness, direction
++
++ also width of joints, filling, roughness, smoothness, direction
difficult in the presence of groundwater an
immediate lining is necessary 4
Heading
++
++
+j
5
Rotary core drilling
++
+
+
in rock in sliff cohesive soils
to judge rock 1naterial and filling of joints
in directional drilling, known st!'ata or combination of different drilling directions, with TV probing
j+
-
-
-
+
6
Percussive core drilling (also wilh tube or hose)
only in soils; narrow strata also detectable, often changes in density an
-~
7
Grab drilling (dry)
-
+
--
-
-
-
-
in soils at strata thickness of ::::: 50 cm; in adn1ixtures of coarse graveol, stones an
8
Flush drilling
+in conjunction with borehole geophysics: +
--
++ + +-
--
- -
--
very suitable, partly optimum investigation 1nethod suilable, generally sufficient results partly sufficient, sufficient only if supplementing by other investigation tnethods and for special problems not sufficient, partial results to he expected in exceptions only
55
1.3 Geotechnical field investigations
5
6
7
8
Suitability of direct investigation methods Geological faults
Investigation of groundwater
Borehole/ field-tests
++
+
-
++
++
+
Optimum san1pling, especially recommendable for invesligaling weathered/loosened zones in rock
++
+-
+-
For deep foundations in difficult ground, e.g. subways, power plants
often difficulties due to water and lack of wall stability
++
+
++
on heading level only
an tests possible
+
+
++
often core loss in faults
in rock
·water pressure tests, PBP tests, SPT,BDP n1easurements of primary stresses
+in soil
+-
-
only by comparing adjacent boreholes
-
++
only by comparing adjacent boreholes
only by correlating of boreholes
+
Remarks
Caverns, large tunnels, reservoir darns, field tests Most frequent investigation niethod. In case of complex geological conditions and difficult structures, completed by shafts and headings
permeability tests, SPT, BDP
High quality drilling method for coarse soils and changing strata, appropriate ·with difficult structures in such soils
+
Suitable for coarse soils
permeability tests, SPT, BDP
-
+ geophysical borehole measuremen ts
Simple deep exploration, groundwater gauge
56
Klaus-Jiirgen Melzer and Ult Bergdahl
1 \ \ ith drilling, soil and rock as well as \Vater samples can be obtained also fron1 greater depths and in addition, tests can be conducted in the borehole. The drilling itself is not hindered in case of groundi,vater, hoi,vever, the presence of the ground\vater has an influence on the selection of the sampling equipn1ent.
Small-scale drilling requires less sophisticated equipment compared to the norn1al drilling equipment, hcl\vever it generally provides only small samples unsuitable for soil mechanics investigations. Samples of higher quality (Table 4) can only seldomly be obtained. Therefore, DlN 4020 and DIN 4021 require that small-scale drilling is used for preliminary investigations under strong restrictions only and the drilling required for design investigations nlust not be replaced hy sn1all-scale drilling. The latter prin1arily serves the purpose of supplementing other investigation methods and for example of examining the ground at the base level of foundations. A combination of sn1all-scale drilling with sampling and high quality penetration testing may be used as well, for example in certain clays. The type and extent of the investigation depends on the lahoratory and field investigation programme designed by the geotechnical expert and with this on the type and extent of the planned structure. DIN 4020 (Section 6.2.4.3) gives guidelines for the spacing of investigation points (e.g. 20-40 m for high-rise and industrial structures) and for the investigation depth for simple structures. large-area structures (e.g. industrial complexes). linear structures (e.g. roads and airfields), special structures (e.g. bridges) and water retaining structures. The reference level for the investigation depth is the lowest level of the structure or the structural element or of the excavation depth respectively. In cases \Vhere the stability of slopes or effects on neighbouring structures have to be considered. the investigations have to he extended heyond the area covered hy the structure. DlN 4021 descrihes the investigation of the ground by excavations, drilling. small-scale drilling and sampling. In the follo\ving reference will be made to nlajor deviations fron1 ENV 1997-3, especially \Vhen essential requirements are not met by DIN 4021.
2.2 Investigation of soils In ENV 1997-2, Tahle 1, soil samples for lahoratory investigations are divided into five quality classes according to soil properties. remaining unchanged during the sampling process and the suhsequent treatment (transport etc.). This tahle was also included in ENV 1997-3. 12. The quality classes are described here in Table 2. Undoubtedly, the quality class of a sample for lahoratory tests, ohtained hy using a certain soil sampling method, will depend on the soil type and also significantly on the design of the sampler and the care taken during san1pling, transport, storage and handling in the laboratory. Quality classes l to 5 \Vere introduced in DIN 4021 for the first time in the early 1970's. The quality class describes what paran1eters and \vhat properties can be detern1ined fron1 one class of samples. The system is based on six parameters and properties: • Particle size Z • Water content V-.' • Density Q • Permeahility k • Linear modulus of elasticity Eoed • Shear strength 'tf
57
1.3 Geotechnical field investigations
Table 2. Quality classes of soil samples for laboratory investigations and corresponding sampling categories (after ENV 1997-3, Table 12.1) Soil properties I quality class
1
2
3
4
Unchanged soil properties particle size water content density, density index, permeability compressibility, shear strength
x x x x
x x x
x x
x
x x x x x x x
x x x x x x
x x
x x
x x
x
5
~
Properties that can be determined sequence of layers boundaries of strata, broad boundaries of strata, fine Atterberg limits., particle density, organic content water content density, density index, porosity, permeability con1pressibility, shear strength
Sampling category to be used
.,.,
.•••,
x
·• e•
f~\~~-<::_'(
::· Samples of the highest quality class (Class 1) retain all the indicated soil properties, nlost desirably unchanged. The state and composition of san1ples of the lowest quality class (Class 5) have been changed completely. These samples can only be used to draw conclusions regarding the ground layering. With the introduction of these quality classes, the selection of a suitable drilling and sampling method has improved. Only samples of a particular quality class are necessary to be san1pled to allov.r the correct detern1ination of the required soil parameters. Table 3 gives an overvie\V of drilling methods appropriate for certain soil types. Table 3 also sho\vs the quality class for laboratory tests that can be reached (column 9) and the soil parameters that can be determined from those samples (column 10). Table 4 shows probable applications of sn1all-scale sampling in soils. When drilling methods v.rith non-continuous sampling are applied, one sample has to be taken from each separate layer or each nleter for layers of considerable thickness. These samples should reflect the composition and state of the actual soil conditions as much as possible. Contrary to above, ENV 1997-3, 12.2.l uses an equipment related approach by characterising the san1pling nlethods by means of the follov.ring three sampling categories: • Category A: By using these methods, the intention is to obtain san1ples in v.1hich no or only slight disturbance of the soil structure has occurred during the sampling procedure or in handling of the san1ples. The v.rater content and the void ratio of the soil correspond to that in situ. No changes in constituents or in the chemical composition of the soil have occurred. • Category B: By using these methods, samples contain all the constituents of the soil in situ in their original proportions and the soil has retained its natural water content. The general arrangement of the different soil layers or components can be identified. The structure of the soil has been disturbed.
58
Klaus-Jtirgen Melzer and Ulf Bt:rgc.lahl
Table 3. Drilling methods in soils (after DIN 4021, Table 1) Column
1
2
4
3
5
Drilling method Linc
Soil loosening technique
lJsc of flu~l1ing
Extraction of sample by
6 Equipment
Drilling
Drilling tool
llorehok diameter raugeO
technique
medium
Drilling irn·ohing continuous coring 1
Rornry drilling
No
Rotary dry core
Singk-Lube core barrel
65 to 200
llollow-slernmed auger
6510 300
Single-tube core barrel
65 to 200
drilling
Yes
2
Drilling tool
Drilling tool
Rotary core drilliug
Double-tube core barrel -
3
100 Lo 200
Yes
Drilling Looi
Rotary core drilling
Doubk-tube wre barrel with screwed culling shoe
>lo
Drilling tool
Percussive core drilling
Percussive clay cutter with cutting edge inside; also with sleeve or ho~e, or hollowstemmed auger
Drilling involving continuou:i. 1'1!00\'ery of bulk samples R
Rotary drilliug
No
Drilling tool
Rotary drilliug
9
Percussion
>lo
Drilling tool
Light cable percus- \Vire line with percussion sion drilling shell auger
150to 500
10
Grabbing
No
Drilling Looi
Grab drilliug
400 LO 2500
Drill rods witl1 shdl auger or worm auger
-
100 to 2000
-
~
Wire line with gr
1) Guideline values. ) De is tht: internal tliamt:tt:r of tht: tlrilling tool. 3) The quality classes given in brackets can only be achieved in particularly favourable ground contlitions. 2
59
1.3 Geotechnical field investigations
R
7
--
9
Unsuitable for 1l
Coarse gravel, cobbles. boulders
10
Preferred method forll
Probable qualit)' Sample class (cf. Table 2) unaffected for soil as in with respect to column 8
Clay. silt, silty fine sand
4, (3 to 2)
-
11
Sample 4ualily3i
Applications and limitations
Remarks
Z,(w, (l)
Good interior, omside dried out
Z.w.
-
---
Clay, silt, sand. organic suil
3.(2to1)
((l, Eoed• "If, k)
Clay, clayey and cemented composite soils, boulders
Non-cohesive soils. silt
4, (3 to 2)
Z,w,
3. (210 1) Gravels, cobbles, boulders
-
-
Z,(V•",Q)
(Q. Eoc:d' 1 1"' k) ~
Z,w, (Eoed· 'f)
-
Cohesive soil: 2, (1)
Z,w. (g, Eocd· 1'.f, k)
Non-cohesive soil: 3, (2)
Z.(w)
Plotting of driving chart on the basis oI number of impacts
4
z
Cohesive soil: 2. (1)
(Eoed · 1:(, k)
2, (1 I
Oay. silt
(!,
soils with a parLid_e size Larger than De/3 2'
Dense soil~ with a parlide size Larger than De/3
Clay, silt and soils with a particle size up to D 0 /3
Gravel, soils '"'ilh a particle size up to De/3 ~
---
Composite and pure sands with a particle si1e Larger than 0.2 mm, as well as gravels, firm and stiff clays
Clay. silt, fine sand
Roulders. cobbles, gravel, dense sand
Soils with a particle size up to De/5
Z, W,Q,
Z.(w)
Cohesive soil: 2, (1)
Z,w. (Z.w,Q,k)
4.(.1)
-
-
Non-cohesive soil: 3. (2)
All soils above water table, all cohesive soils below water table
-
Non-cohesive soil: 4. (3)
---
Boulders of size larger than De/3
-
Z,(Z,w)
- -Z, (w), below
:Nlax:imum length water table (only of auger: 0.5 m from cuttings drilled with large diameter shell auger)
Gravel aOOvc water Lable, Clay '"d silt above Willer 4, (3) silt. sand and gravel below table. clay below water table water table
Z,(w)
Firm, cohesive soils, boulders of size larger LhanD 0 /2
Z,(w)
Gravel, boulders of si1e less Above water table: 3 thfln D'-'/2, cobbles BcloV•' water table: 5. (4)
(Z) --
-
60
Klaus-Jtirgen Melzer and Ulf Rergdahl
Table 3 (continued) Colmnn
1
3
2
4
5
6
Drilling inl'olving rerol'eQ of incomplete sum pies 1
11
Ro Lary
Yo;
Direct flushing
Yes
Reven;e flow of Reverse drilling fluid circulation drilling
chisd
No
Drilling tool
Light cable percussion drilling
Wire line with valve auger
IOU to 1000
'fo
Drilling tool/ auxiliary
Drilling by chisels
Wire line or drill rods, with chisels
100 to 1000
drilling 12
13
Percussion
Wash boring (rotary Jrilling)
Drill rcxls with roler bit, jct bit, step bit, etc. As in line 11, but with hollow
100 to 500
60 to 1000
-
~
14
flushing
1) Guideline values. 2l D.: is the internal diameter of the drilling tool. Table 4. Methods for small-scale drilling in soils (after DIN 4021, Table 3) C.olumn
j
3
2
- -
5
4
Drilling melhod 2l
6
Equipment
Soil loosening technique
·use of flushing medium
Extraction of s
Drilling technique
Drilling tool
Borehole diameter range 1l
1
Rotary drilling
'fo
With drilling tool
Hand auger drilling
Shell auger. worm auger or spir
60 to 80
2
Hammer driving
No
With drilling tool
Small-scale hammer driving
Hiililmcr-Uriving linkage, with tube sampler
30 to RO
3
Pneumatic
No
With drilling Looi
Small-scale pneumatic drilling
Pneumatic linkage, with tube sampler
30 to40
Linc
1> Guideline ,,alues. See limitations described in subclause 5.3. 3l De is the internal diameter of the drilling tool. 2)
• Category C: Here, the structure of the soil in the sample has been totally changed. The general arrangement of the different soil layers or components has been changed so that the in situ layers cannot be identified accurately. The water content may not represent the natural water content of the soil layer sampled. Table 2 defines which of the three categories A, B or C of the sampling methods should be used in order to obtain a corresponding quality class for laboratory tests. Using this a connection to DIN 4021 has been established.
-
61
1.3 Geotechnical lleld investigations
7
----
8
11
10
9
-
All soils
(5)
Samples unsuitable for soil mechanics tests
Limited to pcnctration ofirrclcvant upper strata
-
All soils
5. (4)
(Z), Z. if core pieces arc produced
-
Recovery from above water table
Gravel and sand in water
s. (4)
(ZI
Can also be used in cohesive soils if water is added
-
All soils, to remove obstructions
5
Samples unsuitable for soil mechanic:<. tests
-
-
3l The quality classes given in brackets can only be achieved in particularly favourable ground conditions.
7 Application~
Unsuitable for 1l
8
9
Preferred method for 1l
Coarse gravel with a particle size larger than De /3 3l and dense soils
Clay to medium gravel above water table; cohesive soils below water table
Soils with a particle size larger than De/2
Soils with a particle si?.e uptoD<;:/.'i
11
10 Samples4l
and limitations
Probable quality class (cf. Table 2) for soil as in column 8
Sample unaffcct- Remarks cd with respect
Above water table: 4, (3)
Z,(w)
Below water table: 4
z
Cohesive soil: 3. (2)
Z.w,(Q)
Non-cohesive soil: 4, (3)
Z,(w)
'· (2)
Z, w, (Q)
Lo
-
Only Lo be used [or small depths
-
Firm and coarse-grained soils 4)
Clay, sill. fine, sand
The quality classes given in brackets can only be achieved in particularly favourahle conditions.
While DIN 4021 only defines five quality classes that can be obtained, ENV1997-3, 12 specifies minimum requirements for the sampling equipment - especially for category A - to be used for taking samples of a required quality. On the other hand. DIN 4021 is generally more strict regarding certain dimensions such as inside diameters, cylinder length etc. [1, 2J because the Eurocode states essential requirements only. For each borehole, a qualified field foreman has to record the results on site in a borehole log according to DIN 4022 Part 1 which reflects the results of the drilling, using the
I
62
Klaus-JOrgen Melzer and Ulf Bergdahl
nomenclature for the different soil types specified in this standard. DIN 4022 Part 3 has to be applied when drilling methods with continuous core sample recovery are used because the sample material can be inspected only after opening the liner or tube. The fine strata also have to be described.
2.3 Investigation ofrocks The above quality classes for soils are not applicable for drilling in rock because other aspects are relevant to the assessment of rock properties, e.g. degree of weathering, discontinuities, joint planes, striping and dipping planes (see ENV 1997-1, 3,3,2), These topics are also detailed in DIN 402L Table 5 shows different drilling methods in rock with respect to suitability and results. Table 5. Drilling methods for rock investigations (after DIN 4021, Table 2) 1
Colu1nn
2
3
4
Drilling method Line
Breaking the rock
Flush. Extraction of medium samples by
5
=r
E4uipment Drilling techni4ue
1 llrilling involving continuous coring
Drilling tool
I, attached Drilling tool to drill rods
Rotary core drilling
Single-tube core barrel. usually with hardfaced j core
no
Drilling tool attached to drill rods
Rotary dry core drilling
Single-tube core barrel, with hardfaced core
Rotary drilling
yes
Drilling tool attached to drill rods
Rotary core drilling
Double-tube core barrel with hollow bit
4
Rotary drilling
yes
Drilling tool attached to drill rods
Rotary core drilling
Triple-tube core barrel
5
Rotary drilling
yes
Drilling tool attached to drill rods, with wireline extractable inner barrel
Wireline core drilling
Wireline core barrel with hollow bit, or triple-tube barrel
6
Rotary hammer driving
yes
Drilling tool attached to drill rods
Percussive rotary core drilling
Rotary percussion clay cutter
1
Rotary drilling
yes
2
Rotary drilling
3
2 Drilling involving recovery of incomplete samples 7
ll
Rotary drilling
Guidelines.
yes
Drilling tool attached to drill rods
Rotary open hole drilling
Solid bit, roller bit
63
1.3 Geotechnical field investigations
Frequently, rotary open hole drilling is used in rock and also in soils for preliminary investigations, for example to assess the level of rock surfaces, weathered zones in rock, or the occurrence of cobbles or boulders in soils. Besides the penetration resistance (measured in sec/0.2m of penetration), the following parameters can he recorded using the MWD-technique (Measuring While Drilling): pushing pressure, revolutions/min, applied torque, fluid pressure and fluid volume (J/min). Together with the drill mud flushed up, these parameters give indications of the ground layers penetrated. Based on these results, additional drilling using high-quality drilling methods is planned to determine the ground strata accurately and for core sampling purposes. Table 6 shows properties of rock materials and rock mass that can be determined by drilling. Fifteen mechanical properties of rock materials and rock mass respectively, are listed in the table which can, cannot, or can only partly be determined from drilling or from tests in the borehole.
6
Equipment Borehole outer dia111eter range 1 l
8
7
Drilling tnethod less suitable for 1 I
10
9
Satnple Cores 1l
Drill cuttings Ren1arks
100 to 200
Rock of tnediutn Jointed, soft rock to high hardness
100 to 200
Rock of medium Soft, erosive, water- None to high hardness sensitive rock; short core runs
To prevent overheating of the bit, core runs shoult not exceed 0.5 m
50 to 200
Erosive, watersensitive rock
All types of rock
As 1
-
50 to 200
-
All types of rock
A~
t
-
50 to 200
Erosive, watersensitive rock
All types of rock
As 1
-
Rock of medium Medium to hard to high hardness rock
Asl
With drive device at the equipment or as in-borehole hamtner
As 1
-
100 to 200
50 to 200
-
None
Sieve residue, Flushing tnedium tnay suspended cause disturbance of matter core material
Table 6. Charncteristics of rock material and rock mass that can be detc1mincd by drilling (after OTN 4021, Table 5)
1.3 Gt:ott:chnical field investigations
65
For each borehole, a qualified field foreman has to record the results on site in a borehole log according to DIN 4022 Part 2. The log shall contain all the facts and observations made during the core drilling in rock. In general, it has to be stated that ENV 1997-3, 13 is more comprehensive than DIN 4021 regarding rock sampling, especially in view of the requirements for sampling equipment and quality control [1, 2]. H'n.vever, DIN 4021 shows a higher degree of detail in descriptions of the drilling methods (Table 5) and the properties of rock materials and rock mass that can be determined from borings (Table 6). Similar to soil sampling, ENV 1997-3, 13 defines the following methods for rock sampling: • Category A: By using these sampling methods, the intention is to obtain samples in which no or only slight disturbance of the rock structure has occurred during the sampling procedure or in handling of the samples.. The strength and deformation properties, water content, density, porosity and permeability of the rock sample correspond to the in situ values. No change in constituents or in chemical composition of the rock mass has occurred. • Category B: By using these sampling methods, the samples contain all the constituents of the in situ rock mass in their original proportions and the rock pieces have retained their strength and deformation properties, water content, density, porosity. The discontinuities in the rock mass may be identified_ The structure of the rock mass has been disturbed and thereby the strength and deformation properties, water content, density, porosity and permeability of the rock mass itself. • Category C: By using these sampling methods, the structure of the rock mass and its discontinuities have been totally changed. The rock material may have been crushed. Some changes in constituents or in chemical composition of the rock material may have occurred. The rock type and its matrix, texture and fabric may be identified. Furthermore, ENV 1997-3, 13.2.3 defines the following parameters for the degree of rock recovery from rotary core drilling that should be evaluated in context, not individually: • Rock quality designation (RQD): The sum length of all core pieces that are 10 cm and longer, measured along the centre line of the core, expressed as a percentage of the total length of the core run. • Solid core recovery (SCR): The sum of the length of all core pieces, expressed as a percentage of the total length of the core run. A core piece nlust possess one full diameter but not necessarily a full circumference. • Total core recovery (TQR): The total length of core sample recovered, expressed as a percentage of the total length of the core run. ENV1997-3, 13.3.2 recommends the following methods as sampling techniques for the categories A-B that are inevitable depending on the structure and the decomposition grade of the rock and on the requirements of the laboratory testing to be performed: • Category A or B: Rotary core sampling in which a tube with a cutter at its lower end is rotated into the rock mass thereby processing a core sample. • Category A or B: Drive sampling in which a tube or a split-tube sampler having a sharp cutting edge at its lower end is forced into highly or completely weathered rock mass either by a static thrust or by dynamic impact. Drive samplers are usually piston samplers or open tube samplers.
66
Klaus-Jiirgen Melzer and Ulf Bergdahl
• Category C: Shell or auger sampling where the sample is taken from the actual drilling tool. • Category C: CuttiTI!::,'S sampling in which the rock mass, remoulded or crushed, by cable or rod handled percussion or cutting tools is brought up to the surface by means of a bailer or circulation of a transporting substance. • Category C: Block sampling made by hand cutting from a trial pit, shaft or heading or
by using specially made block samplers.
The selection of the appropriate method is to be made in accordance with the required sample quality for the classification of the rock mass and for the laboratory tests to be performed. Furthermore, precise requirements are defined for core barrels for sampling according to category A and for sampling with rotary core drilling for the categories A and B: ENV 1997-3, 13.3.3 and 13.4. Regarding rotary core drilling, special attention is drawn to the different requirements on equipment checks and controls before and during sampling operation: ENV 1997-3, 13.4.1. This move towards improvement of quality assurance is continued in the Eurocode in the requirements for the documentation (ENV 1997-3, 13.5). Here, the requirements of DIN 4021 and of ENV 1997-3 are in agreement regarding the labelling of the core samples. However, in the latter, information on the sampling category and the sampling equipment are added. In addition, a sampling log is required that must contain the usual information also detailed in DIN 4021 and DIN 4022Part 2 but in addition, the signature of the qualified field foreman or the project manager. The following details (ENV 1997-3, 13.5.2) have to be reported: -
date of sampling: position and elevation of drilling location; borehole direction, inclination and orientation; whenever possible the depth of the free groundwater level; the method of pre-drilling if used: the use of casing and depth of casing tip; the use of drilling lluid and the level of the drilling lluid in the borehole; colour and colour shifts of drilling fluid; loss, if any, of drilling lluid; drilling fluid pressure and circulated volume; the specification and type of sampler used; the diameter or the size of the sample; the depth (top and bottom of the sample) and the length of the sample; the core run interval; pressure on the cutting edge; the rock mass type, discontinuities and grade of decomposition based on the visual inspection of the sample by the field foreman and his judgement of the sampling category: - any obstructions and difliculties encountered during the sampling operation (including unsuccessful sampling attempts).
67
1.3 Geotechnical field investigations
2.4
Obtaining special samples
Methods for taking special samples are included in category A. 2.4.1
Soils
Special samples of fine-grained soils and sands can be obtained quite simply from the base of construction excavations and roads, foundation base levels, slopes and trial pits by means of thin walled cylinders with sharp cutting edges. This method is particularly suitable in cohesive soils of firm consistency and in fine sands of medium density. The test is standardised in DIN 18125 Part 2. In loose and dense cohesionless and cohesive soils of stiff and very stiff consistency, the equipment for obtaining special samples from trial pits according to DIN 4021 should be used (Fig. 1). Cubes with side length of 10 to 30 cm can also be cut out from cohesive soils. f>100 4>96
11 1 ·11
l
2
~ YI
.1
2 ~t:I
b)
3 I ..
I+ 5 ig
~9
6
~
7 8 •,
'
...
~·-··-m----·--r
'•
. ttf. 4s0 < a) . :
c)
·i···\\\. J.~~-J~._J//·······.· · ·
Fig.1. Ohtaining special samples from trial pits (after DIN 4021) a) Arrangement of sampler, b) Sampler tube, c) Sampling process 6 Guide hood 1 Percussion drill rods 2 Drop \\-'eight 7 Sampler tu be 3 Anvil 8 Guide plate 4 Driving device 9 End caps (sealed with adhesive tape) 10 Metal plate for limiting depth of penetration 5 Ring mark
68
Klaus-Jtirgen Melzer and U\f Bergdahl
Ohtaining undisturbed samples from boreholes is more difficult and time consuming because the normal drilling operation has to he interrupted. Nevertheless, it is necessary because only in this way will lahoratory investigations of soil properties yield reliable results. tlowever, it is not always possible to ohtain completely undisturhcd samples from cohesionless soils. In this case, penetration testing is suitable and generally sufiicicnt as a complementary investigation.
Table 6 of DIN 4021 contains details about obtaining special samples, the required equipment, the suitahility of various equipment and the achicvahlc laboratory quality classes of the sa1nples obtained with the corresponding equip1nent.
2.4.2 Rocks In general, only rotary core drilling is suitahlc for the ground investigation in rock hccausc only with this method it is possible to obtain sufficiently large and undisturbed core samples, accurate idcntiiication of the rock and the determination of the rock properties by strength tests (ENV 1997-3, 13). With rotary open hole and percussion drilling, cuttings arc only obtained which arc just suitahlc for the idcntiiication of the rock type. In water sensitive strata or in rock with strong discontinuities, double and triple-tube core barrels have to be used to avoid the flushing mediu1n disturbing the core sa1nple.
2.5
Investigation of groundwater conditions
ENV 1997-3, 14 contains the corresponding requirements for groundwater measurements. Furthermore, DIN 4021, 8 describes the different types of water in the ground and the prohlcms with groundwater ohscrvations during drilling operations. It stresses the point that groundwater gauging stations are necessary to obtain reliable data, and describes their arrangement for short and long-term ohscrvations. DIN 402I also contains guidance for measuring the direction of flow and the flow velocity of the groundwater and dcscrihcs how to obtain water sa1nples (for pumping tests see Chapter 2.9). It describes the test arrangements using single and douhlc packers necessary for drilling in rock to measure the water pressure in different aquifers, and the associated packer test for determining the permeability of the rock mass.1be following Figs. 2 to 7 show some examples. In the case of more than one aquifer and the borehole being drilled with a single run of casing, it is only possihlc to get an approximate measurement of the groundwater level or piezo1netric level in the upper1nost aquifer (Fig. 2a-d). In general, an adequate seal along the casing through each aquifer cannot be achieved. Therefore, the measurement of the piezometric level in the lower aquifer can he distorted (Fig. 2c and d). If the piezometric level of a second, lower aquifer is to he measured, the first run of casing has to be sealed by drilling into the aquiclude. A second run of a casing is brought down inside the first and the drilling continued until the lower aquifer has been reached (Fig. 2h); the piczomctric level of the lower aquifer can then he determined in the second casing. Packers also have to be installed for ineasuring the pressure head of groundwater in fissured rock. A single packer seals off the measuring section c, that reaches down to the bottom of the borehole (Fig. 3). Double packers can be used to define the measuring section c between the two packers (Fig. 4). If pressure heads in different joint systems arc to measured within one borehole, 1nultiple packers inust he used.
69
1.3 Geotechnical field investigations
d)
b)
Fig. 2. Possible effecl on waler level measuremenl when drilling lhrough an aquiclude (after DIN 4021) a) Correcl measuremenl of groundwater level b) Correcl measuren1ent of the piezometric level of the lower aquifer c) and d) Erroneous measuren1ent of the piezometric level of the lower aquifer 1 Groundwaler level 2 Piezometric level of lo'Yl-·er aquifer 3 Aquifer 4 Aquiclude (day layer). measured water level, direction of llow
2 3 4 5 6 7 8 9 10
Spreading device Inner tube of packer Outer tube of packer Annular space Rubber sleeve Piezomeler Clan1ps Observalion jar Recording pressure gauge Compressed air bottle
c Measuring section Fig. 3. Arrangement of single packer and air pressure gauge for water pressure measuremenls (after DIN 4021)
The arrangement of groundwater gauging stations obviously has to take into account the ground conditions, the hydrological requirements, the engineering task and the length of the observation period. An installation plan for each groundwater gauging station has to be documented (Figs. 5 to 7). The piezo1neter for a gauging station consists of a sump pipe, a filter pipe and extension pipes that can be closed off at the top v. hilst allowing for ventilation. The filter pipe is surrounded by filter sand (Fig. 5). Figs. 6 and 7 show examples 1
6 Perforated section of inner and outer tube 7 Clamps c Measuring section Fig. 4. Double packer arrangement (after DIN 4021)
5
__ , -~-·.
~I
·-.I::
I
. ..!.:
_. -B
I Cap
2 Extension pipe 3 4 5 6 7 8 9
Filter pipe Sump pipe Concrete cover Frost-resistant soil material Seal Drill cuttings Filter gravel
4
Fig. 5. Arrangement of a piezometer with free groundwater in the uppermost aquifer. For example above ground level with precautions against frost heave (after DIN 4021)
of various arrangements of groundwater gauging stations (top below and above ground level etc.). Contrary to DIN 4021, where mainly ground water measurements with open system are treated, ENV 1997-3, 14 covers in addition measurements of groundwater pressures with closed systems, i.e. the measurement of pore pressure in fine-grained soils. The requirements for records and the presentation of groundwater gauging station results are given in ENV 1997-3, 14.5 and 14.6 (see also DIN 4020, 8.1). DIN 4023 is relevant for the presentation itself.
Fig. 6. Arn1ngement of a single piezometer below ground level, v.·ith a group of aquifers (after DIN 4021)
1 2 3 4 5 6 7 8 9
Cap (tightly fitting) Casing Extension pipe Concrete cover Anchor Frost-resistant soil Frost line Seal Dill cuttings
Fig. 7. Arrangement of a ground\vater gauging station with the top above ground level and protection against frost heave (after DIN 4021)
3 3.1
Ground investigation by penetration testing General
For a penetration test, a thin rod is pushed or driven into the ground or turned around its longitudinal axis. From the magnitude of the penetration resistance and/or from its variation with depth conclusions can be drawn regarding the strength or sequence of the strata. Compared to trial pits, shafts, headings and drillings, penetration tests are regarded
72
Klaus-Jtirgen \lfelzer and Ulf Bergdahl
as indirect investigations, i_ c. direct visual inspection or sampling of the strata is generally not possible. Penetration tests arc indirect investigations which always have to be supplemented by direct investigations (e.g. key boreholes with sampling) for an accurate identification of the ground because the measured value of the "penetration resistance" hy itself docs not allow any conclusions regarding the soil type. On the other hand, the penetration resistance diagram can he used as additional information to allow the selection of sampling
depths. The derivation of geotechnical parameters has to be viewed carefully. Many investigations on the topic of establishing reliable relations between penetration resistance and gcotcchnical parameters, c. g. cohesion, angle of shearing resistance, modulus of elasticity either directly or indirectly (via consistency or relative density, etc.) have been made. Approaches to find direct relations between hearing capacity of foundation clements.
e.g. the skin friction and the pile resistance are also well known. However, the validity and the suitahility of such relations has to he evaluated critically for each case and area because of potential superimposing influences. For instance in cohesive soil, the penetration resistance at the pcnctromctcr tip can be relatively constant; it mayhc, however, that this result is falsified by skin friction along the rods. Difficulties can also occur in the interpretation of results ohtained in cohesion less soils. For instance, the penetration resistance depends not only on the relative density hut also on the degree of uniformity and the compactibility of the soil. In this case, the determination of the relative density is valuable only if the grain size distribution or the maximum and minimum voids respectively, arc known [6-lOJ. Especially in silty cohesionless soils, the measured penetration resistance can be higher than the one corresponding to the actual relative density, due to false cohesion. Peaks also occur in the penetration resistance measured in gravelly soils because of cobble admixtures. These peaks should be disregarded in the evaluation. The widespread use of penetration testing in practice and numerous research programmes have through the years led to equipment related improvements giving re producible results at compatible conditions and to reliable relations for the derivation of geotechnical parameters, for example [l 1-13]. However, it has to be pointed out that all possibilities to derive geotechnical parameters shown in the following sections are examples that are valid only for the corresponding conditions investigated (c. g. soil types etc.), because it is not possihle to establish relations which are valid world-wide. Furthermore, the origin of the individual examples has to be observed. For instance, all equations regarding dynamic probing, cone penetration tests and borehole dynamic probing as quoted from DIN 4094 in the following sections, are deterministic relations taken as conservative estimates. Other examples have been taken from statistical regression analysis or arc just tables with a range of gcotcchnical parameters. Therefore, different safety concepts have to he considered to suit the application. For this reason, it is recommended that the original source is checked for a closer review of the corresponding examples and that any local experience is collected. In the meantime, the development of some penetrometers and the presentation of the test results are being coordinated on international level [14, 15J and were initially harmonised on European level in the ENY 1997-3. Among other tests, this document contains the essential requirements for the following tests:
1.3 Gcotcchnical field investigations • • • • •
73
Cone penetration test ( CPT) Standard penetration test (SPT) Dynamic probing (DP) Field vane test (FVT) Weight sounding test (WST)
The Ger1nan standardisation work in the newly edited DIN 4094 is consistent with these international efforts..
3.2
3.2.1
Dynamic probing
Equipment and test procedures
Dynamic probing as mentioned in ENV-1, 1.3.3.10.2, and in accordance with DIN 4094 is the in situ 1neasure1nent of the penetration resistance from driving a cone vertically into the ground. A hammer of a given mass at a constant height of fall is used to drive the cone, while the nu1nber of blo\vs Nio for a penetration depth of 10 cm is counted (ENV 1997-3, 6 also allows N2o). The dynamic penetrometer consists of a cone and preferable hollow rods. Co1n1non penetron1eters are listed in Table 7. In the new edition of DIN 4094-3, only the light penetrometer DPL, the heavy penetrometer DPH and the superheavy penetrometer DPG (hammer mass= 200 kg, height of fall = 50cm, cone cross section = 50 cm 2 [16, 17]) appear in the standard itself. The light pcnclrometer DPL-5 and the medium heavy pcnctrometers DPM, which are used on a regional level only, appear in an inforn1ative annex. ENV 1997-3, 6 abJTces generally with DIN 4094 except in Table 7 above, which conlains a DPSH wilh the dimensions of the standard penetration test instead of the DPL-5. The trend to penetro1neters with higher hammer masses can be ohservcd also more recently in Japan, Canada and the U.S.A. [18]; the background to this is the desire to he ahle to also investigate strata of very high strength e.g. Lills, gravels, soft rock etc .. The diameter of the cone is somewhat larger than that of the rod to reduce skin friction, allo\ving the penetration resistance of the cone to he measured more accurately (Fig. 8). Retrieving the penetrometer from the ground is easier if it is equipped with a sacrificial cone that is not fixed to the rod instead of a retainable cone. The rods have to be turned at least 1,5 revolutions after each meter of penetration to the reduce skin friction and to ensure that the rod threads are kept tight. If a torque measuring \vrench is used, the skin friction can be estimated from the n1easured torque. To avoid skin friction, a fluid n1ediun1 (preferable water of drinking quality) can be injected through horizontal or upwards holes in the hollo\v rods near the cone. Son1eti1nes, a casing is used for the sa1ne purpose.
l_
Fig. 8. Cone for dynamic probing (d = l; after DIN 4094-3)
Table 7. Types and suitability of equip1nent for penetration testing (after DIN 4094, Table 1) 1 No. Designation
3
2
4
I
6
5
7
Code
2
3
4
5
6
7
D}'namic probing light
DPL
DPL-5 Light dynamic penetrometer
SPT CPT
12 '
Of limited use in (classi.ficatio 11fler DIN 4022 Part 1)
Remarks, pn:viuus codes
1-ledium and dense gravels, firm clayey 11nd silly soils
Previously LRS 10
Tip diamcter 1)
Ma'>S of Height hammer 1l of fall
OD/ID of rod 2 1
"·=
m, kg
h,m
mm
10
35.7 ±0.3
110
0.50 ±0.01
2216
25.2 ±0.2
110
0.50 ±0.01
22/6
6
Nrn
8
±0.l
35.7 ±0.3
30 ±0.3
0.50 ±0.01
3219
18
N10
2D
Dense gravels
Previously MRSB
35.7 ±0.3
30 ±0.3
0.20 ±0.01
22/6
6
N10
15
Dense gravel-,. firm cbyey <1nd silty soils
43.7 ±0.3
50
3219
18
N10
25
-
±0.5
0.50 ±0.01
Application only in certain ~gion!>. previously MRS A Previour.ly SRS 15
50.5 ±0.S
163.5 ±0.5
0.76 ±0.02
Wlo
30
N3o
35.7 ±0.3
1-
-
321-
-
lj_c, f~
5
D}'namic DPM 10 prohing medium Medium DPM-A 10 dynamic pi::m:trometer Dynamic DPH 15 probing heavy Standard Penetration Ter.t Cone penetration
11
Cone tip area Ac, cm2
I
1
10 8 9 Equipment 1-leasured Max. test .\tfass of driving device values 4l depth belov·· ~tarting point without t,mSl hammer 31 max, kg JO 6 N10
20
10
±0.l
Clayey and silty Application only soils and dense in certain regions, coar:;e-gr11ined soils previously LRS 5
I
te~t
0.45 61
-
rods 40
Soils with stones, dense gravels, fi1m clayey and silty soib
'
lJ Manufacturing tolerances 2 l Manufacturing tolerances need not to he specified 3 l This consists of the driven parts (anvil and guide rod) but excludes the penetrometer. Moving parts for raising and releasing the hammer arc also excluded 4 l \Vhere N10 is the number of blows per 10cm penetration depth, N:m th~ number of hlows per 30c1n penetration depth, Qc is the cone penetration resistance in MPa. f is the local unit skin friction in MPa 5l Approximate values for soil conditions of mediu1n strength 6l The starting point is the respective bottom of the borehole
Electrical cone (CPT-E) or mechanical cone (CPT-M)
I
1.3 Geotechnical field investigations
3.2.2
75
Evaluation
3.2.2.1 General The results of dynan1ic probing can be evaluated qualitatively if • strata are investigated by drilling and sampling; • the homogeneity or inhomogeneity, respectively, of the ground, or of a fill, is to be evaluated; • especially loose or firm layers in fills or the rock surface (with heavy equipn1ent) are to be investigated; • compaction controls arc to be performed. by comparing the penetration resistances before and after compaction (see also Chapter 2.12 of Volume 2 of this Handbook). The following investigation depths can normally be reached using the different dynamic probing methods: DPL: 10 m: DPM: 20 m; DPH: 25 m: DPSH: 25 m; DPG: 40 m. In DIN 4094-3 exan1ples are given of equipment and gcotcchnical related influences to be observed in the evaluation of the test results. One of these, is that the penetration resistance in cohesionless soils with the san1e relative density, is lower below the groundwater level than above at the same conditions. Corresponding relations to correct for this influence, are given in DIN 4094-3.
3.2.2.2 Derivation of geotechnica1 parameters Shear strength The results from dynan1ic probing tests are used nlainly to derive the strength and con1prcssibility of primarily cohesionless soils. First, an example is shown on how to derive indirectly the angle of effective shearing resistance q/ fron1 results of dynan1ic probing (see also ENV 1997-3, Annex E.1 and DIN 4094-3). Extensive investigations have demonstrated [7] that the following general equation represents the best relationship between the penetration resistance (in this case the nun1ber of blows Nio) and the relative density of cohesion less soils:
lu = a1
+ a2
logN10
(1)
Table 8 contains examples of the coefficients shown in Eq. (1) for different cohesion less soils for both the light (DPL) and heavy (DPH) dynamic penetrometers. The equations are valid for penetration tests performed above the groundwater level and for a depth larger than about 1 m where this is the critical depth from which under the same conditions, the cone penetration resistance becomes almost constant. Above this depth, the cone penetration resistance increases considerably with depth. The angle of effective shearing resistance q:/ can then be determined fron1 tests or by proven relationships using the relative density In, predicted from the dynamic probing test results, with the above equations. ENV1997-3, Annex D.3 contains an exan1ple of the relation between In and cp' for silica sands, which arc differentiated qualitatively by the degree of uniforn1ity and grain size. A practical exan1ple for deriving the angle of shearing resistance of gravelly soils by means of this indirect method, in conjunction with the design of harbour sheet pile walls, is described in [9J.
76
Klaus-Jiirgen Melzer and lJlf Ilergdahl
Table 8. Examples of cot:-ffir.:ients in Eqs. (1), (3) and (4) for deriving rt:-lativt:- density 10 and stiffnt:-ss coefficient v from results of dynamic probing above groundwater level (after bNV 1997-3 and DIN 4094-J) Soil Conditions Relative density Iv Stiffness coefficient v r.:lassifir.:ation Ic 2) (DIN 18196) DPL DPH DPL DPH
u''
a,
a2
SE
::;3
-
0.15
0.260
SW,GW
~6
-
-
-
TL.TM
-
0.75-1.30
-
-
a,
•2
b1
b2
b1
b2
0.10
0.435
71
214
161
249
-0.14
0.550
-
-
-
30
-
4
50
-
6
I) Dt:-gret:- of uniformity, d60/d10· 2> Consistency, unit:
Valid ranges: For the relative density: 3 ::: Nio ::; 50. For the stiffness coefCicient in Sb: with DPL: 4 ::; Nio ::S 50; with DPlI: 3 ::= Nio ::=: 10. For the stiffness coefficient in TL, TM: with DPL: 6::: Nio ::,:: 19; with DPH: 3 :::: N10 ::; 13. Soil r.:lassification according to DIN 18196: SE: poorly graded sands; S\:V: well graded sands; G\:V: well graded sand-gravel mixtures; 'IT.: low plastir.:ity days; TM: medium plastir.:ity days.
Compressibility
The following is an example of deriving directly the stress dependent modulus of linear elasticity from results of dynamic probing tests ahove the groundwater level (ENV 1997-3, Annex E.3 and DIN 4094-3). The definition of the modulus of linear elasticity Eoed derived from oedometer tests and used for the calculation of the settlement of spread foundations, is the basis for the determination of compressihility: Eoed = V ·Pa [(uv'
, ]w + 0.5up)/p,
(2)
with: v
w UV'
u~
p, Wp WJ.
~
stiffness coefficient stiffness exponent; for sands and sand-gravel mixtures: w = 0.5; for slightly plastic clays with low plasticity index (wp ~ 10; WL ~ 35): w = 0.6 effective vertical stress at the hase of the foundation or at any depth he low it due to the overburden of the soil effective vertical stress caused hy the structure at the hase of the foundation or at any depth below it atmospheric pressure plasticity index liquid limit
If soil shear deformations during the settlement process are to be considered, the corresponding modulus of linear elasticity can be assumed to be approximately 0.75 Eoed·
77
1.3 Geotechnical field investigations
Corresponding investigations in cohesionless and cohesive soils [7, 19) resulted in the following equations for the determination of the stiffness exponent v of Eq. (2) above. • For sands and gravelly sands:
v = bi
• For slightly and medium plastic clays:
+ b2
log N io
v = b 1 + b, · N10
(3) (4)
The stress dependent modulus Eo~d· according to Eq. (2), can then he derived directly using the coefficients from Table 8 for determining v from Eqs. (3) and (4) and with w = 0.5 for cohesionless soils and w = 0.6 for cohesive soils. 3.2.2.3 Bearing capacity of piles
Results from dynamic prohing tests have hcen used for some time to predict the drive ability of piles and sheet piles. as well as of the bearing capacity of piles (see EN 1997-1, 7 and also [16, 17, 20-221). This is due to similarities in the driving and testing techniques used. 3.2.2.4 Relations between the results from different penetration tests
It should also he noted that a nun1her of relations have heen estahlished hetween the results from different dynamic prohing tests and het"'rccn these and the results of standard penetration and cone penetration tests, see DIN 4094 and e.g. [7. 19]. It has to be stated that the various types of penetrometers have different penetrability and sensitivity for variations in soil types. It can therefore he appropriate to use different types of penetrometers in parallel for a certain project to obtain the best information about the ground to be investigated in the n1ost econon1ical \Vay possihlc. 3.3 3.3.l
Standard penetration test Equipment and test procedures
The standard penetration test mentioned in EN 1997-1, 3.3.10.2, covers, according to ENV 1997-3, 5, the determination of the resistance of the soil at the hot tom of a horcholc to the dynamic penetration of a split barrel sampler and the recovery of disturbed samples for soil identification purposes. The test consists in driving the san1pler (outer diameter: 51 n1m, inner dian1eter: 35 mm) by dropping a hammer of 63.5 kg mass from a height of 76 cm on to an anvil. The number of blows necessary to achieve a penetration of the san1pler of 30 cm (after its penetration under gravity and below a seating drive of 1.'1 cn1) is defined as the penetration resistance N (blows/JO cm penetration). The standard penetration test is the oldest form of dynamicprohing [11, 23]. Its first known use goes back to heginning of the 20th century. The original atten1pts to standardise the equipment stem from the early 1930's in the U.S.A. Even today, the standard penetration test is the n1ost \Videly used in situ test for bearing capacity and stability investigations [12, 24]: ref. [11] contains a very good survey. The hcst known standards for SPT are ASTMD 1586 in North America and BS 1377 in Great Britain. which are referred to on a world-wide hasis. National standards arc also availahlc, for example in Australia, Brasilia, Denn1ark, India, Japan and Sweden. With the "International Test Procedures
78
Klaus-Ji..irgen Mel:rer and Ulf Bergdahl
for SPT" [14], the Technical Committee TC 16 of ISSMFE succeeded for the first time in harmonising the test on an international level. This document \\'as the basis for ENV 1997-3, 5. However, difficulties occurred in interpreting the results because the actual value of the energy which is induced in the rods has to be known. In addition, energy losses can occur
due to the rods in the borehole not being supported. Today, methods for determining these energy losses or corresponding values from experience arc available fll, 24-26], and are induced in various standards, e.g. ASTM D 4633 and ENV 1997-3, 5. Recent extensions of the equipment and the test procedures include devices for measur-
ing the torque at the rods [23, 27] to obtain additional information about the soil types encountered. In the early 1950's in Germany, this uncertainty about the energy losses led to an essential modification of the equipment, whilst maintaining the original technical specifications (height of fall etc.), with the following aims: • transmission of as much of the full energy as possible on to the anvil; • essential reduction of the rod influence. The e4uipment was improved [3, 28, 29] by the drive mechanism being encapsulated in a water tight casing directly above the penetrometer (Fig. 9) .The equipment as a whole is lo\\'ered into the borehole. The hammer is released by means of an automatic release mechanism. The sampler is closed by a cone (apex angle = 60::i) because disturbed or
2
J
5
6
x
d
l Rope 2 Stuffing box 3 Automatic release mechanism 4 llammer 5 Water proof casing 6 Anvil 7 Penetrometer
3 to 6 Driving device
FiJ!:. 9. Borehole dynamic probing BDP (after DIN 4094)
L3 Geotechnical field investigations
79
undisturbed samples (depending on the soil type) can be obtained from the borehole itself between penetration tests. The use of a closed sampler or solid penetrometer (of about 90cn1 length) with a cone (apex angle: 60°) for performing tests in gravelly soils and in soft weathered rock is also current practice in countries like Australia, Great Britain, Portugal, Spain and South Africa [11]. The abbreviation for this test is generally SPT(C) or SPT(cone). Since the early 1950's, the equipment described in Fig. 9 was standardised in DIN 4094 and also covered by ENY 1997-3, 5. It is defined as follows - Borehole dynamic probing (BDP) is a penetration test where the penetron1eter is driven into the ground from the bottom of a borehole for a defined penetration depth. As in the case of SPT, the number of blows NJo is determined for a penetration depth of 30 cm after the penetration under gravity and an initial drive of 15 cn1. Recently, the use of additional weights, mounted directly above the penetrometer, are recommended for investigation depths of > 20 m below water level (see DIN 4094-2 and [30]). Special care has to be taken when performing the test in cohesionless soils below the groundwater level - for example, the soil below the bottom of the borehole could be disturbed by the drilling. Using drilling tools causing suction, should be avoided. It is also possible that the penetration test is performed with the casing in soil subjected to buoyancy. 'fhe sol I would then be constrained between the penetrometer and the casing, leading to an increased nun1ber of blows. Therefore, a lowered water level in the casing has to be avoided by for example, maintaining the water or drilling fluid level in the borehole at a sufficiently high level at all times. The standard penetration test is primarily perforn1ed in key boreholes to obtain indications about the strength and deformation properties of the ground.
3.3.2 E>alualion 3.3.2.1 General
The opportunities to applying SPT results for different design purposes is considerable. The test is mostly used for the determination of strength and deformation characteristics of cohesionless soils, however, valuable data can be also detern1ined for other types of soil under certain circumstances, e.g. [31 ]. Table 9 gives an overview of the current application of SPT results on an international level for geotechnical design. ENY 1997-3, 5 and [11] give examples of corresponding applications. In applying relations between the SPT results and geotechnical paran1eters, the following conditions should be considered in addition to the effects of the different performance of the test and the equipment used. The soil type to which a relationship was established has to be descTibcd because relative density not only influences the number of blows in cohesionless soils but also the compactibility, the grain size and possible cementing [7, 32]. This of course affects the derivation of geotechnical parameters. The same is also valid for the other penetration test methods covered by this chapter, e.g. [6--9]. It is also necessary to know whether, and by which nlethod, the nun1ber of blows used in the relation, has been corrected in respect of said energy losses. As for the dynamic probing tests (Section 3.2.2.1 ), the following has to be considered in evaluating the SPT results obtained in cohesionless soils; at the same relative density, the penetration resistance below the groundwater level is sn1aller than above the groundwater
80
I
Table 9. Exan1ples of the application of SPT results in international geotechnical design (following [11]) Derivation of geotechnical pa:ramete1s
• • • • •
Angle of shearing resi~tance of cohesionlesl-i soils Undrained shear strength of clay~ Unconfined compressive strength of weak rocks Modulus of elasticity or stiffness coefficient. respectively, of cohesionless and cohesive soils Maximum shear modulus
Direct calculations
• • • • • •
Settlements of spread foundations on sand Acceptable bearing pressure of foundations on sand Acceptable bearing pressure of rafts on sand Li4uef<:1ction potential of sands Shaft and end resistance of piles Sheet pile drive ability
level. DIN 4094-2 gives some relationships to correct for this effect (see also Section 3.3-2-2).
3.3.2.2 Derivation of geotechnical parameters Shear strength The follo\.ving example shows a possible method of deriving indirectly the angle of effective shearing resistance qi' for cohesionless soils. Sitnilarly to Eq. (1), the following general relation between number of blows N3o and relative density In applies: TD = c1
+ c2
log N3o
(5)
Table 10 shows examples for the coefficients c1 and c2 in Eq. (5) for BDP results obtained in different cohesionless soils above the groundwater level. Using the relative density In, determined from SPT/BDP results, the angle of effective shearing resistance q::' can be derived. For instance, DIN 1054 and ENV1997-3, Annex D.3 contain corresponding estimations of cp' for different cohesionless soils. For more detailed investigations of the relation between the penetration resistance of dynamic penetrometers and cone penetrometers in cohesionless soils on the one hand and their relative density and angle of shearing resistance on the other hand, reference is made to [6-11, 28, 33-35J. A good overview of the options to derive the shear parameters for cohesive soils, limestone and soft rock can be found in [11, p. 83 ff.] 1• Compressibility Similarly to the evaluation of dynatnic probing results, the stiffness coefficient v in Eq. (2) can be derived directly from the number of blo\.vs NJo as shown in the following example.
1 In the following: text, ref. [11] is the seconda.ry reference; the page numbers refer to the original source.
81
1.3 Geotechnical field investigations
Table 10. Examples of coefficients in Eqs. (5) to (7) for deriving relative densi1y Io and stiffness coefficient v from BDP results (after DIN 4094) Soil classification (DIN 18196)
Conditions
Relative Density lo
,,
d,
d,
0.10
O.JRS
146
217
0.18
0.370
-
-
-0.03
0.455
-
-
-
un
lc 2)
above GW
under GW
SE
:53
-
x
-
SE
:53
-
-
x
e>6
-
x x
SW,GW -
-
Stiffness coefficient v
c,
~-
TL,TM
-
0.75LJO
-
50
4
-~
1> Degree
of uniformity, d 60 /d 10 . 2 l Consistency, unit:
Valid ranges: For the relative density: 3 :::=: NJ(J :;: 50. Por the stiffness coefficient: in SE: 3 ::: N3o _:: 25; in TL, TM: 3 _:: N 3o ::: 23. Soil classification according to DIN 18196: SE: poorly graded sands; SW: well graded sands; GW: well graded sand-gravel mixtures; TL: lo\v plasticity clays; Ttvf: n1ediun1 plasticity clays.
Investigations into cohcsionlcss and cohesive soils [7, 19] resulted in the following equations to determine directly the stiffness coefficient in Eq. (2):
+ d1
• For sands:
v = di
log N3o
• For slightly and medium plastic clays:
v = d 1 + d2 · N,o
(6) (7)
With the coefficients d 1 and d1 from Table 10, the stiffness cocffidcnt v can be derived and by applying w = 0.5 for sands and w = 0.6 for the cohesive soils considered, the stress dependent modulus of elasticity Eoed is determined. ENV 1997-3, Annex D.4 gives an exan1ple for determining directly the settlen1ent of spread foundations in cohcsionlcss soils from SPT results. 3.3.2.3 Bearing capacit:y of' spread foundations and piles
Spread foundations Numerous attempts have been made since the late 1940's to determine the bearing capacity of spread foundations in cohesionless soils fron1 SPT results. However, these have to be accepted as methods that result in rough estimates only. On the other hand, some methods, developed during the last 25 years, use statistical evaluations of settlement observations of structures as a basis for determining the relationships between allowable bearing pressure, settlements, foundation geometry and SPT results [11, p. 95 ff.]. Even in these cases however, large deviations can occur. Because of these uncertainties, international practice prefers to derive the geotechnical paran1eters for shear strength and con1pressibility
82
Klaus-Jtirgen Melzer and Ult Bergdahl
from SPT results and use these as input to the design methods when only SPT results are available.
Piles Boundary conditions are more favourable for determining the bearing capacity of piles
(pile base resistance, shaft resistance) from SPT results. Methods are available for cohesive and cohesionless soils., limestone and soft rocks [11, p. 101 ff.]. These are mainly based on the results from pile load tests on various pile types. The approach is similar to that used in Germany (Section 3.4.2.3). 3.3.2.4 Relations heh,.'een the results from different penetration tests Finally, reference has to be made to relationships established between SPT/BDP results
and those from dynamic probing and cone penetration tests in DIN 4094 and ref& [7, 19, 36-38] (see also Section 3.4.2.4).
3.4
Cone penetration lest
3.4.1 Equipment and test procedures The cone penetration test (CPT) - mentioned in EN 1997-1, 3.3.10.1 and according to ENV 1997-3~ 3 - consists of a penetrometer being pushed vertically into the soil at a relatively constant rate of penetration of 2 cm/sec. The penetrometer comprises a series of rods ending in a penetrometer tip, consisting of a cone and a cylindrical shaft. During the penetration, the resistance of the cone and, if possible, the local friction on a sleeve (friction sleeve) located in the cylindrical shaft are measured. The cone resistance qc (penetration resistance Oc divided by the cross sectional area of the cone Ac) and the local unit skin friction fs (frictional force Os acting on the sleeve divided by its area As) are used for further evaluation.
Today the electrical cone is the most used equipment on a world-wide basis. One example of this is shown in Fig. 10. Generally, the cone has a cross sectional area of 10 cm 2 . During the recent 10 years, a cone with a cross sectional area of 15 cm2 (followed by a series of rods with a cross sectional area of 10cm 2 ) has also come into use [12] to increase the penetration depth and measurement accuracy but also to allow the incorporation of other measuring devices into the cone. Compared to the electrical cone, other equipment [12] for example the mechanical ("Dutch") cone penetrometer is now only seldomly used. At the beginning of the 1970's, the additional measurement of the pore water pressure using the so called piezocone was introduced. According to ENV 1997-3, 3, the cone penetration test CPTU is a CPT which includes the measurement of the pore water pressure generated at the base of the cone during the penetration. Fig. 11 shows an electrical cone, and Fig. 12 shows the corresponding definitions. Other equipment allows the measurement of the pore water pressure in the middle of the cone and at a defined distance above the friction sleeve [12]. Corrections and the methods of interpreting these test results are also given in reference [12]. The widespread use of the cone penetration test also outside Europe, which started in the 1970's., increased the need for international harmonisation. Initially, recommendations for CPT were made by the Technical Committee TC 16 of the ISSMFE [14] which were
83
1.3 Geotechnical field investigations
Push rods
Gap and seal
2 3 4
5 6 7 8
Cone, Ac= 10cm 2 , apex angle = 60" Load cell Strain gauges Friction sleeve, A~= 150cm2 Adjustment ring Waterproof cable bushing Signal cable Rod connection
Fig. 10. 'lip of the cone penetration test CPT (after DIN 4094)
Possible Friction sleeve
Fig.11. Scheme of a piezocone for the cone penetration test CPTU (after ENV 1997-3)
,_____..,
do - 35.7 mm
Fig. 12. Critical dimensions for a piezocone test CPTU (after ENV 1997-3)
followed recently by recommendations for the CPTU [15]. These y,.·ere also the basis for ENV 1997-3, 3. In addition, there exist a number of comprehensive national standards for example in the Netherlands, Nory,.·ay, Sweden and the U.S.A. [12]. As the content of DIN 4094 was less comprehensive, it made lts new edition, DIN 4094-1, even more important. During the last 30 years, the cone penetration test has gone through an enormous development not only because of its widespread use outside Europe, but also because of changes to the equipment. Ref. [12] 2 gives an excellent overview of the current situation and [39J summarises the state of the art for aspects like earthquake and environmental engineering. Today the standard penetration test is perhaps still the leading test used world-wide. 2 In the following text. [12J is the secondary reference; the page number refers to the original source.
84
Klaus-Jurgen Melzer and Ulf Bergdahl
1-iowever, the cone penetration test has reached or even passed it for many applications because of a higher accuracy in the interpretation of the test results and the numerous possibilities it offers in equipment and operations. This has also intensified the investigations on comparisons of the results from the two tests methods l40J, to transfer the SPT know hov..· to the evaluation of the cone penetration test results (and vice versa). In the long run, the application of the two test methods will most probably follow the trend in Germany during the last 50 years: Within a ground investigation project, the cone penetration test (or if applicable, a corresponding dynamic probing test) will be used as the main test, and the SPT will be used as a valuable supplement in the key boreholes, including sampling as for example required by DIN 4094 or EN 1997-1, 3. The state of development of the technical equipment is as follows: Generally, a deadweight of up to 100 kN is sufficient to overcome the total penetration resistance. The dead v..·eight is usually obtained by self-propelled thrust machines (mostly trucks). For light weight machines the counterweight can be increased by using screw anchors. A corresponding measuring capacity of the cone of 50 MPa is usually sufficient to measure the cone penetration resistance. There is also off-shore equipment now available, with dead-weights up to 200 kN, for penetrations of very stiff and very dense grounds (tertiary clays, glacial sands, soft rock). The corresponding cones possess measuring capacities of up to 120 MPa [12, p. 8ff., 41 ]. By comparison, the cone penetration resistance is usually low in cohesive soils; a value of 5 MPa can already characterise a soil of very stiff consistency [42] and values of q, < 1.5 MPa could indicate a firm to stiff consistency [33]. This means, hov..·cvcr, that an "all-purpose cone" [41] would need to have a measuring capacity from 1.5-120 MPa which is hardly feasible. Therefore, [15] recommends cone classes of different measuring capacities depending on the required use. While the electrical cones with and without the capability to measure pore water pressure, belong to today's standard equipment, the following additional measuring devices made possible by fast sensor development, were introduced for practical applications during the last ten years: • Cones for measuring lateral stresses [12, p. 172ff.J o Cones with prcssurcmctcrs [12, p. 175 ff., 43, 44] • Cones for seismic measurements [12, p. 179 ff., 45] o Acoustic cones [12, p. 190 ff., 46, 47] • Cones for measuring permeability [12, p. 80 f., 48] • Cones with liquid samplers to obtain pore fluid for chemical investigations [12, p. 199 ff., 49, 50] • Cone for measuring electrical conductivity/resistivity [12, p. 193 f., 49, 50] • Cones for radiometric measurements [12, p. 186 ff., 51-53] • Cones v..·ith build-in cameras [54, 55] o Vibratory cones [12, p. 132] 3.4.2
E•aluation
3.4.2.1 General
The aim of the evaluation of cone pcnctrometcr test results is principally the same as that for the results from the dynamic probing and the standard penetration test. The primary aim is a qualitative evaluation of the ground strata (together with the results from key boreholes). However, in this case the sensitivity is larger than from the dynamic pcnctrometcrs.
85
1.3 Geotechnical field investigations
The ability to measure the local unit side friction fs on the friction sleeve, in addition to the cone penetration resistance qc, has already led early to the use of the parameters q, and C as a means to classify the penetrated soil strata [56, 57], Fig, 13 shows such an exan1ple. Further investigations demonstrated [12. p. 51 ff.. 58, 59] that the accuracy of the prediction can be itnproved by using the corrected cone penetration resistance q1 (Fig. 12) and/or the pore water pressure itself instead of qc. This had led to the recon1mendation in ENV 1997-3, 3 to use the results from investigations with the CPTU for soil classification purposes. Further improvements were ohtained using refined statistical evaluations f59], additional evaluation methods (e, g, Fuzzi logic) [60] or by using cameras in the cone, There is no doubt that classification systen1s, as the exa1nple in Fig. 13 shows, can be a valuable tool in identifying the penetrated soil strata. However, it must he stated that such a syste1n established for a certain geographic/geological region cannot necessarily be applied in other areas without additional calihration (61]. This fact was confirmed hy recent con1parisons of different classification syste1ns (58]. Therefore, DIN 4094-1 and ENV 1997-3 insist that in addition to indirect ground investigations (here: cone penetration tests), direct ground investigations (e.g. key boreholes) with san1pling and laboratory investigations are also perlormed. Thanks to the variety in available 1neasuring techniques, a considerable nun1ber of parameters representative of different soil properties can now be quantitatively determined. Table 1 in [12] presents a good overview of what can be obtained with con1n1on field testing. Besides the consolidation ratio, sensitivity, permeahility etc., the following evaluation options should be mentioned: the description of in situ stress conditions including the coefficient of earth pressure [12, p, 61 [, 88[. 172ffl, seismic properties [43]. soil liquefaction [12, p. 166 !., 39, 62-64]. porewater pressure distribution [12. p. 74 ff.] and with increasing investigations of soil contamination, the quality of pore liquid, electrical resistivity and conductivity [12, p. 194 ff., 49, 50]. The application of cone penetration test 100 Friction ratio R1= f/qr. · 100
50 20 ~
"'~
ill
10 5
-
=-"'-=-
-~
§
2
i.
0,5
§
-=
=-==~--
----CLJIY---T_-_- 1----_
u
0,2 0,1 0
2
3
4 5 Friction ratio, %
•
7
•
9
10
Fig. 13. Exiunple of a semi-logarithmic relation between the cone penetnition resisnince rind the friction ratio in various soils (designation after DIN 4022) fron11neasure1nents by the GEOSOND company
86
Klaus-Jiirgen Melzer and Ulf Bergdahl
results in geotechnical design has, on an international level, at least reached and mayhe
even passed, the application of SPT results (see Table 9). Advanced 1neasuring techniques and nun1erous basic investigations in test chan1bers [12, p. 291 ff., 65, 66] have both contrihuted to this fact. In these tests, the influences of individual para1neters, such as in situ stress conditions, have been investigated systematically. This has contrihuted essentially, at least qualitatively, to the clarification of the penetration processes around the cone. However, the results cannot he directly transferred to reality hccause of limited test con-
ditions (partially too small chambers; "non-grown" soils) [12, p. 291ff., 39] although the transfer was possihlc in so1ne isolated cases as descrihed in [67].
3.4.2.2 Derivation of geotechnical parameters Shear strength
In the follo\ving sections, so1ne examples are presented for deriving geotechnical paran1eters from CPT results. Firstly, there are two examples of the indirect determination of the angle of effective shearing resistance (r' of cohesionless soils. In these cases, the relative density In is initially derived and, by the means of this parameter, qi' can he determined fron1 a corresponding relationship. From Eqs. (1) and (5), confirmed by recent investigations [12. p. 81 ff.], DIN 4094 gives the following general equation as an exa1nple of the derivation of the relative density Io from the cone penetration resistance qc:
lo=
e1
+ ez
(8)
logq,
Table 11 contains exa1nples of the coefficients e1 and e1 for sands and sand-gravel n1ixtures for CPT \¥ith a 10-cm 2 cone. 'lbese relations are valid for CPT performed ahove the groundwater level and for a depth larger than about 1 m where this is the critical depth Table 11. Exan1ples of coefficients in Eqs. (8), (13) and (14) for deriving relative density lo and stiffness coefficient,. from cone penetration resistance qc (in MPa; 10-cm2-cone) above groundwater level (after DIN 4094)
I)
Soil classification (DIN 18196)
Conditions
Relative density lo
uri
Ic2J
er
e,
fr
f,
SE
~3
-
-0.33
0.73
113
167
SW
:':6
-
-
-
-13
463
SW,GW
:':6
-
0.25
0.31
-
-
TL,TM
-
0.75-1.30
-
50
15.2
-
Stiffness coefficient v
Degree ofuniformil)'. d60/d10· 2 l Consistency, unit:
Valid ranges (in MPa): For the relative density; 3 ::=: qc ::=: 30. For the stiffness coefficient: in SE, SW: 5 :::; qc :::; 30: in TL, TM: 0.6 _::: qc _::: 3.5. Soil classification according to DIN 18196: SE: poorly graded sands; SW: well graded sands; GW: well graded sand-gravel mixtures; TL; low plasticity clays; TM: niedium plasticity clays.
87
1.3 Geotechnical fie:-ld investigations Table 12. Example:- of a re: la lion be:-l\Y't::-en cone penelration resislance ql: ( 1O-cm 2 -cone) and relative density In tor naturally moist medium sands (atter r68-70])
Cone penetration resistance qc, MPa
Relative density
<2.5 2.5-7.5 7.5-15.0 15.0-25.0 > 25.0
<0.15 0.15-0.3.1 0.35-0.65 0.65-0.85 > 0.85
Designation
ln Very loose Loose Medium dense Dense Very dense
from which under the same conditions, the cone penetration resistance becomes almost constant. Above this depth, the cone penetration resistance increases considerably with depth. The second example (Table 12) shows in tabular form, the relation between cone penetration resistance qc and relative density lu for moist uniform medium sands ("Berlin Sands") based on nu1nerous tests [68-70]. With comparable boundary conditions, this relation could be used also for deriving indirectly q:i' from qc via lu. Comparative penetration tests have shown that the cone penetration resistance in nonuniform cohesion1ess soils is smaller than in uniform soils at the same relative density [6, 7]. This is due to the higher co1npactibility::: (emax - emin)/ernin of the non-unifor1n soi1s. Additiona1 investigatlons [7, 8] revealed that not only the compactibility, but also the average grain size, influences the cone penetration resistance at the same relative density (Section 3.l). This means that an abso1ute determination of the relative density frotn the cone penetration resistance itself is not possible. For this, the grain size distribution and the maximum and minimum void ratios must be known. In addition, the existence of groundwater has a certain influence on the penetration resistance. Consequently, DIN 4094 differentiates between cases of "with and without groundwater" in lhc correlations for deriving In (see also Tables 8, 10 and 11). The tables also show the boundary conditions (soil types etc.) for which the correlations have been estabhshed. In case of the CPT, however, the influence of the ground water may often be negligible. A number of theoretical and empirica1 invcstigatlons of the relationship between the relative density lo and the angle of effective shearing resistance q:r' are also available in [9, 12, p. 90ff., 35, 39] and ENV 1997-3. Most of lhcsc relations arc dependent on the type of cohesionless soil investigated. 1be stress dependency of q/ is also increasingly being considered. Some exa1nples for deriving the angle of effective shearing resistance q:i' directly, are given below. The similarity of a cone penetration test with a deep foundation led to attempts to derive q:r' empirically as well theoretically from the cone penetration resistance, as can be seen fro1n a number of investigations [12, p. 90ff., 39]. Sotne of the references refer again to the stress dependency of q/. The dependency on the type of cohesionless soil is also 1nentioned. Table 13 shows a tabular relationship between cone penetration resistance 4c and the angle of shearing resistance
88
Klaus-Jurgen Ivielzer and Clf Bergdahl
Table 13.Example of a relation for deriving the angle of shearing resistance q/ and the drained Young's
1nodulus Em fro1n cone penetration resistance qc for natural cohesionless soils (quartz- and feldspar sands) (after [71])
Cone penetration resistance q.;;,:\1Pa
Angle of shearing
Drained Young's modulus2 1
resistance 1) cp', deg.
Em, :\1Pa
0-25 2.5 5.0 5.0-10.0 10.0-20.0 >20.0
29-32 32-35 35-37 37-40 40-42
<10 W20 20-30 30-60 60-90
1> The
values are valid for sands. Por silty soils a reduction of3° should be made. For gravels 2° should be added. 2> Em is approximated by the stress and time dependent secant modulus. Values given for the drained inodulus correspond to settleinents for 10 years. They are obtained assu1ning that the vertical stress distribution fol!O\\'S the 2:1 approximation [71, p. 64ff.]. Purthermore, some investigations indicate that these values can be 50 °/o lo\ver in silty soils and 50 °/o higher in gravelly soils. In overconsolidated cohesionless soils the modulus can be considerably higher. When calculating settlements for ground pressures greater than 2/3 of the design bearing pressures in ultimate li1nit state, the inodulus should be set to half of the values given in this table.
gravel mixture (U
=
5.7) [10, 70, 73, 74]. !his relation can be described by the following
equation with a valid range of 6.9 MPa < qc < 42.5 MPa:
q/
=
26.8 + 4.5 lnq, ± l''
(9)
with qc in MPa. The trend and order of magnitude agree with the deterministic equation in DIN 4094-1
for narrowly graded sands (SE, U <: 3) within the range of 5 MPa <: q, <: 28 MPa:
qi
= 23
+ 13.5 log q,
(10)
with qc in MPa.
Theoretical and empirical investigations are also available for the derivation of the undrained shear strength cu in cohesive soils [12, p. 63 ff., 33, 39, 75]. As examples, the two follov..·ing equations are mentioned, which were also included in ENV 1997-3, 3.
With q, from CPT: Cu=
(q, - Ovo)/Nk
(11)
But preferably with q1 from CPTU: Cu= (q, - Ovo)/Nkt
(12)
where: total vertical stress (due to overburden) factors to be estimated from local experience Nk can take values betv..·een 11and19 and N1 1 between 8 and 20 respectively depending on the actual cohesive soil and its plasticity index [12, S. 64 ff., 75]
89
l .3 Gcotcchnical field investigations
Compressibility
Especially in international practice, the Young's modulus Em is frequently used as a geotechnical design parameter. Investigations under controlled laboratory conditions indicate that Em in sand under drained conditions depends primarily on the relative density, the consolidation ratio and the actual stress condition. (~onscqucntly, the methods for the determination of Em from the cone penetration resistance qc reflect this fact [12, p. 93). A simple example is given in Table 13 [71) (sec also ENV 1997-3, Annex B.l ). The Eurocode contains further examples for settlement calculations of spread foundations in sands (ENV 1997-3, Annex B.2 and [12, p. 158f.)). For similar investigations in cohesive soils reference is made to [12, p. 71 f.] and [33). In Germany, the n1odulus Eoed fron1 oedometer tests is primarily used for settlement calculations. The same type of investigations, as used for the dynamic penetration tests in cohesionless and cohesive soils [7, 19], resulted in the following equations for deriving the stiffness coefficient v in Eq. (2) directly from the cone penetration resistance q, (in MPa): • For sands:
v = f1
+ fz ·log qc
• For slightly and medium plastic clays:
(13)
v = f1
+ 12 · q,
(14)
Examples of the coefficients f1 and fz are given in Table 11. lJsing w = 0.5 for the investigated sands and w = 0.6 for the corresponding cohesive soils, the stress dependent n1odulus Eoed can be derived directly. On the other hand, a direct correlation between the modulus of elasticity Eoed fron1 settlement measurements (c. g. from plate loading tests •Nith model foundations) and the cone penetration qc is not possible because paran1eters such as loading conditions, shape and size of the foundation and thickness of the compressible layer beneath the foundation, have an additional influence [10]. The •Nell known relation Eued = uqc (for values of a sec [33] and ENV1997-3, Annex B.3) should therefore be considered as a rough approxi1nation only. 3.4.2.3 Bearing capacit)· of' spread foundations and piles
Spread foundations As already mentioned when discussing the relation bet\\ een cone penetration resistance and angle of shearing resistance (Section 3.4.2.2), it \Vas obvious to correlate theoretically the coefficients in the equation for calculating the bearing resistance of spread foundations \Vith qc because of the similarity of a cone penetrating the ground. In practice however, early successful attempts were made to estimate the bearing resistance of spread foundation directly from CPT results [12, p. 157 f.]. In Germany, the evaluation of numerous large-scale load tests showed that cone penetration resistance and bearing resistance arc directly proportional to each other [10, 70, 73, 74]. This method for direct application in the design of spread foundations is reflected in the standardisation (sec DIN1054). 1
90
Klaus-Jtirgen Melzer and Ulf Bergdahl
Piles The determination of pile bearing resistance (see also EN 1997-1, 7.6.2.3) can be viewed as the original intention of quantitative evaluation of the cone penetration test results
because the transferability of the results appeared to be obvious. Consequently, there are more empirical approaches, mainly validated by pile load tests, available today than theoretical methods. State of the art methods are detailed in ll2, p.151, 76] respectively. There are, however, indications that the empirical use of CPTU results is more accurate
[77]. ENV 1997-3, Annex B.4 gives an example of a common method that stems from the early use of deriving pile bearing resistance from CPT.
Gern1any primarily followed the approach to correlate the results from pile load tests with the CPT results. This was based on a large number of related parameters (ultimate pile
base bearing resistance, normalised settlement = settlement/pile diameter) from pile load test results in cohesionless soils with known cone penetration resistances Qc from which conservative estimates were taken [78]. These comprehensive investigations are retlected in the Gern1an standardisation codes (DIN 1054, DIN EN 1536)_ These standards contain required minimum values of cone penetration resistances in the ground in the case of the bearing resistance of driven displacement piles_ For bored piles, values for pile base resistance and skin friction are given as a function of the cone penetration resistance from CPT within a range of 10 MPa < qc < 25 Ivlpa. For the pile base resistance, the normalised settlement of the pile head is given as additional parameter.
3.4.2.4 Relations between results fron1 different penetration tests The opportunity to derive the bearing capacity of foundations directly from the results of cone penetration tests, has led to numerous relationships between the results from different penetration test methods (e.g. SPT and CPT), see l7, 12, p. 149ff., 36-38, 40]. These efforts were enforced to utilise and to complement the comprehensive existing knowledge for the future. Table 14 shows exan1ples of some of these relationships.
3,5 3.5.1
Field vane test Equipment and test procedures
The field vane test (FVT) is an in situ test (ENV 1997-3, 8); it is performed with a rectangular vane, consisting of four plates fixed at 9o:i angles to each other, pushed from the bottom of a borehole (or excavation pit) to the desired depth and then rotated (loaded by torque). The ratio of the height Hof the vanes to the diameter D must be 2:1. The vane should be equipped with a device that allows the torque of the vane to be separated from that of the extension rods. A casing or a slip coupling can be used for this purpose. The test is used in very soft to very stiff cohesive soils to determine the undrained shear strength and sensitivity. It maybe used also to determine the undrained shear strength of silts and clayey glacier deposits. The reliability of the test results varies with soil type. In Germany, the field vane test has been standardised in DIN 4096, which was adapted to suit ENV 1997-3, 8 by the new edition DIN 4094-4. The equipment consists of the vane apparatus i.e. the vane and shaft (with protective sleeve, if appropriate), the rotating device, the extension rods (if appropriate) and the measuring device for measuring the torque and the angle of rotation (if appropriate) (Fig. 14).
91
l.J Geotechnical field in ...·estigations
Table 14. Examples of the average ratios of cone penetration resistance Qc {in MPa) to number of blows N3o and N 1o, respectively, for son1e cohesionless and cohesive soils above groundwater level (following DIN 4094 and [19]) Ratios of penetration test results
Soil classification (DIN 18196)
Cohesive soils
Cohesionless soils
BDP
DPH
DPL
BDP
DPH
DPL
q.;;/N_~o
40::/N10
Qc/N10
40::/N3<1
Q1:/N10
4c/N10
SE
0.5
0.7
0.25
-
-
SW. Sl
0.7
LO
0.35
-
GE,GW,GI
1.1
1.5
'
-
TL,TM 11 1i
-
-
-
-
-
-
0.55
1.00
For le= 0.75 - 1.30.
Valid ranges: BDP: in SE: J ::S N 3o:::: 50; in SW, SI: J ~ NJo ~ 40; in GE, GW, GI: 3 ~ N 30 :-:; JO; in TL, TM: 3 < N3o ~ 14. DPH: in SE, SW, SI: 3:::: Nm:::: 30; in TL, TM: 3:::: N 10 :::: 19. DPL: in SE: J :-:; N1o <: 60; in SW, SI: 3 :S N 1o ::S 25; in TL. TM: 9 ::S Nio ::S 60. Soil classification according to DIN 18196: SE: poorly graded sands; SW: well graded sands; SI: poorly graded sands with son1e grain dian1eter missing: GE: poorly graded sand and gravel; GW: well graded sand and gravel; GI: poorly graded sand and gravel with son1e grain dian1eter missing; TL: low plasticily clays; TM: medium plas.ticity clays.
p"'
Top
view-~
Rotation angle
/ /
I
P1
I
Initial/ position
--- Spring for torque measurement: T=2P·a
Casing
Position before test
~~
H~w_
T
---- Position during test ~
'' _J1--
Vane {top view) Fig~
14. Schen1e of the field vane test
0.36
Klaus-Jtirgen Melzer and Ulf Bergdahl
92
The vane height/diameter ratios are H/D = 100 mm/50 mm (FVT 50) and H/D = 150 mm/75 mm (FVT 75). The selection of the vane dimensions depends on the strength of the soil. For example FVT 75 is for low consistency and FVT 50 is for higher consistency. The vane apparatus is pushed into the soil until the required depth is reached; driving, vibrating or rotating are not allowed during the push-in process. When a casing is used to reduce skin friction, the apparatus is pushed into the soil only after the casing has reached a required depth; then, the apparatus is rotated; the required depth should be ::: 5 D but at least 0.3 m below the bottom of the borehole/pit. The rate of rotation should be 0.5°/sec in soft soils at low sensitivity and 0.1-0.2°/sec in soils with high sensitivity respectively. The maximum torque Tmax,u required to shear the soil along the undisturbed cylindrical soil surface for the first time, is measured (the angle of rotation is also recorded on occasions to obtain additional infor1nation about the shear behaviour of the soil). After the initial shearing process and the recording of Tmax.ii, the vane is rotated at least ten titnes with a rate of rotation of 10°/sec. After that, the above shearing procedure is repeated and the maximum torque T lllilx for this remoulded condition is recorded. 3.5.2
Evaluation
3.5.2.1 General The 1naxi1nu1n shearing resistance is determined by the following fortnula frotn the 1neasured torque, with Das vane diameter, assuming a simplified stress distribution along the failure surfaces of the sheared soil cylinder [3, 79]: Cfv =
0.273 T max.u/D 3
(15)
where: Cfv = maximum shearing resistance of the soil during the initial shearing process T max,u = maximum torque during the initial shearing process = vane dia1neter D
For the determination of the shearing resistance for the remoulded condition is replaced by T max:
c,,
=
Crv,
T max,u
0.273 Tm,,/D 1
(16)
The sensitivity Srv determined from the field vane test is defined as the ratio Crv/Crv· 3.5.2.2 Derivation of geotechnical parameters The 1neasured shearing resistance cannot be separated into effective friction and cohesion because the effective horizontal stress conditions in the soil being investigated are not known. Therefore, the field vane test can only be applied where the soil can be assumed to be frictionless for undrained conditions., i.e. in saturated normally consolidated cohesive soils of soft to stiff consistency. The shearing resistance Cfv can then be determined from the FVT as equivalent to the shear strength cru during soil failure under undrained conditions (for normal clays). At low shear stresses - for example creep movements in slopes - the shear strength of high plasticity clays is smaller [80]. Therefore, the shearing resistance obtained from FVT has to be corrected by means of empirical factors: Cfu=µ·Cfv
(17)
1.3 Geotechnical field investigations
93
The correction factor µhas to be determined from local experience. ln general, it is correlated to the plasticity index or the liquid limit and perhaps to the effective normal stress. The correction factorlncreascs in the case of ovcrconsolldated clays wlth increasing plasticity index [81---83] or in the case of norn1ally consolidated clays with decreasing liquid limit [84]. ln other cases - for example earth pressure calculations - the derived Cfu values are considered as nlinimun1 values because they were nleasured prin1arily ln vertical failure planes, where they are under normal conditions smaller than in horizontal or inclined planes. In these cases, Cfu can be increased [81 J. Examples of the correction factorµ arc given in ENV 1997-3, Annex G and in DIN 4094-4. ln fissured clays and in heavily silty or sandy clays., the correction factorµ has sometimes to be reduced to as low as 0.3. The undrained shear strength Cfu derived from the results of field vane tests is mainly used for the calculation of bearing resistance of spread foundations and piles or for stability analyses of slopes using analytical methods. The use of common field tests in environmental lnvestigatlons (see also Section 3.4.1) has also led to the first appllcatlons of the field vane test in this area of site investigations [85].
3.6
Weight sounding test
3.6.l Equipment and test procedures The weight sounding test (WST) mentioned in EN 1997-1, 3.3.10.3 was developed by the geotechnical department of the Swedish Railway Administration in about 1915 and became a national standard by 1917. lOday, the method is the most commonly used penetration test in Scandinavia and Finland. The weight sounding test is normally used for preliminary investigations in differing soils. The test results could also be used for design and inspection investigations ln nlost con1n1on soils but are primarily applied in very soft to stiff cohesive soils and very loose to dense coheslonlcss soils. ln very dense sand and gravel and tills pre-drilling could be necessary. The results are generally used to evaluate the thickness and extent of different soil layers but also for the assessment of the design parameters for spread foundations and piles. The first international harn1onisation of the weight sounding test took place in 1989 [14]. The method was also included in the European standardisation (ENY 1997-3, 7). The weight pcnetrometcr in its original form consists of a screw shaped point (diameter: 25 mm), a set of weights (1 x 5 kg, 2 x JO kg and 3 x 25 kg), a number of rods (diameter: 22 mm) and a handle (Fig. 15). The point ls manufactured fron1 a 25 mm square steel bar with a total length of 200 mm. The bar has an 80 mm long pyramidal tip and is twisted one turn to the left over a length of 130 mm (see ENY 1997-3, Fig. 7.1). It is used in general as a static pcnetromcter in very soft and very loose soils where the penetration resistance is less than 1 kN (corresponding to a total load of 100 kg). The weight sounding test can be performed manually or mechanically. Today, most tests arc performed mechanically (by hydraulic nlachines) and the recording of loads and number of halfturns is made automatically by means of electrical sensors. ln the static phase of the test, the penetron1eter should be loaded in stages as follows: 0.05 kN, 0.15 kN, 0.25 kN, 0.50 kN, 0.75 kN and 1.00 kN. The load is then adjusted from these standard loads to keep the penetration rate at about 50 mm/sec. If the penetration resistance is greater than 1.00 kN or the penetration rate is less than about 20 mm/sec the
94
Klaus-Jiirgen Melzer and Ulf Bcrgdahl
I
~:;;Jc~
Weights 25 kg
n=--
Weights l 0 kg Clamp 5 kg
100m
h,",-~-~~-~~~,,,;,;4/'}l;;;,:_:,_~
Wood
WST22
0
2
Scraper
rL
1.00m
6
0_80m
Rod, diameter: 22 mm
_____ Screw-shaped
___i _L__ __
0.20 m
cight sounding test
8 0 1
cone
Fig. L.c;. Test equipment for the manual \ \1
7
Load,kN
0204060 h110,2 m
Fig.16. WST results presentation WST 22:
Weight sounding test, Ro
ht/0.2 tn: fh(Sp~80):
Revolution per 0.2 m penetration Pre-boring to the designated depth (encrusted :-.urface layer); dia1nctcr: 80 mm
pcnctro1nctcr has to be rotated. The load of 1.00 kN is then maintained and the nun1bcr of halfturns required to give a 0.2 m of penetration is recorded.
The \\'eight sounding test is terminated at a depth when a certain penetration resistance is reached or when the penetron1ctcr cannot be driven any deeper, i.e. the so called "firm bottom" for the weight penetrometer is reached. 111e criteria chosen for the termination of a weight sounding test depends on the ground conditions and the purpose of the investigations. When the so called "firm bottom" criteria is used, the final resistance should be checked by sledgehammering on top of the penetrometer, by blows using the \\'eights or by a percussion machine to ensure the "firm bottom" has actually been reached. The chosen procedure must be recorded in the test report. The results fron1 a weight sounding test are presented in diagrams showing the penetration resistance versus depth (Fig. 16).
3.6.2
Evaluation
3.6.2.1 General
When considering the evaluation of the weight sounding diagran1s., both the n1agnitude of resistance and its variations are used. One has to remember that the variations of the resistance can also depend on the variations in the soil layer sequence. In very soft to firm
95
1.3 Geotechnical field investigations
clays the penetration resistance is often less than 1 kN or the resistance against turning is rather constant and low with less than 10 halfturns/0_2 m of penetration. As the sensitivity of the clay also influences the penetration resistance, the strength of the clay cannot be determined directly from the penetration resistance without a separate calibration from each site.
In very loose to loose sediments of silt and sand rather low and constant penetration resistances are also obtained. In medium dense to dense silts and fine sands higher (10-30 halfturns/0.2 m of penetration) resistances arc obtained, which remain rather constant with depth. In sand and gravel sediments, the variation in penetration resistance increases with the grain size. ""11en evaluating the weight sounding test results from silty sands and coarse gravel, one should note that a high penetration resistance does not always correspond to higher density or strength and deformation properties. The soil layer sequence evaluated from the weight sounding tests and any additional sampling on a site, including the "firm bottom" criteria, is used for the evaluation of the suitability of a site for a certain structure, for the evaluation of the type of foundation {spread or pile foundation) and for the derivation of geotechnical parameters.
3.6.2.2 Derivation of geotechnical parameters Weight sounding test results arc used as the basis for the design of foundations in cohcsionless soils. In l71 j, it is shown how shear strength and deformation properties can be derived from weight sounding test results and can be used as input for the design methods used for spread foundations (Table 15). In fine-grained silts and clayey cohesionlesssoils, gcotcchnical parameters should be determined by specific tests - for example by in situ pressuremeter tests or in the laboratory using good quality samples. Table 15. Example of a relation for deriving the angle of shearing resistance q/ and the drained Young's modulus Em for natural cohesionless soils (quartz- and feldspar sands) from the results of weight sounding tests (after [71])
Weight sounding test. half rcvolutions/0,2 n1
Angle of shearing resistance l 'i er'' deg.
Drained Young's modulus 2 l Em,MPa
0-10 10-30 20-50 40-90 >90
29-32 32-35 35-37 37-40 40-42
10-20 20-30 30--60
<10
60-90
lJ lhe values are valid for sands. For silty soils a reduction of 3"' should be niade. For gravels 2° should be added. 2 J Em is approximated by the stress and time dependent secant modulus. Values given for the drained modulus correspond to settlen1ents for 10 years. They are obtained assun1ing that the vertical stress distribution follows the 2:1 approximation [71, p. 64ff.]. Furthermore, some investigations indicate that these values can be 50 °/u lower in silty soils and 50 °/u higher in gravelly soils. ln overconsolidated cohesionless soils the modulus can be considerably higher. When calculating settlements for ground pressures greater than 2/3 of the design bearing pressures in ultimate limit state, the modulus should he set to half of the values given in this table.
96
Klaus-.Ttirgen Melzer and lJlf Rergdahl
3.6.2.3 Bearing capacity of piles The weight sounding test results can also he used directly for pile design [86]. The required length of end hearing concrete piles can he determined from the so called
"firn1 botton1" criteria where the tests have been terminated. Norn1ally, the required end bearing pressure is achieved at the "firm bottom" or up to 2 m deeper. However, dynamic probing is considered to be a n1ore accurate n1ethod in determining the length of such piles. Norwegian experience with friction piles indicates how the average weight sounding resistance along the pile length can also be used to calculate the n1agnitude of the skin friction (hearing capacity of the pile) in sands.
4 Lateral pressure tests in boreholes 4.1
Equipment and test procedures
The equipment for lateral pressure tests in boreholes (see also ENV 1997-3, 10-4) can generally he defined as follows f87]: The equipment normally consists of a cylindrical device that can apply a uniform pressure to the pocket \Vall in soil and rock; the pocket is created specially for the test. The term "pocket" is intentionally used rather than horcholc
to distinguish between the pocket created specially for the lateral pressure test and the horehole created for advancing hctwcen test positions. The borehole dian1eter should
be either equal or larger than the pocket diameter. Methods for creating the pocket are summarised in Tahle 2 of f87]. During the test, the volume change and the radial or
lateral displacen1ent of the cylindrical device are measured. From the results, strength and deformation properties of soils and rock, as \veil as for fills (quality control), can he derived. The first investigations of this type were described by Kogler [88, 89]. In the 1930'~ he developed a lateral pressure device (horehole jacking probe) where two cylindrical halfshells are pressed mechanically against the pocket wall. This device was later replaced by a cylindrical probe closed on all sides by a rubber nlembrane \.Vith steel plates at the top and the bottom of the cylinder. The probe was inllated by air pressure [90, 91]. ln the 1950's, this method was developed further hy MCnard into the three-cell pressuren1eter test (upper guard cell, test cell, lower guard cell). The lateral pressure device developed by Goodman for application in rock f92] should also be noted. After this, the world-wide dissemination of the "prehored pressuremcter" (PBP) of the Menard type and of other devices began. In France f93] and Great Britain f94], "self-horing pressuren1eters" (SBP) for applications in soil and rock were developed independent of each other, to reduce, as far as possible, the disturhance of the pocket walls during drilling and lowering of the probe, with the drilling device integrated into the probe. Finally, offshore application initiated a third generation of lateral pressure tests: the "pushed-in" or "full displacement pressuremeter" (FOP) [43, 44, 95]. In this case, the pressuremeter is incorporated with the cone in a cone penetrometer. In the hroadest sense, the NGI dilatometer, as further development of the !lat dilatometer (DMT) hy Marchetti, belongs to this group of tests [96, 97]. An overview of the present state of development of lateral pressure tests in boreholes can be found in [87].
97
l.3 Geotechnical field investigations
A numhcr of countries have now standardised certain devices such as the prebored prcssuremeter. The best known are probably the French Standard NF P94-110 and the American ASTM D 4719. Other standards for self-boring pressure meters arc in preparation. The first international harmonisation on an European level is described in ENV 1997-3, 4, where the essential requirements for the equipment and test procedures arc defined. Suhsequently, DIN 4094-5 was written to cover the equipment commonly used in Germany (PBP). Whilst internationally the equipment is divided into the three methods of hringing the probes into place (see above), DIN 4094-5 describes the equipment commonly used in Germany (PBP) as follows: The dilatometer (Fig. 17) is a cylindrical device where a flexible rubber membrane is used to apply uniform pressure (by gas or fluid) to the walls of the pocket (borehole). The displacement of the pocket is measured by displacement transducers in selected radial directions. The applied pressure is measured at the same time. The pressuremeter (Fig.18) is a cylindrical device where a flexible rubbcrmembraneis also used to apply uniform pressure to the walls of the pocket (borehole). The displacement of the pocket is determined by measuring the volume of fluid injected into the test cell The applied pressure is measured at the same time. The borehole jacking probe (Fig. 19) is a device where two half-shells made of steel are pressed diametrically against the pocket walls (borehole) by hydraulic pressure. The expansion between the half-shells is measured by displacement transducers. The applied pressure is measured at the same time. The types and operational possibilities for these prcssurcmctcrs arc summarised in Table 16. Recent experiences with some of the equipment arc published in [19, 98- 101 ]. Tahlc 17 gives an overview of some data for self-boring and full displacement pressuremeters. For ingtancc in [104-107] and f43, 44] comparahlc investigations with these two types of equipment are presented.
Fig. 17. Scheme of the dilatometer equipment (after DIN 4094-5)
98
Klaus-Jurgen Melzer and Ulf Bergdahl
Fig. 18. Scheme of the pressuremeter equipment (after DIN 4094-5)
Guide rod
\
Data acquisition
Cable_
___IB~ol
Pressure control Hydraulic pump
::__s
Sediment catcher
Borehole jacking probe
Fig. 19. Scheme of the borehole jacking
test equipment (after DIN 4994-5)
The special case of the flat dilatometer (DMT) - sec also ENV1997-3, 9 and [96, 97] - includes the dete rmination of the ground strata (supplemented by key boreholes), in situ stress conditions, the shear strength and deformation properties of cohesive soils and sand with a blade-shaped probe (Fig. 20). The fla t dilatometer has a thin circular steel membrane mounted on the outside of one side of the blade. The test is especially suitable for use in soils where pa rticles are small compared to the size of the membrane (e. g. days, silts, sands).
Table 16. Types and suilC:1bility of dilatometer, pressuremeter and borehole jacking tests (after DIN 4094-5) Nu.
I
I
2
3
4
T"'t designation (international)
Codo
Device for borehole expansion
Type of loading
Rock dilatometer test
ROT
One test cell, membrane
Radialsymmetric
5 l't-lea~urement
of
borehole expansion \Vith electric displacement transducers at tl1e pocket wall
6
7
8
Test result
Recommended borehole diaml-1cr
Recommended use at a modulus of compressibility Roed•l't-IP(I
mm Dilatorneter modulus
86
I
g Limited use in
Eo
Rock with strong discontinuities
Creep index
soft soils
101
Up to 50000
I
eod
ko Soil dilatometer test
2
SOT
One test cell, membrane
Radials~mmctric
\Vith electric displacement transducers inside lhe membrane
73
Up to 50000
Firm rock without discontinuities
66
10 to 1500
Firm rock without discontinuities
5 to 1 000
Firm rock without discontinuities
101
Eo Creep index
ko
I
rv1e11ard pressuremeter test
3
Dilatometer modulus
MPT
One test cell, two guard cells (top and boltom) membranes
Radial-
Volumo;;:trically
s~metric
Menard modulus
EM Limit pressure of the soil. PLM
Rorel1olc jacking test
4
B.JT
One test cell, two cylindrical 11~ lf shells
Diame1rical
\Vith electric displaco;;:mo;;:nt
Borehole
IOI
jackin~
146
transducer~
modulus
En Creep index I
kn
r-'
Table 17. Overview of some commercially available self-boring and pushed-in pressuremcters (following [87]) Group, international code
lnterrwtionul designation
Rc£crcnce
Self-boring, SBP
Cambridge self-boring pressure meter (CSBP)
Pushed-in. t'DP
0 0
Lenjph/diameter
Use in
84
6
All soils containing little or no gravel
Volumetrically
132
2
AU soils o:.:mtaioing little or no gravel
One test cell, membrane
Electric transducers three radial directions
73
5.5
Hard clays.. H~ry dense sands and weak rocks
[103]
One test cell, membrane
Electric transducers three radial directions
44
10
All soils except gravel
Pence! Pressuremeter
[87J
One test cell. membrane
Volumetrically
31
4
AJl soils except gravel
Fugro McClelland cone pressuremeter
[12]
One test cell, membrane
Electric transducers three radial directions
44
10.3
All soils except gravel
APEAGO minipressiometre
[87]
One test cell. membrane
Volumetrically
32
10
All soils except gravel
Device for pocket expalL'don
Measurement of hon!hole displacement
Diameter
[94J
One test cell membrane
Electric transducers, three radial directions
PAF76
[93J
One test cell, reinforced membrane
Weak rock self-boriog pressuremeter (RSBP)
[102]
Cambridge cone pressure meter
mm
IOI
1.3 Geotechnical field investigations
Control and Calibration unit Pressure source Pneumaticetectric cable Push rads
Pressure tube - Ground cable
Blade
141016 mm 94to96mm
:~
I E
,g
' I
1.1mm
l!lg. 20. Sche1ne of the DMT equip1nent and the measuring principle (after ENV 1997-3)
The hasic the test method is as follows. The hlade is pushed vertically into the soil by a thrust nlachine {for exan1ple, as used in cone penetration tests). At the selected test depth, the contact pressure po is initially measured while the memhrane is just about to lift off the blade. Subsequently. the pressure Pl (applied by gas) which is necessary to deform the memhrane for 1:I0 mm is measured. Fig. 20 shows the equipment and the measuring principle. Recent exan1ples of investigations are published in l108-112J. The para1neters obtained frotn these lateral pressure tests, such as the 1nodulus of elasticity EM from the MPT (Tahle 16 and Section 4.2), are not real geotechnical parameters but equip1nent specific para1neters. Therefore, it should be noted that the tests have to be performed and evaluated exactly in accordance with the standard procedures for each test in order to obtain reproducible and reliable results l87J. In addition, it is itnportant to gain local experience with the test to he a hie to use the results for design purposes. Depending on the particular equipment, the main steps of the test procedures are: l. Calibration hefore the test (pressure or measuring system, volume or displacement transducer syste1n, system compliance, correction factor for memhrane stiffness).
102
Klaus-Jtirgen Melzer and L'lf Dergdahl
2. Preparation of the test pocket (pre-drilling, special drilling of the pocket or push-in) and insertion of the probe minimising the disturbances of the pocket walls. 3. Performance of the test and the corresponding data acquisition, pressure application in constant load steps (stress control) or the creation of stages of constant pocket deformation (strain control), initial load and unload-reload cycles. 4. Recording of the test results (raw data). 5. Evaluation and correction of the measured values (hydrostatic pressure, membrane stiffness, system compliance, pore water pressure). 6. Reporting (number of the borehole and the test, equipment and component types used, borehole log etc., see also ENV 1997-3, 4.6). 7. Calibration after each test series (see step 1 ).
4.2
Evaluation
4.2.1 General 1be determination of equipment specific parameters from each test is far more complex than in the case of all other field tests treated in this Chapter. For example, the evaluation of the test data of a pressuremeter test (MPT) are summarised below (see also ENV 19973, 4 and DIN 4094-5).
The Menard modulus of elasticity Em and the limit pressure PLM are determined from the corrected test results according to Fig. 21. The diagram shows the injected fluid volume V versus the applied pressure p (upper part of Fig. 21) and Ll. V / ll.p versus p (lower part of
•E :g 0
E1o1~2,00~·Y
4V1- v,
(Y-Yc+Yr)
IP u.i
Pd 4p
1<
l
Applied pressure
I I
I I
I I
I I I
--'----
I
I I -----1,2m'1
'
I
,, Applied pressure
Fig. 21. Deternlination of EM and PLM from the results of a pressuTemeter test (afte• DIN 4094-5)
103
1.3 Geotechnical field investigations
Table 18. Examples of the determination of the modulus of elasticity Eoed from different lateral pressure tests Test
RDT
MPT
BJT
Modulus of elasticity
Test results
Explanations in
Eoed
Eq.no.
Eou(l - v) oed - (l + \•)(l 2v)
(18)
Modulus of elasticity: EM= 2.66V. lip/L>V Limit pressure: PLM from Fig. 21 or Pu.1=1.7pr-0.7oHs
Eoed =EM/Cl.
(19)
ENV 1997-3; DIN 4049-5
Modulus of elasticity: Es= f · d ·Lip/Lid Creep index: ko = (d2 - di)jlog(t2/ti)
EsuO - v) E oe ' -- (1 +v)(l-2v)
(20)
DIN 4094-.5
Eoed = RmEDMT
(21)
ENV 1997-3
Modulus of elasticity: Eo = (1 + v) · d/ Lid· i>p, Creep index: ko = (d2 d1)/ log(t2/t1)
Fig. 21 ). The limit pressure PLf\.f is defined as the pressure required to double the volume of the test cell and corresponds to the injected fluid volume ofV =Ve+ Yr where Ve is the deflated volume of the prohc and Vr is the injected volume measured at Pr, the latter being the pressure where .6. V / .6.p is a minimum. Table 18 contains the essential test results fron1 son1e types of equipn1ent. For further details reference should be made to the relevant standards. Due to the wide variety of the types of equipment, which enables the user to test soils and rock, and due to the almost 50 years of experience in this area, a numher of parameters can nowadays be derived from the test results which represent certain soil properties (see Tahlc 1.1 in [12], Table 10 in [87]). Among others, these are; the consolidation ratio, relative density, soil liquefaction, horizontal stress conditions, stress-strain relations, pore water pressure and permeability. The results of lateral pressure tests at international level arc applied to the gcotechnical design for spread and pile foundations. Series of large-scale tests and rigorous standardisation, e.g. [98] and Fascicule 62, together with detailed laboratory investigations, i.e. [113], have contrihuted to this fact. The applications of the test results in the design of sheet pile walls, in slope stability analyses and in tunnel design are also well known [87J. Principally, one has to distinguish, as in the case of penetration test results, hetween two methods of application [87 J: on the one hand en1pirical or theoretical derivation of geotechnical paran1eters serving as input to design methods or on the other hand empirical or semi-empirical methods, where the test results serve directly as input into the design methods. However, it has to be noted that for of all empirical and semi-empirical methods local experience plays a decisive role.
104
Klaus-Jtirgen Melzer and lJlf Bergdahl
4.2.2
Derivation of geotechnical parameters
Shear strength For the determination of the angle of effective shearing resistance q/ of sands using these methods, the approach is generally to develop a model for the behaviour of the sample soil and adjust it according to the test results. q/ then can be derived from this semiempirical model. Examples for empirical and semi-empirical methods are given in [44, 87, 113, 114]. Only results from SBP tests are generally used for this approach and not many examples are available today [87]. In addition, the SBP methods strongly depend on the local conditions for which they were specifically developed. The undrained shear strength Cu in cohesive soils can be determined directly from the upper part of the pressure-deformation diagram of an SBP test. However, empirical and semi-empirical methods arc used in the case of results from PBP tests. For instance, the limit pressure PLM is correlated with Cu from laboratory tests or in situ field vane tests [87, 99, 101]. Eq. (22) from ref. [99], where further relations are given, is an example for deriving Cu from MPT results in clays: (22) where: limit pressure according to Fig. 21 = horizontal stress at the tested depth
PLM =
uh
Eq. (23) is an example of the derivation of Cu from DMT results according to ENV1997-7, 9: Cu=
0.22o~ 0 (0.5 KnMT)l. 25
(23)
\\/here: u~ 0 = average normal stress at the tested depth before insertion of the probe KoMT = horizontal stress index (Table 18) Pressuremeter tests are suitable for the determination of the shear modulus for soils and also for rocks during initial and cyclic loading [43, 44, 100, 101, 106]. According to the relevant evaluations procedures (e.g. NF P94-110, ASTMD 4719), the shear modulus for the initial loading condition is determined from the middle, almost linear elastic part of the curve of the test results (Fig. 21 ): (24) where: V 0 = volume of the test cell before loading V m = average value of the volume in the almost linear elastic part of the test curve A modified evaluation method is suggested in [JOO] to make the results from MPT and SBP tests compatible. Compressibility In Table 18, equations for deriving the modulus Eoed from the results of tests arc summarised (Eqs. 18-21). Eqs. (19) and (21) for MPT and DMT are empirical relations. The values for a in Eq. (19) and Rm in Eq. (21) arc given in ENV 1997-3, 4.
1.3 Geotechnical field investigations
105
While deriving the modulus Eued from RDT and BJT results (Table 18), the following should he noted (sec DIN 4094-5, Fig. D.1 ). Experience shows that the moduli for unloading conditions Enu and EBu, respectively, deter1nined as the secant modulus from the middle section of the unloading curve, arc close to the Young's modulus Em of the material being investigated. The middle section is defined as the part of the pressure-defor1nation curve from 30 to 70 'Yo of the pressure hctwccn the upper turning point of the C)'Clc and the full unloading pressure (representing O0/o ). Assuming that rock and soil exhibit linear elastic, homogeneous and isotropic behaviour, Eoed can he derived from Eqs. (18) and (20), respectively. 4.2.3 Bearing capacity of spread foundations and piles Spread foundations The direct application of MPT results in the calculation of the bearing resistance of spread foundations is an excellent example of how a semi-empirical calculation method can be systematically converted into a standard method, see Fascicule 62 and (98]. For instance, the hearing resistance under vertical loads can he determined in accordance with ENV 1997-3, Annex C.1 by the following equation: R/ A' = CTvo
+ k(PLM -
Po)
(25)
where: R = resistance of the foundation to vertical loads A' = effective base area avu = total initial vertical stress at the level of the foundation base PLM = representative value of the Menard li1nit pressures beneath the foundation base Pu = Ku(av - u) + u; with Ku normally equal to 0.5, av as the total vertical stress at test level and u as the pore pressure at the same depth k = bearing resistance factor depending on soil type and PLM; given in ENV 1997-3, Table C.1 as a function of B, Land De B = width of the foundation L = length of the foundation De = equivalent depth of the foundation
ENV 1997-3, Annex C.2 also gives an example for MPT results serving as input to a method for calculating settlements, which is of special importance for spread foundation design. Pi1es The bearing resistance Q of piles can also be determined based on MPT results as follows (ENV 1997-3, Annex C.3): Q =A· k(PLM - Po)+ P:E(q,; · z;)
(26)
where: = base area of the pile equal to the actual area for closed ended piles and part of A that area for open ended piles PLM = representative value of the li1nit pressure at the hasc of the pile corrected for any weak layers below
106
Klaus-Jiirgen Melzer and Ulf Bergdahl
K 0 (ov - u) + u; \Vith K 0 normally equal to 0.5, ov as the total vertical stress at the test level and u as the pore pressure at the same depth k = bearing resistance factor depending on soil type, PLM and pile type; given in ENV 1997-3, Table C.4 P = pile perimeter qsi unit shaft resistance for the soil layer i, given by Fig. C.1 and l'able C.5; for both see ENV 1997-3, Annex C.3 Zi thickness of soil layer i p0
There are also methods available for estimating the settlement of pile foundations [87, 115]. Furthermore, it is important to note that a series of well-tried methods is available for determining the horizontal resistance of piles [87, 116]. 4.2.4
Comparison with the results from other field tests
If the results from lateral pressure tests in accordance with Tables 16 and 17 (including DMl') are used in conventional design, it has to be shown that the geotechnical parameters derived from these results correspond to those parameters used in traditional design methods. This has led to series of investigations to compare geotechnical parameters from lateral pressure tests with those determined from common laboratory tests (e.g. triaxial tests) and from other field tests (e.g. DP. SPT, CPT). Examples are given in [43, 99-109, 111].
5 Determination of density 5. 1 Sampling methods Field tests for determining the density are important, especially in cohesionless soils, because it is not possible to obtain undisturbed samples from boreholes (see Section 2.4). In Germany, the required tests are standardised in DIN 18125-2. Essentially, all tests follow the same principle: a defined volume of soil is measured in situ and its mass \Veight is determined. From this the density is given by: p=m/V
(27)
where: m = mass weight of the sample (moist or dry) V = volume of the sample Whilst the determination of the mass by weighing is relatively simple, the selection of the method for determining the volume depends on the soil type encountered. For instance, recovering undisturbed samples is possible with sampler tubes from trial pits and the base of excavations, roads, foundations etc. if the soil does not contain gravel, i.e. particles a diameter larger than 2 mm. In this case, the replacement methods should be used, i.e. the cavity produced by the sampling procedure is filled with a standardised replacement material in a standardised \Vay. The volume of the cavity is then determined by the volume of the replacement material necessary to fill the cavity. The different tests are defined by the means of determining the volume of the cavity. l'able 19 contains an overview of the different methods.
107
L.3 Geotechnical field investigations Table 19. Designation and suitability of tests for volume determination (following DIN 18125-2)
Code
Method
Designation of test
Applicable in
alter DIN 18125-2
Cohesive soils soils
Cohesionless
Stones and boulders!)
A
Cutting cylinder
DIN 18125-2-F-A
Without coarse grain
Fine to inediun1 sands
-
B
Balloon
DIN 18125-2-F-B
All
Fine to medium sands. gravel-sand mixtures, gravel with little sand
-
F
Replacement by fluid
DIN 18125-2-F-F
All
Fine to nlediu111 sands. gravel-sand mixtures, gravel \Vith little sand
-
G
Replacement by gypsu1n
DIN 18125-2-F-G
All
Fine to medium sands, gravel-sand nlixtures. gravel with little sand
-
s
Replacen1ent by sand
DIN 18125-2-F-S
All
Fine to medium sands, gravel-sand niixtures
-
Sch
Trial pit
DIN 18125-2-F-Sch
All
Fine to medium sands, gravel-sand mixtures
-
-
-
-
l
-
All -
I)
With little admixtures.
In cases where soils have to be investigated in depths that cannot be reached by the above close to surface methods. the density could be determined by radiometric methods (see Section 52), by dynamic probing (including SPT) or by cone penetration tests (see Sections 3,2 to 3.4),
5.2 Radiometric methods In radiometric methods, the radiation of radioactive isotopes is measured by Geiger counters and the results are correlated to the density and the moisture content of the soil (it was for this reason that the method was formerly called the "isotope penetrometer test''). Two types of radiation methods arc used: 1. Gamma radiation (y radiation), consisting of electromagnetic waves of high energy or gamma particles (y-y penetrometer), 2. Neutron radiation (n radiation), consisting of electrical neutral particles with the mass number 1 (neutron pcnctromctcr). The equipment consists of a radiation source, a detector for measuring the radiation intensity and an impulse counter. The combination of the radiation source and the detector is called a radiometric probe. Two main types of equipment arc used: devices for closcto-surface operations, e.g. compaction control ("close-to-surface probes"), and probes
108
Klaus-JOrgen Melzer and 1 Jlf Bergdahl
Dimensions in mm
1-
...- Geiger-MUiier counter
I
~
:;:
! '
, I
Lead shield Bortrifluoride -, ~
counter
: ~1. ., I ~I -Oir~-t---+--+-- Ra source L ~
~Jdh-~ 1
jl· --
Ra-Be source-..______
i
l
N
~
Fj-~~-~
:;:
=-t
;:!
1I
D
Fig. 22. Example of a y-y pt:nt:lromeler (left hand sidt:) and a nt:ulron penetrometer (right hand side) without radiation protection (aftt:r [71)
used for deep investigations of the ground ("depth-probes"). Fig. 22 shows the arrangement of the components for a y-y penetrometer and of a neutron penetrometer as depthprobes.
The use of radiometric methods is subjected to legal regulations and legal permission has to be obtained. Regulations for radiation protection control, transport, storage and calibration of the radiometric probes apparently still restrict their use. DIN 18125-2 refers to [117] where the methods are described in detail (definitions, terminology. equipment, calibration, performance of measurements, radiation protection etc.). Radiometric devices for use at greater depths are sometimes already integrated within the cones of cone penetration penetrometers [12, p. 186ff., 51-53, 118]. For the evaluation. the density p, the water content wand the dry density Pd are plotted against depth (Fig. 23). Their application for the compaction control of fills is well established. The combination with key borings and for example, cone penetration tests [119J results in valuable information about the ground strata of natural soils and their properties. This type of application
109
1.3 Geotechnical field investigations
p28
Measuring point P2s
Water contentw o o1 02 03 04 o~, 0 0102030405 2. ~· m'\.o ~ Density Q lO 15 -~ and Qd
1.5 ro(t'm3) I
~ ~
;;=
= ~~
1.00
~
• Gravelly sand
w
8
,ll
w -Q I
-~
r '
2.00 I" E
1
!
I
'
~
GW
Q
w IQ
Q,
E
~ c
~
~-
I
,, I
-
Fig. 23. Results of gamma radiation and neutron measurements for determining the natural density p, the ,,,,.ater content wand the dry density Pd
has gained in1portance recently by con1bining the radiometric device with CP'fU equipment. In general, radiometric nlethods arc suitahlc for use in cohcsionlcss soils [54, 118]. Examples of their use in clayey soils are reported in [53].
6 Geophysical methods 6.1
General
Geophysical investigation methods can he used in conjunction with key horeholcs • in prelin1inary investigations of large-scale projects for detern1ining the stratification of the lop layers; • in design investigations, to complement the geotechnical investigation; • for locating geological joints, discontinuities and anomalies in the strata; • for locating historical or unknown objects and cavities in the ground; • for locating seepage and gradients in the groundwater ftow; • for determining geophysical parameters; • for controlling contaminated groundwater fronts, salt water fronts etc. Tahles 6 and 7 of Supplement 1 of DIN 4020 give overviews of surface and horehole nlethods and their characteristics. The interpretation of the test results requires experience and special knowledge. In numerous practical applications it has proven appropriate to combine different independent methods to avoid misinterpretations [120-123]; the cost remains justifiahle because the requiren1ents for equipment and personnel are relatively small for most of the available methods. In Germany, common geophysical nlethods were investigated on a scientific basis within the framework of an extensive research programme "Methods for the ground investigation and description of landfills and toxic waste deposits". The results were published in [124]. A theoretical study is also available from Finland [125].
110
Klaus-Jtirgen Melzer and Ulf Bergdahl
"Geophysics" on the Internet
• Deutsche Geophysikalische Gesellschaft (DGG): http://www-seismo-hannover.bgr.de/dgg/dgg.html • DMT-Gesellschaft filr Forschung und Prtifung, Essen: http://www.fp.dmt.de • llarbour Dom, K()ln: http://www.harbourdom.de • GeoPager: http://\VW\v.geopager.de
• Deutsche WWW-Server-Liste der Geologie Clausthal: http://w\vw.inggeo.tu-clausthal.de/geo-server/geoserver-germany.html
6.2
Brief descriptions of some methods
• Soil dynamics and soil seismic testing: see Chapter 1.8 and [12, p. 179. 126] • Gravimetric methods: the measurement of anomalous deviations, unit is 1ngal, to explore underground close-to-surface hidden objects or cavities in sufficiently level terrain. The application of gravimetrics should always be combined with other geo-
physical methods. • Radiometric methods: see Section 5.2. • Geo-electric methods: by pushing two electrodes into the ground at a set distance a, a direct current field is created in the soil. Using this field, the specific electric resistivity in [nm] is determined from the potential difference in a soil mass reaching to a depth of about a/4. Approxitnate values for the resistivity are given in [127]: rock. solid: > 5000 >lm,rock, weathered: 100-1000 >lm, sand, moist:> 100 >lm,sand, wet:> 50 Qm, silt, moist: > 20 Qin, fresh water: 20 Qm. An extension of this 1nethod consists of progressively increasing the distance of the electrodes outside the nleasuring probes. • Geo-radar: Uses a transmitter/receiver antenna (transducer) pulled over the ground surface inducing electromagnetic impulses into the ground. The signals reflected from, for exa1nple, the strata boundaries in the ground, are registered. 1l1e wave propagation depends primarily on the dielectric properties and conductivity of the ground. At discontinuities, the signal is spread, reflected, inflected and partially absorbed. Approximate values for penetration depths in soils: up to 10 m [127]. • Geomagnetic methods: The nleasure1nent of anomalies in the ground's magnetic field, caused by a ferromagnetic rock mass or other objects (for example, unexploded bombs, cables etc.). With 1nagneton1eters (sensors set at t\VO different heights above the ground; Fdrster probe or proton-magnetometer) the intensity and gradient in a vertical plane are measured. A depth of only about 4 nl can be reached because 1neasurable values caused by objects decrease with the third power of the depth. • Electromagnetic methods (TEM): lJsing a mobile probe, an artificial (reacting to all metals) electromagnetic field is created. After turning off the transmitted current, the voltage induced into a receiver spool is registered. The method is characterised by high measuring speed and insensitivity against technical disturbances. Obtainable measuring depths are similar to those for geomagnetic 1nethods.
1.3 Geotechnical field investigations
111
• Soil thermionics: The measurement of temperature anomalies below a depth of 1.5 m with temperature sensors (at depth increments of about 1 m) placed in driven hollow rods. Measuring accuracy is to± 0.1'"' [128. 129]. The primary application of this method is the location of leakages in the ground.
7
References
[1 J DIN Deutsch es Institut ftir Nor munge. V.: Bauen in Europa-Felduntersuchungen und Laborversuche fi.ir die geotechnische Bemessung. Beuth Verlag GmbH, Berlin/Wien/Zilrich 200 l. [2] Stiilben, F., Eitner, V.: Wesentliche Anforderungen bei der Probenentnahme in Boden und Fels nach Eurocode 7. bbr 11(1999),30-·33. [3] Schultze, E., Afuh.1; fl.: Bodenuntersuchungen ftir Jngenieurbauten, 2nd. Edition, SpringerVerlag, Berlin/Heidelberg/New York 1967. [4] S\vedish Geotechnical Society: Geotesnisk Fti.lthandbok. SGF Rapport 1:96. Velent AB, Stockholm 1996. [SJ Bakke, T., Braathen, O.A., Eilerrsen, 0., Myklebust,!.: Quality assurance of fieldwork. Nordiska Ministerradet, TemaNord 1997:590. Kopenhagen 1997. [6] Kahl, H., Muhs, H., .Meyer, l-V: Ermittlung der GrOBe und des Verlaufs des Spitzendrucks bei Drucksondierungen in ungleichfbrmigem Sand, in Sand-Kie:;,-Gemischen und im Kies. Mitteilungen der Degebo, No. 2 l.1968. [7J Melzer, K.-J.: Sondenuntersuchungen in Sand. Mitt. Inst. f. Verkehrswasserbau, Grundbau und Bodenmechanik an der TH Aachen, No. 43, 1968. [8J Afelzer, K.-J.: Measuring soil properties in mobility research: relative density and cone penetration resistance. Technical Report No. 3-652, Report 4, U.S.A.E. Water\vays Experiment Station, Vicksburg 1971. [9J Afelzer, K.-J.: Relative density -Three examples fro111 research and practice. Special Technical Publication 523, American Society for Testing and Materials, Philadelphia 1973, 463---477. [10] Muhs, fl.: 50 years of deep sounding with static penetrometers. Tn: A half century in geotechnics. Jubilaumsheft zu Ehren von Professor A. Hamdi Peynircioglu, Techn. lJniv. Istanbul, 1978, 40-54. [llj Clayton, C.R.!.: The Standard Penetration Test (SPT): methods and use. Construction Industry Research Jn formation Association, Report 143, London l 995. [12] Lunne, T, Roberrson, PK., Pot-t'ell, J.J.M.: Cone penetration testing in geotechnical practice. E & FN Spon/Routledge, London, Ne\\' York 1997. [13] Robertson, P.K., Afuyne, P.l-V: Proc. 1st TC on Site Characterization, Atlanta, A.A. Balkema, Rotterdam, Brookfield 1998. [14] Int. Soc. Soil Mech. Found. Eng.: Report of the Technical Committee on Penetration Testing of Soils - TC l6 with Reference Test Procedures CPT-SPT-DP-WST. Swedish Geotech. Inst., Information 7, LinkOping 1989. [l5] Int. Soc. Soil Mech. Geot. Eng.: International Reference Test Procedures for Cone Penetration Test (CPT) and Cone Penetration Test with Pore Pressure (CITU). Report of the Technical Committee on Ground Characteri:;,ation from In Situ Testing-TC 16. Proc. 12th Europ. CSMGE, Amsterdam 1999 (corrected 2001), Vol. 1, 2196--2222. [16] Krtirner, H.-J: Untersuchung der bearbeitungstechnischen Bodenkenn\verte mit schwerem Ramm-Druck-Sondier-Gerat zur Beurteilung des Maschineneinsatzes im Erdbau. VerOff. Inst. r. Maschinenwesen im Baubetrieb der Universitat Karlsruhe, Reihe FI No. 14. 1976. [17] Kriirner, H.-J: Gertitetechnische EinftuBparameter bei Ramm- und Drucksondierungen und ihre Auswirkungen auf den Eindringwiderstand. VerOff. Inst. f. Maschinenwesen im Baubetrieb der Universitiit Karlsruhe, Reihe FI No. 26, 1981.
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Klaus-Jiirgen Melzer and Ulf Bergdahl
[18] Koester, J.P., Daniel, C., Anderson, M.: In situ investigation of liquified gravels at Seward, Alaska. Proc. Conf. on Innovations and Applications in Geotechnical Site Characterizations, Geo Denver 2000. ASCE Spec. Publication No. 97, Denver 2000, 33-48. [19] Biedermann, B ..· Vergleichende Untersuchungen mit Sonden in Schluff. Forschungsberichte aus Bodenmechanik und Grundbau, published by Prof. Dr.-lng. E. Schultze, Heft 9, Aachen 1984. [20] Rollberg, D.: Bestimmung der Tragfiihigkeit und des Rammwiderstands von Pfiihlen und Sondierungen. Forschungsberichte aus Bodenrnechanik und Grundbau, published by Prof. Dr.-Tng. E. Schultze, No. 3, Aachen 1977. [21] Magnusson, 0., Anderson, H., Astedt, B., Holrn, G'.: The drivability of friction piles based on penetration testing. Proc. 14th ICSMCIE, Hamburg 1997, Vol. l, 539--542. r22] Schurnacher, L.: Spitzendruck bei Sondierungen, EinfiuBgrOBen und Prognose. Bautechnik 76 (1999), 568--580. r23J Decourt, L.: A more rational utilization of some old in situ tests. Proc. 1st IC on Site Characterization, Atlanta, 1998, Vol. 2, 913-918. [24] Butler, J.J., Caliendo, JA., Goble, G.G.: Comparison ofSPTenergy measurements methods. Proc. 1st IC on Site Characterization, Atlanta 1998, Vol. 2, 901-905. [25] Farrar, I.A.: Su1nmary of Standard Penetration Test (SPT) energy ineasure1nents experience. Proc. 1st IC on Site Characterization, Atlanta l 998, Vol. 2, 919-926. [26] Abou.-Matar, H., Goble, G.G.: SPT dynamic analysis and measurements, ASCE,Journ. Geotechnical and Geoenvironmental Engineering l 23 ( l 997), 92 l-928. [27] Lutenegger, A.I., Kelley, SP.: Standard Penetration Tests \Vith torque ineasure1nents. Proc. 1st IC on Site Characterization, Atlanta 1998, Vol. 2, 939-945. [28J Menzenbach, E.: Die Anwendbarkeit von Sonden zur Prlifung der Festigkeitseigenschaften des Baugrundes. Forschungsberichte des Landes Nord-Rhein-Westfalen, No. 713, Westdeutscher Verlag, Koln 19S9. r29] Schultze, E.: Diskussionsbeitrag. 5th ICSMFE, Paris 1961, Vol. 3, 183-184. [30] Kiekbusch, M., Siebenkorn, G'.: Der Standard Penetration Test (SPT). bbr 8 (1999), 2-7. l31] Barros, JM. C., Pinto, C.S.: Estin1ation of maximum shear modulus of Brazilien tropical soils from Standard Penetration Test. Proc. 14th ICSMGE, Hamburg 1997, Vol. l, 29-30. [32] Kokusho, T: Formulation of SPT N-value for gravelly soils with different particle gradings. Proc. 14th ICSMGE, Hamburg 1997, Vol. I, 523-526. [33] Sanglerat, G.: The penetrometer and soil exploration. Interpretation of penetration diagrams - theory and practice. Elsevier Publication Company, Amsterdam/London/New York 1972. [34] Schultze, E., Melzer, K.-J: 1l1e determination of the density and the modulus of compressibility of non-cohesive soils by sounding. Proc. 6th ICSMFE, Montreal 1965, Vol. 1, 354-358. [35] Teferra, A.: Beziehungen zwischen Reibungswinkel, Lagerungsdichte und Sondierwiderstanden nichtbindiger Boden mit verschiedener Kornverteilung. Forschungsberichte aus Bodenmechanik und Grundbau; published by Prof. Dr.-Ing. E. Schultze, No. 1, Aachen 1975. (36] Coutinho, R. Q., Oliveira, J. TR.: Geotechnical characterization of a Recife soft clay - Labororatory and in-situ tests. Proc. 14th ICSMGE, Hamburg 1997, Vol. 1, 69-72. (37] Danziger, F..A.B., Politano, C.F.., Danziger, B.: CPT-SPT correlations for so1ne Brazilian residual soils. Proc. 1st IC on Site Characterization, Atlanta 1998, Vol. 2, 907-622. [38] Vianna da Fonseca, A., Fernades, M.M., Cardoso, A.S.: Correlations between SPT, CPT, and Croos-Hole testing results over granite residual soil of Porto. Proc. 14th TCSMGE, Hamburg 1997, Vol. I, 619-622. [39] Mitchell, J.K., Brandon, TL.: Analysis and use of CPT in earthquake and environmental enginering. Proc. 1st IC on Site Characterization, Atlanta 1998, Vol. 2, 69-97. l40J Susuki, Y, Sanetnatsu, T, Tokimatsu, K.: Correlation between SPf and seismic CPT. Proc. 1st IC on Site Characterization, Atlanta 1998, Vol. 2, l 375-1380. [41] Peuchen. J: Comercial CPT profiling in soft rocks and hard soils. Proc. lst lC on Site Characterization, Atlanta 199~, Vol. 2, 1131-1137.
1.3 Geotechnical field investigations
113
[42] Weiss, K.: Die Haupthodenarten in Berlin als Baugrund. Vortriige Baugrundtagung Berlin .521. Deutsche Gesellschaft fiir Erd- und Grundbau, Essen 1978. [43] Post, M.L., Srnits, M. TJ.H., Kolk, H.J.: Comparison of cone pressuremeter data with results from other in-situ and laboratory tests. Proc. 14th ICSMGE, Hamburg 1997, Vol. 1, 569-572. r44J Powell, J.J.M., Shields, C'.H.: 1be cone pressuremeter-A study of its interpretation in Holmen sand. Proc. 14th ICSMGE, Hamburg l 997, Vol. l. .573-.57.5. r45J Mayne, PW, Robertson, PK., Lunne, T: Clay history evaluated from seismic piezocone tests. Proc. 1st IC on Site Characterization, Atlanta 1998, Vol. 2, 1113-1118. [46] lloulshy, G. T, Ruck, B.M.: Interpretation of signals from an acoustic cone penetrometer. Proc. 1st IC on Sile Characterization, Atlanta 1998, Vol. 2, 1075-1080. [47] Men~e, P.: Acoustic emissions cone penetration testing (AE-CPT). Proc. lst IC on Site Characterization, Atlanta 1998, Vol. 2, 1119-1124. [481 Burns, SE., Mayne, PW: Penetrometers for soil permeability and chemical detection. Georgia Institute of Technology, School of Civil and Environmental Engineering. Report GIT-CEEGE098-1. Atlanta 1998. [491 C'ampanella, R.G., Davies, M.P: In-situ testing for geo-environmental site characterization: A mine tailing example. Proc. 14th ICSMGE, Hamburg 1997, Vol. 1, 43-46. [501 Campaneffa R.G., Kristiansen, H., [Janie!, C., [Javies, M.P: Site characterization of soil deposits using recent advances in piezocone technology. Proc. !st IC on Site Characterization. Atlanta 1998. Vol. 2, 995-1000. [.51] SRE: Radio-isotope cone penetrometers, Brochure, Soil and Rock Engineering Co., Ltd., ()saka 1999. [521 Mirnura, M., Shrivastava, A.K.: Application of RI-cone penetrometers in sandy foundations. Proc.14thICSMGE,Hamhurg l997, Vol. 1, 547-550. r53J Miniura, M., Shrivastana, A.K., Shibata, T, Nobuyama, M.: In-situ measurements of wet density and natural water content with RI-cone penetrometers. Proc. 5th Int. Syrup. Field Measurements, Singapore 1999, 559-564. r54J Raschke, SA., Hryciw, R.D.: Vision cone penetrometer for direct subsurface soil observation. ASCE, Journ. Geotechnical and Geoenvironmental Engineering 123 (1997), l074-l076. r55J HryciH·'. R.lJ., Raschke, SA.: In-situ soil characterization using vision cone penelromeler (VisCPT). Proc. 1st IC on Site Characterization, Atlanta 1998, Vol. 2, 1081-1086. r56] Begem.ann, H.K.S.: Improved method of determining resistance lo adhesion by sounding through a loose sleeve placed behind the cone. P,.oc. 3rd ICSMFE. Ziirich 1953, Vol. L 213-217. [.57] Be~emann, ll.K.S: The friction jacket cone as an aid in determining the soil profile. Proc. 6th ICSMFE, Montreal 1965 Vol. 1. 17-21. [581 Berry, K.M., Olson, SM., Larnie, M.: Cone penetration testing in the Mid-Mississippi River Valley. 1st IC on Site Characterization, Atlanta 1998, Vol. 2, 983-987. [59"] Hegazy, YA., Mayne, PW: Delineatinggeo.slratigraphy by cluster analysis of piezocone data.1st IC on Site Characterization, Atlanta 1998, Vol. 2, 1069-1074. f60J Zangh, Z., Tamay, M. T: Statistical lo fuzzy approach toward CPT soil classification. ASCE, Journ. Geotechnical and Geoenvironmental Engineering 125 (1999), 179-186. [61] Melzer, K.-J..· Uber Erfahrungen mit der Ilegemann-Spitze. Bauingenieur 43 (1968), 340342. [62] Rohertson. PK., Wride, C.E.: Evaluating cyclic liquifaction potential using cone penetration. Canadian Geolechnical Journal 35 (1998). 442-459. [63] Salgado, R., Boulanger, R. W., Mitchell, J.K.: Lateral stres.s effects on CPT liquifaction resistance corelations. ASCE. Journ. Geotechnical and Geoenvironmenlal Engineering 123 (1997), 726735. [641 Zangh, L.: Assessment of liquifaction potential using optimum seeking. ASCE, Journ. Geotechnical and Geoenvironmental Engineering 124 (1998), 739-748. [651 Konrad, 1M.: Sand state from cone penetrometer tests: a framework considering grain cruhing stress. Geotechnique 48 ( l 998), 201-21.5.
114
Klaus-Jtirgen Melzer and UlfBergdahl
[66J Salgado, R., Mitchell, JK., Jamiolkowski, M.: Calibration chamber size effects on penetration resistance in sand. ASCE, Journ. Geotechnical and Geoenvironmental Engineering 124 (1998), 878-888. [67 J Tanaka, fl., Tanaka, Af.: Characterization of sandy soils using CPT and DMT Soils and Foundations 38 (1998), 55--05. [68] Kahl, H., Muhs, H.: lJber die Untersuchung des Baugrundes nlit einer Spitzendrucksonde. Baulechnik 29 (1952), 81-88. [69] Muhs, H.: Die Priifung des Baugrundes und der BLlden, Handbuch der Werkstoffpri.ifung, 2nd Edition, Kapitel XXIll, p. 819-988. Springer-Verlag, Berlin-GOttingen-Heidelberg 1957. [70] Muhs, H.: Neue Erkenntnisse iiber die Tragfiihigkeil von ftachgegri.indelen Fundamenlen aus GroBversuchen und ihre Bedeutung fiir die Berechnung. Bautechnik 46 (1969), 181-191. [71] Bergdahl, U., Ottosson, E., Malmborg, B.S.: Plattgrundlagning. AB Svensk Byggtjanst, Stockholm 1993. [72J Meyerhoj; G.G.: General Report. Proc. European Symposium on Penetration Testing, Stockholm 1974, Vol. 2.1, 41-48. [73] Muhs, H., Weij3, K.: Untersuchung von Grenztragfiihigkeit und Setzungsverhalten ftachgegriindeter Einzelfundamente in ungleichfOnnigen nichtbindigem Boden. Mitteilungen der Degebo, No. 26, 1971. [74] Muhs, H.: On the relation ot' the bearing capacity factors, the modulus of elasticity and the cone resistance. Proc. Europ. Symp. on Penetration Testing, Stockholm 1974, Vol. 2-1, 141-142. [75] .Eidt, H. T., Stark, T.D.: Undrained shear strength front cone penetration test. Proc. !st IC on Site Characterization, Atlanta 1998, Vol. 2, 1021-1025. [76] Bandini, P., Salgado, R.: Methods of pile design based on CPT and SPT results. Proc. 1st IC on Site Characterization, Atlanta 1998, Vol. 2, 967-976. [77] Eslami, A., Felleniu.s, R.H.: Pile capacity estimated fron1 CPT data -Six methods compared. Proc. 14th ICSMGE, Hamburg 1997, Vol. 1, 91-94. [78J Grundlagen zur Festlegung von Sicherheitsanforderungen fiir bauliche Anlagen. Beuth Verlag 1981. [79] Zweck, H.: Baugrunduntersuchungen durch Sonden. Bauingenieur-Praxis, No. 71. Ernst & Sohn, Berlin/ Miinchen 1969. [80] Helenelund, K. V: Methods for reducing undrained shear strength of soft clay. s. . .·edish Geot. Inst., Report No. 3, LinkOping 1977. [81] Bjerrum, L.: En1bank1nents on sott ground. State-ot'-the-art Report. Proc. ASCE Cont'., Purdue, Indiana 1972, Vol. 2, 1-54. [82] Bjerrum, L.: Probletns of soil mechanics and construclion of soft clays. State-of-the-art Report. Proc. 8th ICSMFE, Moscow 1973, Vol. 3, 111-159. [83J Norwegian Geot. ln~l.: Veiledning for utfo1·else ov vingeborr; Melding No. 4, Oslo 1982, Rev. 1 1989. [84] Swedish Geotechnical Society: Reco1n1nended standard for lield shear test. SGF Report 2:93E, Velenl AB, Slockholm 1996. [85] Zllleskj, J., Kos, J., Salllk, J., Halama, V., Karlin, P, Horejs{, V.: Vane Lest used for very soft soil-like materials characterization. Proc. 1st IC on Site Characterization, Atlanta 1998, Vol. 2, 1201-1206. [86] Bergdahl, U., Bronts, B., Muromachi, T.: Weight sounding test (WST): International reference test procedure. Proc. 1st Int. Symposium on Penetration Testing, ISOPT-1, Orlando 1988, Vol.1, 71-90. [87] lnt. Soc. Soil Mech. Geot. Eng.: Pression1eter testing in onshore ground investigations. Report of the Technical Committee on Ground Characterisation from In Situ Testing - TC 16. Proc. 1st IC on Site Characterization, Atlanta 1998, Vol. 2, 1429-1468. [88] K6gler, r:: Baugrundpriifung int Bohrloch. Bauingenieur 14 (1933), 266-270. [89J Kogler, F, Scheiding, S.: Baugrund und Bauwerk. Ernst & Sohn, Berlin, 1938, 4th Edition 1944. (90] Menard, L.: Mesures in situ des propri6tes physiques des sols. Ann. des Ponts et Chausse.es 127 (1957), 357-377.
1.3 Geotechnical field investigations
115
[91 J Gibson, R.E., Anderso11, WF: In situ measurement of soil properties with the pressurementer. Civ. Engng. and Puhl. Works Review 56 (1961),615--618. [92] Goodman, R.E.: Measurement of rock deformability in boreholes. Proc. 10th Symposium Rock Mechanics, Austin 1968, University of Texas. [93] Baguelin, F, .Tezequel, J.F, Shields, D.11.: The pressuremeter and foundation engineering. Lab. Central des Pants et Chaussees, Paris, and R.M. Hardy & Ass., Kanada. 1978 [94] Wroth, C.I'., Hughes, .T.l\1.0.: An instrument for the in-situ measurement of the properties of soft clays. 8th ICSMFE, Moscow 1973. Proc., Vol. 1.2, 487-494. [95] Reid, W2\1., St. John, H.D., Fyffe, S., Ridgen, WI.: The push-in pressiometer. Proc. Symposium on the pressiometer and its marine application. Paris, 1982, 247-261. [96] 2\1archetti, 5.: In situ tests by flat dilatometer, ASCE Journal of Geotechnical Engineering, No. 106, (1980), 299-321. [97] Schnierffnann, J.H.: Suggested method for performing the flat dilatometertest. ASTM Subcommittee 18.02. ASTM Geotechnical Testing Journal 9 (1986) 99-101. [98] Amar, S., Bagu.elin, F, Canepa, Y., Frank, R.: New design rules for the bearing capacity of shallow foundations based on Menard pressuren1eter tests. Proc. 1st IC on Site Characterization, Atlanta 1998, Vol. 2, 727-733. [99) Bahar, R.: Properties of clays fron1 Menard pressuremeter test results. Proc. Ist TC on Site Characterization, Atlanta 1998, Vol. 2, 735-740. [100J Gamhin, M.I'., .rezequel, J.F: A new approach to the Menard PMT parameters. Proc. lst IC on Site Characterization, Atlanta 1998, Vol. 2, 777-782. [101] Hughe.\~ .T.l\1.0., Gertje, H., Wu, M..l.: Pressuremeter testing for drilling shafts in gravelly clays. Proc. 1st IC on Site Characterization, Atlanta 1998, Vol. 2, 789-793. [102] Clarke, B.G., Allan, PG.: Self-boring pressuremeter for testing weak rock. Proc. 12th ICSMFE, Rio de Janeiro 1989, Vol. 1,211-213. [103] Whiters, /\i'.J., Howie, I.A., Hughes, J.M.O., Robertson., PK.: The development of a full displacement pre.ssuremeter. Special Technical Publication 950, American Society for Testing and Materials, Philadelphia 1986, 38--56. [104] Cunha, R.P: Quantification of the soil disturbance generated by selfboring pressuremeters. Proc. 1st TC on Site Characterization, Atlanta 1998, Vol. 2, 753-758. [105] llughes, J.M.O., Ca1npanella, R.G., Debasis, R.: A simple understanding of the liquifaction potential of sands from self-boring pressuremeter tests. Proc. 14th TCSMGE, Hamburg 1997, Vol. I, 515-518. [106] Macklin, S.R., Yinisiri, S., Soga, K.: Assessment of the strength and stiffness of the London Clay Formation at Crown \Vharf, T.ondon. Proc. 5th Int. Symp. Field Measurements, Singapore l 999, 565-570. [107] Pinto, C.S., Ahra1nento, .~1.: Pressuremeter tests on gneissig residual soil in Sao Paulo, Brazil. Proc. 14th ICSMGE, Hamburg 1997, Vol. I, 175-176. [108] Mello Vieira, M. V C., Danziger, FA.B., Alfneida, 2\1.S.S., Lopes, PC.C.: Dilatometer tests at the Sarapui soft clay site. Proc. 14th TCSMGE, Hamburg 1997, Vol. 1, 161-162. [I 09] .~1ulabdic, M.: Comparison of piezocone, Marchetti dilatometer and vane test results for the Danube-Sava canal. Proc. 14th ICSMCJE, Hamburg 1997, Vol. 1, 561 563. [I IO] Redel, C., Blech1nann, D., Feferbau1n, S.: Flat dilatometer testing in Israel. Proc. 14th ICSMGE, Hamburg 1997, Vol. I, 581-584. [111] Tanaka, A., Bauer, G.E.: Dilatometer tests in a Leda clay crust. Proc. ]st TC on Site Characterization, Atlanta 1998, Vol. 2, 877-882. [112] Toran~ M., Calabrese, 2\1.: Jn situ determination of ch by flat dilatometer (DMT). Proc. 1st IC on Site Characterization, Atlanta 1998, Vol. 2, 883-888. [I 13] Biarez, J., Ganzbin, M., Gomes-Correia, A., Flavigny, E., Branque, D.: Using pressuremeter to obtain parameters to elastic-plastic models for sand. Ist IC on Site Characterization, Atlanta 1998, Vol. 2, 7 47-7.12.
116
Klaus-Jilrgen Melzer and lJlf Bergdahl
[ l 14] Fukugawa, R., .Muro, ·r, Hata, K., Hino, N: A new method to estimate the angle of internal friction using a pressuremeler lest. lsl IC on Sile Characterization, Atlanta 1998, Vol. 2. 771-775. [ l 15] Fujiyasu, Y, Orihura, K.: Elastic modulus of weathered rock of Jurong Formation in Singapore. Proc. 5th Int. Symp. Field Measuren1ents. Singapore 1999. 183-186. [116] Seeger, K.: Beitrag zur Ern1ittlung des horizontalen Bettungsmoduls von Boden
8
Standards
ASTM D 1586-84: Standard test method for penetration test and split barrel sampling of soils. American Society for Testing and Materials, Philadelphia 1992. ASTM D 4633-86: Standard test method for stress wave energy measurements for dynamic penetrometer testing systen1s. American Society for Testing and Materials, Philadelphia 1986. ASTM D 4719-94: Standard test nlethod filr pressuremeter testing in soils. American Society for Testing and Materials, Philadelphia 1994.
1.3 Geotechnical field investigations
117
BS 1377: Part 9: British standard methods of test for soils for civil engineering purposes. Part 9: In situ tests. British Standards Insitution, London 1990. DIN 1054: Baugrund - Sicherheitsnachweise im Erd- und Grundbau. Draft, 2000. IJINEN 1536: Ausfilhrung von besonderen geotechnischen Arbeiten (Spezialtietbau)-Bohrpf:.ihle. 2000. IJIN 4020: Geotechnische Untersuchungen fiir bautechnische Zwecke; einschl. Beiblatt 1: Anwendungshilfen, Erlauterungen, 1990 (New edition in preparation). IJIN 4021: Baugrund- AufschluO
1.4 Properties of soils and rocks and their laboratory determination Paul van Sous and ]an BuhdC
1
Soils and rocks - origins and basic terms
In the civil engineering context, soil means the sedimentary material of the upper part of the Earth's crust, which is relatively loose, not strongly cemented or highly co1npressed. Generally, soil can be worked 'Nithout drilling or blasting.
Rock, on the other hand, is the hard, rigid deposit with strongly cen1ented constituents. Its properties arc governed by the ccmcntation, and by any joint system (discontinuities),
along which the cohesive effect of cementation would be lost. Products due to the weathering of solid rocks that have not been transported are called residual soils. If transported by wind, water or ice, they for1n deposits (sedirnents). In organic soils, along with the 1nineral constituents. re1nnants of organic n1atter arc present. Solid rocks are classified according to their mode of formation into three main categories: igneous rocks (c. g. granite) derived from molten material, sedimentar_v rocks (e.g. sandstone) formed by sediments settling in water, and metamorphic rocks formed by the re-crystallization of igneous or sedimentary rocks due to high pressure and/or temperature. The soil and rock forming processes, expressed in the cycle of weathering - transport deposition and rock-forming. can be interrupted or renewed at any stage. Differences in the origins and histories of soils and rocks therefore result, which can explain the vast diversity and heterogeneity of their gcotcchnical properties. Soil and/or rock within the influence zone of structural loading is frequently called subsoil, despite different usage of the ter1n in the pedology. Besides this role of supporting structures, soil or rock often represent building materials when the structure is constructed of then1.
2 2.1
Properties of soils Soil layers
Soils deposited under constant conditions for1n practically continuous ho1nogeneous strata-layers. In engineering practice. their properties may be considered constant within individual layers. However, even in apparently ho1nogeneous layers the properties of soils change from point to point. The extent of the fluctuation depends on the origin and the nature of the property in question: 1noraines transported by ice are less ho1noge11eous than clays deposited in stationary water and, for example, the scatter in density of the solid particles in a given soil is very 1nuch sn1aller than the variability of its coefficient of
permeability.
120
Paul von Soos and Jan Bohat
Inhomogeneity of soils hecomes ohvious only after studying a larger numher of soil sarnples. This inhotnogeneity steins either fron.1 a random variation of soil properties, or from a systematic one which depends on direction, or from the comhination of the two. A good representation of the variation of soil properties can be obtained by sounding (see Chapter 1.3). Statistically satisfactory descriptions of the properties may require a large nurnber of samples. However, for economic reasons the interval of variation is often determined only for basic parameters that can he easily ohtained (water content, particle size distribution, Atterberg limits). Established correlations then allo\v us to limit the determination of more costly para1neters to statistically significant intervals. CJenerally, expertise is required for effective choice of samples. Moreover, the results should he critically examined to find whether sa1npling or testing could cause any systen1atic or randon1 errors. For a reliable assessment of the characteristic values for ground parameters used in design, all aspects of ground conditions should be thoroughly investigated (DIN 4020 [25, 47]).
2.2
Soil samples
In German practice, distinction is n1ade between "individual satnples" (eletnents cut frotn the soil, for example undisturbed sa1nples), samples taken, for example, from the face of an excavation and Iinally samples procured after the soil was excavated, for example from a dun1p. The samples to be tested in the laboratory must satisfy the following criteria: 1. The properties investigated must he changed as little as possible (see also Chapter 1.3). Fulfilling this requiren1ent depends on the 1neans of sa1npling, handling and transporting, storing and processing in the laboratory, as \Vell as on the nature of the particular soil. Gravel without fines, for exa1nple, can hardly be satnpled successfully for density testing. Where there is any douht ahout the applicability of lahoratory testing for a particular property ( e. g., pern1eability, cotnpressibility, etc.), suppletnentary field investigation should be carried out (Chapter 1.3). 2. The mass and dimensions of san1ples to be tested in the laboratory 1nust be suiiicient for all necessary testing. The required n1ass and din1ensions of sa1nples depend on the 1naxitnun1 grain size and the size of the apparatus, or on specific recommendations (Table l). Some properties can exhibit anisotropy (e.g. per1neability, con1pressibility, stiffness and strength) and testing of oriented specimens may therefore be re4uired (e.g. setting up specimens perpendicular to the direction of sa1npling). Testing procedures may change soil properties, so material \Vhich has been used for an analysis, e.g. for detern1ining the \Vat er content, n1ay not be appropriate for further testing, for say, particle size distrihution. The size of the samples should therefore correspond to the sun1 of all the required specin1ens.
2.3
Laboratory investigation - performing and evaluating
In the laboratory, physical properties ( e. g. grain size) or conventional characteristics (e.g. Atterherg limits) are detertnined. To guarantee comparahle results, unification of testing procedures and of data evaluation is necessary. In Gern1any, standardization is represented by the DIN Codes of Practice (DIN 18121 to DIN 18137 [46a-s]), by the regulations of the
1.4 Properties of soils and rocks and their laboratory determination
121
Road Research Society (TPBF- StB [59a, b]), by recommendations of the German Society for Geotechnics (DGGT, [53a-j}), etc. Internationally, the most common references are the documents by the American Society for Testing and Materials (ASTM [3-5]) and the American Association of State Highway and Transportation Officials (AASHTO, [I)), or British Standards (BS 1377 [25, 26]). Recommendations for laboratory testing have also been produced within the International Society for Soil Mechanics and Geotechnica/ Engineering (ISSMGE) by different Technical Committees of the Society (e.g., the document by the regional European Technical Committee ETCS issued in 1998 [83]) and by the International Society for Rock Mechanics (ISRM). Despite the codification of testing techniques and equipment, there is generally a scatter of the results. Repeatability of the results obtained by a single operator is generally better than cornparability of the results from different laboratories. An experimental determination of the variation by comparative studies may involve a further error, induced by the distribution of the samples to individual laboratories, portioning/subdividing of the samples etc. Variations due to individual testing techniques (see column e in Table 1) are markedly smaller than the variations exhibited by natural soils in situ (column/in Table 1). Nevertheless, they should still be considered in data evaluation and quality control [ 144]. Testing variations result in randorn errors, while systematic errors may be related to the testing procedure and can be caused, for example, by wall friction, or by the influence of end platens. Tf the testing technique cannot be improved, systematic errors should be taken into account in evaluating the tests (see also EN 1997-1, 2.4.3).
2.4
Soil properties and laboratory testing
Soils consist of solid particles and voids - pores- filled by liquid (water) and/or gas (air), forming two-phase or three-phase systems. Some soil properties and characteristics either depend only on the nature of the solid particles, (e.g. particle si7.c distribution, mineral composition, density of solid particles, shape and roughness of particles and limit densities) or express some interaction between the solid and liquid phases, (e.g. Atterberg limits, water absorption, or water contentcompaction relationships). For the testing of these properties, the only sampling requirement is to preserve the shape of particles and their si7.c distribution (sample quality 4 according to DIN 4021, see Chapter 1.3). Other properties and characteristics depend on the proportion of solids in the total volume of the soil. and also partly on the arrangement of particles (fabric) and the proportion of the liquid and gaseous phases, (e.g. density, water and air permeability, capillarity, swelling pressure, or slake-durability). Sample quality 2 is required to test them. Stress-strain relations and strength properties depend on the fabric and on the forces interacting between the solid particles. They can be measured only approximately, even with sample quality 1, since all sampling cause changes in stresses. The tests for particle size distribution, Atterberg limits and organic content arc called classification tests. The water content together with the consistency limits, or the current density with the maximum and minimum limit densities, represent the tests identifying the soil state. An overview of the typical characteristics of different types of soils is given in Table2.
122
Paul von Soos and Jan Boh
Tablt! 1. T,ist of important laboratory tests on soils a ~
b
-~
Fine-grained soil 1
Sieving
2
Sedimen
I
Minimun1 sample
TeH
c
d
Required sam/ilc qua 1ty DIN 4()21
CharacteriHics
4
Grading curve
4
Grading curve
Gravel
Sand
-
U, Cc, dw
I talion
Particle density
3
Loss on
4
I
50g
''
50g
50 g
up to 2000 g
4
Q,
> 10 g
> 100g
upto10kg
4
v,,
4
Ye.
ignition content -
-
-
7
70cm 3
Density
6
--
Minimum and maximum ·'density
8
t
-
10 g uf gruum.l sample
Ctlrbonate
5
-
400cm'
1 to 3kg
r
'
> 2200 cm 3
with 3: n, e, S, with 7: D, 10
2
6 to90kg
Q, Qd;
maxe '
> 10 g
Water
up to 10kg
100 g
minn mine max Qd
maxn
4 '
min
Qd
I, w; together with 9: le
3
content -
9
Auerberg limits
200g
-
-
4
10
Shrinkage limit
200g
-
-
4
w,
2gof pulverized sample
-
-
4
w,
Water
11
adsorption
WL• Wp,
Ip
i
I
200to 1000cm 3
up to 10 l and more
2(1)
k, (k.), (k.)
permeabilily
Specimens 0 >50mm
Air pem1eabilily
Specimens 0 >50mm
200 to 1000cm 3
up to 10 I and 1nore
2(1)
k,
~ 1000cm 3
4(2)
h\p1 hh
6 to 90kg
4
Qvpt' Wup1
Water
12
I i
13
14 !
15
16
---
I Proctor ! te_st OeUometer tesl
17
Unconfined compression lt'st
18
Triaxial compression test
19 -
-
Capillary rise
Shear box tesl
I Specimens 0 > 50mm
;;:>::
50cm 3
i ' I
-
3to12kg
3to12kg
mod
'
I Specimens 0 ~50mm
Specimens0 Specimens 0 250 mm or ~70mm 0 ;::= 6dm"' !
1
I 1000cm 3
q"; E
1
1.p', c'; 1.?u• Cu; 1.p.,., C.,.
4
I
(10 l) '
I
s'(cr); e(rr); Eoed(rr);
1
t
1000 cm 3
mod w0 r,
C 0 , C,; Ca
Specimen' 0 [, only fm mixed~ 36 mm grained soils h~20 Specimens0~6dmax;h~20 at least 3 specimens as in 17
Q.,p1 ,
!
'
L: after Lumb [107], T: after Tiedemann [ISIJ,A: afterASTM, F: after ··Ringanalysen Forschungsgesellschaft fiir das StraBenwesen ', KO!n.
E;~ -
1.4 Properties o( soils and rocks and their laboratory determination
e
f
')
-
-
2+~( 0/o)1) A 0.7 °/o
0.15°/n 0.2 O/o
(F) (L)
=_§_
(F)
~
I'
(L)
(F) (L)
1 to 2 °/o
Q,:
1 day
DIN 18123
2 Jays
DIN 18123
1day
DIN 18124
I
-- -
-
-
~
1 day Q:
(T) (T)
0.7o/o 0.6°/o
DIN-Code
All data: (R)
2days
max g": 1.3 °/o ming": 1.8 °/o
h
-
"
Repeatability
A < 0,002mm: 11.4 °/o (L)
0.3 o/o 1.4 °/o
v
"
5+~0(0/o)l) (F)
g Ti!ne for one test
In situ scatter
Test data scatter (Variance V=Comparability
123
3 to 8°/o
(A) (A)
'
r
--
DIN 18128 --
DIN 18129
1 hour to 1 day 1 to 2 days
DIN 18125 Part I DIN 18126
1 day
DIN 18121
- -
w: 10 to 22 °io Ip: 10 to 27 °/o 6.0 O/o wP: 8. 7 °/n WL:
(L) (L)
I
1-2°/o 2.9 O/o
(L); wL: 9 to 28%1 (L) wp: 6 to 16°/n ! I,.: 18 to 40%
!
1 to 3 days
-
DTN 18122 Part 1 DIN 18122 Part 2 DIN 18132
r-
1 to 2 days
' 3 hours to 1 day
I
I
---
20 °/o
-
I to 14 Jays
DIN18130
1 to 3 day<:
-
l day
-
I -
I I -
-
~
I -
1.30/o (F~ 4.0 to 25o/o (F t: 1-6°/o (L)
0.5 o/o 2.0 to 7.0 °/o 0, 7 O/o
(F) (F) (L)
-
Cc: 10°/n
-
Cv'.
(L) (L)
Qopl: rl>pl :21 Q0
32 °/o
-
Sand tan
Clay: D-Test tan
J
2\
DIN 18127
7 to 20 days
DTN 18135
C,: 26 to 53 °/0 E,: 25 to 49°/a
l day
'
DIN18136
I (L) tan
fine-grained soils 6 to 19 °/n Cu: 25 tO 68 °/0 -~
1
l to 3 days ~
Cc: 26 to 47°/n
q": 15 to 62°/o Sand, Silt tan
-
~-
The value corresponds to particles retained on sieve A in o/o. Lower liruit: soil class UL; upper lin1it: ~oil class GW.
2 to 14 days
2 to 7 days
i
DIN 18137 Parl 2 DIN 18137 Part3 -
-
Paul ,,.·on Soos and Jan Boh
124 Table 2. Soil characteristics
4 25 4 30 4 I 11 7 ' 25 7 16 ' I 16 I 28 ', 33 I 55 '
I
I
! I
10
30 -
I
-
16
I!
I
2 5 10 100 30 300
> 60 :
<5
SW. Sl I
I
w,
I
20 40
8
15
i
'
15
GO.GT
Sand, well graded and sand, gravelly
Atterberg limits of paiticles < 0.4 mm
Cu
I
6
Uniformity coefficient
F
-
-
-
100 250
30 80
!
50
I 170
!1lote: The soils designated in column (a) should be interpreted more properly than just according to the group symhols in column (h). Their limits are defined in terms of the parameters given in column (c).
125
1.4 Properties of soils and rocks and their laboratory detennination
The two lines in colu1nn (d) to indicate upper and lower values for the characteristics of the soils in column(a). Fu1ther, values in a row are only valid for colu1nns linked under a headline letter such as forexamrle (e). The values given in (c), (e) and ([)depend only on lhe physical behaviour of lhe solid particles. The values in lhe other colu1nns depend also on the consistency index le or on lhe rela1.ive densily Io.
126
3
Paul von Soos and Jan Bohac
Properties of rocks
A rock mass is con1posed of intact blocks of rock material divided by joints. Consequently, the pcrn1cahility, stress-strain behaviour and strength of rocks arc governed hy the nature, history, geon1etry and frequency of joints, as \\'ell as by any filling nlaterial. The geometrical and physical properties of joints and their influence on the properties of the rock have usually to be determined in situ since the effective volun1es to be tested arc generally very large, typically many cuhlc metres. Only in the case of relatively soft, thin-bedded or fissured sedimentary rocks, core san1ples of adequate dimensions can be sufficiently representative. Rock samples tested in the lahoratory can generally yield only properties of the rock material itself, which are independent of the joints, and cannot give the properties of the rock mass. However, the roughness of the joints and the properties of the filling material can be determined in the laboratory. In rocks, the bonds between particles arc so strong that splitting them into individual particles is in1possible in the laboratory. Therefore, texture and structure are examined on joints and density, permeability, strain and strength properties on the most intact cores possible. The anisotropy of rock is usually even more important than in soils. Moreover, the axes of anisotropy of the rock material and of the rock mass do not have to coincide. Reco1nmendations for the laboratory testing of rocks \\'ere published by the Gern1an National Society of ISRM ("Deutsche Gcscllschaft filr Gcotcchnik").
4
Characteristics and properties of solid soil particles
4.1 Particle size distribution Particle size distribution is one of the most in1portant physical characteristics of soil. It expresses the percentage by mass of individual size ranges. Many geotechnical properties are closely related to particle size, \\'hich is therefore a convenient indicator of the probable soil behaviour, and serves as a main classification attribute for soil description and classification (see Sections 11.1and11.2). The size of soil particles varies roughly fron1 0.0001 to 200 mm. Table 3 shows the individual particle size classes used in the geotechnical description and classification of soils. Silt, sand and gravel sized particles are further subdivided into the subclasses of fine, nlediun1 and coarse (see also Fig. 3). Although the divisions hctwccn individual classes, Table
3. Definition of particle size groups
Grain size range stnaller
than
0.002 0.063 2.0 63.0
lo to to to
greater
than
Name 0.002 tnn1 0.063 mm
Clay Silt
2.0
Sand Gravel Cobbles Boulders
mm
63
tnn1
200 200
mm mm
Fig.1. Definition of particle sized
127
1.4 Properties of soils and rocks and their laboratory determination
especially between sand, silt and clay, arc rather arbitrary, they have proved useful for engineering purposes. In practice, the term "clay" may be used to express either mineralogy or particle size, or type of soil, which may son1ctimcs be confusing. Clay, silt and sand fractions can be represented as a single point in a triangle diagran1 (Fig. 2). A similar diagram may be used for fines ( < 0.063 mm), sand and gravel, i.e., all particles sn1allcr than 63 mm. The distribution of clay, silt, sand and gravel+cobble fractions can alternatively be expressed by the so called grain size distribution index, in which the percentage of the individual fractions is rounded to tens and then stated, starting with the clay fraction. For example, the soil in Fig. 2, with 33 % of clay particles, 52 o/c) of silt and IS o/o of sand would have the grain size distribution index 3520. Particle sizes over 0.06 nin1 arc dctcrn1incd by sieving, the fines, sn1allcr than 0.06 nin1, arc subdivided by sedimentation (DIN 18123 [46e]; ISSMGE, 1998 [83]; AASHTO T88 [1]; BS1377 [26]; ASTM D422, etc.). 0 100%
80
(so 20
80 \
100%00
\v.1~~~-~~~o 20
40
60
80
100~0
0002 to OOOJmm silt
Fig. 2. Representation of particle size classes in the triangle diagram \So/o Exa1nple: Sand Silt 52o/o Clay 33o/o
4.1.1
20 0
00005 OW6 002 02 053 0002 0,063 d
I
zo
53 20
200mm 63
Fig. 3. Particle size distribution curves
Sieving
In sieving, particle size classes are separated by the use of sieves constructed of metal wire cloth or of perforated metal plate in accordance with an appropriate standard (e.g., DIN 4187). An adequate number of sieves should be used to gain a continuous grading curve in the range fron1 0.063 to 125 n1m. The individual particle size classes arc classified by the aperture of the last sieve passed. Therefore, because of the tolerance of the sieve aperture and the large variety of shapes of particles, the size intervals dctcrn1incd arc nominal values rather than strict physical dimensions of the particles (Fig. 1). The results are plotted as cumulative percentages of the dry weight passing each sieve against particle size, as a semi-logarithmic grading curve. A steep portion of the grading curve indicates the prevalence of the corresponding interval of particle size, while a flat part shows a lack of particles. TI1c shape of a grading curve can be
128
Paul von Soos and Jan Iloh
quantified using its three characteristic points D10, D30 and D6o, defined as the 1naximum size of the smallest 10 °/o, 30 1Yo and 60 'Yo of the sample. Both uniformity coefficient Cu = D60/D10 and coefficient of curvature (or coefficient of gradation) Cc = Dj0/(D60 x D10) reftect the shape of the grading curve. Cu = Cc = l represents a single-sized soil. Cu < 5 indicates a uniform (poorly graded) soil, 5 15 a strongly non-uniform soil Cc expresses the relative position of the characteristic points. A s1nall value indicates that D30 lies close to D10, and a high value that D30 lies close to D6(). Most well graded soils have a C, in the range from 0.5 to 2. In preparing specimens for sieving, particles s1naller than 0.06 tnm (i.e. fines) should be washed (the so ca11ed v..·et method). Speci1ncns without fines can however be subjected to oven drying at 105 "C prior to sieving. According to the maximum particle size, representative specimens can he prepared from samples by riffling or quartering. The inini1nu1n dry masses required for sieving are given in Fig. 4.
mm ~ "
IDD
l
!
60
ll! " :ll 1D ~ 20
~
-~
V'
I
Vj
Ii''
,
l '
~
QI""'
I
1 '
-
Fig. 4. Recommended sample mass with respect to particle size
ID 10 "' 1DD ;g
minimum sample mass md
4.1.2 Sedimentation The procedure for sedi1nentation is based on Stoke's law, which states that in a suspension (soil-water in this case) the larger particles sink more quickly than the small ones. Stoke's law assumes the particles to be spherical. The dia1neter of the spheres that settle a particular distance in a measured time interval may then be obtained. The soil-v..·ater suspension is placed in a glass cylinder and at a number of suitable ti1ne intervals the density of the suspension is measured at a reference depth using a hydrometer (Fig. 5). Corresponding equivalent dian1eters of spheres that would sink at the sa1ne velocity as the soil particles are then computed fron1 Stoke's law. Using equivalent diameters as particle sizes, a grading curve of fines can be drawn [34]. This method is applicable for particles from 0.001 mm to about 0.100 mm, using 20 to 50 g of soil in l litre of water. A dispersing agent is added to the suspension to prevent ftocculation. Haas [68] has suggested the use of less than 20 g of soil to keep ftocculation under control and to measure the density of the suspension by the uplift force acting on the hydro1neter to avoid errors connected with the hydro1neter readings.
4.1.3
Sieving and sedimentation
Soils composed of a substantial fraction of both fine and coarse particles arc analysed by a combination of sieving and sedimentation. The fraction of particles larger than 0.125 in1n is
1.4 Properties of soils and rocks and their laboratory determination
129
~
9/cm 1 a995 UJ{}{}
1 I
""'
lOJO
l
¢~
JoLl
volume 50or 100 cmJ
lead ballast
Fig. 5. Jlydrometer after
F1g. 6. Density bottle
Bouyoucos!Casagrande
obtained by washing and then sieved separately, \vhilc fines arc subjected to a hydrometer analysis. The results are then presented in one graph. lf a discontinuity arises in combining the data, priority is given to the results of the sieving f83]. Brittle particles (e.g., of residual soils) are susceptible to breakage by an intense mechanical treatment during specimen preparation. Fines produced by abrasion of the particles may also distort the results of the analysis. The time required for separation of fines from coarse-grained particles for hydrometer analysis nlay be reduced by using the sedimentation technique after Haa.s and by using small amounts of fines (TPBF-StB Part B 5.2 f59b ]).
4.2 Density of solid particles The density of solid particles p~ is their mass Mct divided by their volun1e V8 ; p5 = Mct/Vs. Soil particles sometimes contain enclosed voids which are included in the volume Ys. Therefore an apparent density of solid particles is determined. The standard procedure for determination of density makes use of the density bottle (pycnomctcr) to determine the volume of the soil specimen by the water displacement method (Fig. 6; DIN 18124 [46f]). With particles up to 4mm, about 30g of dry mass of soil is placed into the density bottle_ Distilled water is added and the content de-aired. From the measured masses and the known density of water, Psis calculated. For soils with constituents that react with water (e.g., anhydrite, organic soils), a different control liquid may be used, for example trichlorethylcn, kerosene, toluene etc. To increase the accuracy of nleasuring particle density on larger specin1ens in pycnon1eters, Haas's or Neuber's methods can be used (TPBF-StB Part B3.2 f59a] and DIN 18124 f46f], respectively). In the Neuber nlethod the dry specin1en is de-aired before de-aired water is added in the pycnometer. The procedure has proved effective especially in testing rocks. Typical values of particle density, \vhich depends primarily on mineralogy, are given in Table 4.
130
Paul von Soos and Jan Bohac
Table 4. Particle density of some minerals in Mgm 1 (gcm- 3)
Gypsum Feldspar Kaolinite Quartz Na-feldspar Calcite II lite
4.3 Mineralogical composition of soils Mineralogy is a decisive factor controlling the size and shape of the soil particles, the soilwater interactions and the mechanical hehaviour of soils (plasticity, fluid conductivity, compression, strength). It is governed mainly by the origin of the particular soil. Two main groups of soils can he distinguished: coarse-grained soils consisting primarily of non-clay particles, and fine-grained soils composed of a substantial proportion of clay minerals. Coarse-grained soils originate from mechanical weathering. They are therefore primarily formed by rock-forming minerals, for example quartz, feldspar, mica, calcite and dolomite and are mostly of bulky shape. Clay-sized particles of fine-grained soils are primarily composed of clay minerals which are the products of chemical weathering of rocks. Clay particles are platy, some are needle-shaped or tubular. Dominant clay minerals are hydrous aluminum silicates, whose structure is constructed from two simple structural units, the silicon tetrahedron and the aluminium octahedron. Individual clay n1ineral groups, e.g. kaolinite, halloysite, montmorillonite, illite, chlorite etc, are formed by sheets or chains of the basic structural units and are characterized hy the manner in .vhich t\VO successive ltt.'O- or three-sheet layers are stacked and held together fl13J. In Tahle 5 for example it can he seen that montmorillonite does not exhihit a constant distance between layers, \vhich enables .vater to enter the structure and cause swelling. 1
1
In clay minerals, some of the tetrahedral and octahedral positions are occupied by other cations than those in the ideal mineral structure. For example silicon can he replaced hy aluminium, aluminium by magnesium, and magnesium by iron. As a result of this so called isomorphous substitution the clay particles ohtain a net negative charge. To maintain electrical neutrality, other cations, most of .vhich are exchangeable cations, e.g. Na, K, Ca, Mg, Fe, are attracted and kept within the particles [113]. The quantity of exchan[{eable cations, cation exchange capacity, of typical clay minerals is given in Table 5 in milliequivalents per 100 g of dry clay. The exchange capacity and the nature of exchangeable cations strongly influences the engineering properties of clay minerals. 1
The bulky particles of coarse-grained soils can he distinguished with the use of a magnifying glass or an optical microscope. For studying clay particles and their mineralogy, X-ray diffraction analysis, thermal analysis (DTA), optical (polarizing) and electron microscopy can be used. The X-ray diffraction method makes use of the typical differences in the angles of reflection of different crystals. The X-ray diffraction is particularly well suited for the identification of clay minerals. The common non-clay minerals of soils are also detectahle
131
1.4 Properties of soils and rocks and their laboratory determination
Table 5. Structure of clay minerals St:ruC(Uf'dl rnoi.lel
by this method. DTA is based on the phenomenon that endothermic and exothermic reac· tions are triggered at different temperatures, and that individual minerals exhibit typical characteristic thermograms. The curves obtained in the analysis arc compared with those for known material so that the soil composition can be determined [113].
132
4.4
Paul von Soos and Jan Bohat
Shape and roughness of particles
The shape of soil particles describes their geometric form, which reflects their origin, history and internal lattice structure. The form varies widely and in particles precipitated by organisms may be exceedingly complex. The measure of particle shape is sphericity, defining the degree to which a particle approximates the shape of a sphere (see, e.g., [621). IIowever, for engineering purposes the following shapes of particles may be distinguished: spherical, se1ni-spherical, pris1natic,fiat, tubular or needle-shaped and platy (Fig. 7). According to their roundness, particles can be very angular, angular, subangular, subrounded, rounded, and ivell rounded (Fig. 8). Roundness is related to the sharpness of curvature of the edges and corners. Therefore, roundness nlay be viewed as a measure of the macroscopic roughness of particles, Roundness is geometrically independent of sphericity: particles of both high and low sphericity can exhibit any class of roundness. On the micro-scale, roundness can be studied by scanning electron microscope. ln coarse-grained soils, bulky particles prevail. Their shape and roundness depend on the parent rock and on the history of \Veathering and transport. The more transportation has occurred the higher rounding of the edges and breaking of asperities. However, weathering that can follow may increase the roughness again. With fine-grained soils the shape of particles is controlled by the mineralogy. Most clay minerals are platy, halloysit is needleshaped, and quartz, calcite and dolomite are bulky to prismatic (see Section 4.3). The shape of particles influences soil fabric and anisotropy and their roundness (roughness) affects stress-strain behaviour and strength.
The roughness of grains in a sand can be determined from the exit velocity when poured through a nozzle. Roughness coefficient r is computed from the comparison of the exit times of the investigated sand and of a reference material. For the reference material (crushed quartzite) r = 1.0, for ideally smooth spheres r = 0. Correlations have been suggested between rand shear strength and stiffness ([88, 89]; see Section 10).
Fig. 8. Roundness of particles 1. very angular, 2. angular, 3. subangular, 4. subrounded, 5. rounded-well rounded
Specific surface
The specific surface As is the area of a particle expressed \Vith respect to 1 g of its dry mass
A
a
As=-=-md d · Ps where shape factor a becomes - for bulky grains (e.g. quartz) - for platy particles up lo depth/diameter ratio 0.1 (e.g. kaolinite, illite) - for platy particles up to depth/diameter ratio 0.01 (e.g. montmorillonite)
a=6 a =24 a =204
1.4 Properties of soils and rocks and their laboratory determination
w----
10mm
am a1
133
an.
10
Imm
j O.lmm 'C>
quartz
10µm ~
lµm
kaolin 111 111
aTµm 0.001 0.01
montmorillonitt
i
I,. l//;tv,
~
Ill I I Ill 10
0.1
100
1000 m1/q
Fig. 9. Particle size and specific surface of some minerals
A,
Dependence of the specific surface of some comn1on nlinerals on the particle diameter is shown in Fig. 9. The specific surface obviously increases with reducing the particle size, since the volume and area are proportional to the particle diameter cubed and squared, respectively. Therefore, as the soil solids dimensions are decreased, the proportion of surface area to weight becomes larger. The same applies for bonds, unbalanced surface electrical forces, etc. Thus specific behaviour and properties studied by colloidal and surface chemistry become more pronounced \Vith smaller clay particles. The amount of water that can he adsorbed by the surface of minerals is also proportional to surface area. In the laboratory, specific surface area can be estimated by the amount of, e.g., ethylene glycol or glycerol, adsorbed on the clay surface [113].
4.6 Organic content Even a small adn1ixture of organic matter can bind a substantial amount of water into the soil, increasing porosity and influencing deformation and strength properties. Organic matter content in soils is responsible for high plasticity and shrinkage, high compressibility, low permeability and low strength. Organic content is very well determined by treating the soil with hydrogen peroxide solution. The most common laboratory nlethod determines the loss on ignition of nlass of an oven-dried specimen at 550"C (DIN 18128 [46j); 450°C according to AASHTO T26786 [1 ]). The results are expressed as a percentage of the dry original mass of the specimen. They may easily he distorted however by the additional loss of water hound by clay minerals and by other chemical reactions. Wet combustion can be used for organic carbon content determination, especially for soils with the humus-like easily oxidized organic material other than fresh plant and similar undecayed vegetative matter (AASHTO Tl9487 [1, 132]).
134
4.7
Paul von Soos and Jan Bohac
Carbonate content
Carbonate minerals, calcite and dolomite, can be present in soils in the form of particles, shells, precipitates and segregations. According to the form of occurence, they can markedly influence plasticity, compressibility, strength, collapsible behaviour etc. The carbonate content can be indicated qualitatively by treating the soil with dilute hydrochloric acid (HCl), which results in strong effervescence if carbonates are present (DIN 4022-1 [48]). For quantitative analysis, Scheibler's device can be used (Fig. 10), in which the amount of carbon dioxide produced during the test is measured. Since the reaction of dolomite to HCl is delayed in comparison to the reaction of calcite, the carbonate content of the two constituent parts can be approximately distinguished in the test (DIN 18129 [46k]). nlling bl al -
o
measured volume
Jlii'kd~HC/ / soil specimen position of flJ/J/Jer /Jal/ (C02-air interface) equalizing pressures
Fig. 10. Apparatus for carbonate content after Scheibier a) initial water level b) water level after C02 development c) v.:ater level at volun1e reading
5 Characteristics and properties of soil aggregates 5.1
Fabric of soils
The nature and behaviour of soils depends not only on the properties of individual particles, but also on the properties controlled by the arrangement of the particles in the soil fabric. Its nature depends primarily on the origin, nature and size of the particles. With gravel, sand and silt the influence of molecular attractions and electrical charge is relatively small and the particles can freely occupy soil voids and arrange themselves in single-grain fabric (Fig. 11-1 ). These are rare however in soils with clay-size particles, where interparticle forces induce typical associations of particles: aggregated (face-to-face associated) or dispersed (with no face-to-face association). Particles usually associate in edge-to-edge or edge-to-face flocculated fabrics, or can remain deflocculated. Dispersed edge-to-face or edge-to-edge flocculated arrangement, cardhouse fabric, Fig. 11-2, is commonly found in freshwater sediments. In salt water however, due to high electrolyte concentrations and lower ion concentrations, aggregated and flocculated cardhouse fabric occurs, which is even more open (Fig. 11-3). In residual soils, an open fabric can develop as a result of leaching. In compression, the particles or their parallel aggregations of the cardhouse fabric tend to re-arrange themselves perpendicular to the loading (oriented compression fabric), in
1.4 Properties of soils and rocks and their \ahoratory determination
•~m
1 single grain fabric
2 card house fabric
3 flocculated and aggregated fabric
135
Fig. 11. Fabric of soils
shearing they take the direction of the shear load to form shear planes or shear bands (oriented shear fabric). During kneading or other kinds of dynan1ic compaction the interparticle bonds deteriorate and flocculated aggregations get broken. This is n1anifested generally by loss of strength (sec sensitivity in Section 7.3). On the macroscopic scale, a nlore hon1ogeneous fabric is built up, in \Vhich large pores are generally eliminated, although some macropores may still be present (see Section 5.8). Fabric can be studied by direct observations using an optical or electron microscope, by X-ray diffraction, electrical or thermal conductivity and n1agnetic susceptibility. Indirectly fabric can be described using phase relations, e.g. by pore size distribution. Other indirect methods are represented by (phenomenological) studies of the mechanical behaviour, which can yield an indication of the soil texture and structure. Fabric can therefore be estimated on the basis of the soil permeability, compressibility, strength etc. (113].
5.2
Porosity and voids ratio
The ratio of the volume of voids to the total volun1e of soil, porosity n, can be depicted in a phase diagram where each of the three phases (solid particles, liquid and gas) are represented separately as shown in Fig. 12. The portion of voids filled with water and air are nw and Ila, respectively, where n = nw +Ila. The ratio of the volume of voids to the volume of solid particles is called the voids ratio e. From the phase diagram it can be easily shown that n e Ilw Ila e---· c ---· n=-ew=--; a-1-n' - 1 - n' 1- n 1+c If nlasses and volumes are considered in the phase diagram, a relationship het\veen voids ratio, ratio of mass densities Ps/Pw (see Section 5.3), and water content w (Section 5.5) can he expressed: p, ew=W-
Pw
A list of phase relations is given in Table 6. The degree of saturation Sr= nw/n = e,,..;e = Yw/Yv expresses the portion of the voids filled with water (often given as percentage).
F1g. l2. Definition of porosity and void ratio
Table 6. Phase relations Ps and pw are known a)
~ "' Porosity n
Void ratio
;:!!
'
~
Saturatiou
•> ~
•••
" ~
1 ~
,,
~
. ..• i
=•
.E
n n 1-n Ilw n
' ' 1 -· e
-·
Density of particles Psb)
P.::__::- n· Pw 1- n
Pr - e ·(pr-pw)
Water conlen1 w
n Pw --· 1-n p;
,. Pw p;
Soil density Psil:t
ll -n) · Ps -n ·pw
·----
Water content
-
-
-
-
-
-
\Vater porosity
nw
\\later voids ratio
p
1-
-
a) In practice Pw = 1.0 t!m3 --'- 1.0 g/cm 3 h) In this row psai, fld and fl are known instead of P> c) Relations in this column are valid only for water content of saturated soil (w = w,at)
p
p;
(1 +w) · Ps
p~ -1 Pd
(1 +w) · ~ -1 p
~ _ps PwlO +w) · Ps - p] p
1-· w
~w_
Pw Pw .
(1
- W·
Pd
(~ ~) -
£!!) ·rw+Pct p;
P·Pw p-(l+w)·(pr-pw)
Pw
c+w -'-) p
W·
£!!.
,,
W· -
Pw
(1 . w) ·Pd
p~
(1-:;)
pw+-'-
1 "i w
w
Pw
'w
Soil density
e
PJ
p~-e·pw
1 _,
w
Pw · (Ps -
'
(1 - n) . P;
p
w
W·Pd·Ps Pd)
Ow
-
Dry density Pd
Pd
l
' p; 1 _,
w ~
Gener11lly "t-alid
n
w w
p
l ., w
Pw
--
W·
E;. Pw p
...,,
Table 6 (continued) Ps and Pw arc known aJ
WC)
w air porosity nn = n-n....-
11sat ~
n
ws<1t ·Ps Wsat ·Ps+Pw
Vuid ratio
Ps Wsat·-
Porosity
l'w
Pr·Pw
Water con1ent
w
w
Soil density Psat
(l+w~at )·p~
W·ps W·p~ I Sr·Pw
'
' I+<
,9.,
,
' '• ,,'
"" 2c ,,,. ""
w·11s+na·11w na)·Pw
W·Ps Sr·Pw
n 1-n
(1-l\i.)·w·pw
w
nw
W·ps+Da·Pw
Wges
n
(1-na)·ps·Pw W·Ps+Pw
Sr·Ps-Pw W·ps+Sr·Pw
0-n)·Ps
P ·Pw '1+w)·0 -na)·pw -w-11
Sr·p·pw
p-nw·Pw
(l+w)·Sr·Pw-"' ·p
1-n
•
n Pw 1-n
"
----
I
11r·pw
-
Da·(W·ps+pw) (1-na)·ps
Ps Pr Pw Pr- Pw Ps
w+------· (1 +w)-0-na)·p; ·Pw
p,..
W·Ps+Pw
W.<;.at Ps+Pw
Wat er porosity nw
+na·Pw
(1-1 I
Pw
""
p, I e·rw
'"
'<
l+e
n
<·-
~
-~
0.
,
Cw·-
"5·3
W·ps (1-na)
Sr·W·Ps W·Ps+Sr·Pw
"w
Ow 1+e
5
nw 1-n
'•
(1-n)·ps+nw·P
Ps+ew-Pw l+<
P;
w
p
W·ps+Sr l'w
(l-n)·p~+n·11w
~-
nw flw 1-n
~·oids
Suil Uensi!y
~-±w)·p~·ew
,,
(1+e)-p-e....- ·Pw
0.
;.
w
W·Ps+Pw
ratto
s,
1+e
0.
w
w
I
n ,"·
P> Pr Pr-Pw
Water content
\Valer
Ps-Pw
~p~
Density of particles p.,b)
nw n n
'
Wsat ·ps+Pw
w Sr < 1 -----
s, Pd
~
W·Ps+na·rw W·ps+Pw
Saturation
Dry density
"
'5
Partiall}' saturated
Saturated (Sr - 1.0 wa - 0)
w)·O
Pw n;i)·p,·pw
--------
W·Ps+Pw
a) In practice Pw =1.0 tim3 = 1.0 g/cm 3 b) Jn this row Psat· Pd and pare known instead of Ps i.:) Relations in this column arc valid only for waler content of saturated soil (w = Wsad
,,
W·-
Pw (l+w)·Sr·rs-rw w·p~+Sr-Pw
,,
Pw
,,,
~-
138
Paul von Soos and Jan Bohac
In the capillary zone above the groundwater table (the free water surface), two zones with capillary water can he distinguished. Below the capillary saturation level there is full saturation. Sr = 1. Above the capillary saturation level. there is a partially saturated zone. Its extent depends on the nature of the soil, predominantly on the particle size. Nevertheless. for simplicity full saturation is often considered in ground investigations and in design. Therefore, for most of the laboratory testing full saturation is required in practice. Even on the samples procured deep below the ground water table, Sr < l is often found, since sampling results in a decrease of pore pressures, which can cause the release of gas hubbies from the pore water of the sample.
5.3
Density
Density is mass per unit volume. However, several densities arc used in geomechanics: the total (wet, bulk) density p, the dry density Pct• the saturated density p,,,, the density of particles (solid density, Section 4.2) Ps and the density of waler rw: Mt Md+ Mw r=-=----
v,
v,
Mct Pct=Yt Mct Ps=-
V,
Mw
Pw=-
Yw
where Mis mass, Vis volume, and subscripts w and a denote mass or volume of water and dried specimen (skeleton), respectively; subscripts sat and t mean saturated and total. ln measuring p and Pd in the laboratory, the total volume of the specimen must be determined. With specimens of regular shape. the dimensions can he measured directly. The volume of an irregular specimen can be determined by immersing the specimen in water and either weighing its mass (immersion in water method), or measuring the volume of displaced water. Prior immersing into water, the surface of the specimen must be treated: all the surface air voids must he filled hy a suitable material insoluble in water (plasticine, putty) and then the surface coated by paraffin wax ([73, 83], DlN 18125 [46g]). From the phase diagram, useful relations for densities can then he derived, for example Pd= (1 - n)p,
"*
n = l - Pct/p,
P = (1 + w)pd = nw · Pw +Pd Psat =n · Pw +Pd
5.4 Relative density ln its loosest stale an ideal packing of spheres of an equal diameter (Fig. 13a) will have porosity n = 0.476 (e = 0.908), while the densest packing (Fig.13b) will result inn= 0.259
139
1.4 Properties of soils and rocks and their laboratory determination
I•ig.13. Loosest (a) and densest (b) packing of spheres (a)
(e = 0.350). With typical particle density of soils p, = 2.65Mgm- 3 , the minimum and n1axin1un1 dry densities of spherical particles would be min pct = 1.35 and max pd =
1.96Mgm-3 . Minin1un1 and 111axin1un1 densities of poorly graded (uniforn1) coarse-grained soils will not be very much different from these theoretical values for the packing of equal spheres. However, if platy particles are present, e.g. n1ica, both the lin1it values can increase substantially. For well graded mixtures of gravel. sand and silt sized fractions, both the values, especially nlin n can drop considerably below the values for the packing of spheres. Relative density Dd (ASTM D 4254 [5]; density index In according to ISSMFE rules (see Chapter 1.1, Section 2) is defined using nlini111un1and111axin1u111 voids ratio maxe - e Dd = In = - - - - max e - mine The density index Id (ASTM D 4254 [5]) is defined using density or porosity max n - n Id =D= . 1nax n - nun n
Pd - min Pd max Pd - min Pd
1
+mine 1+ e
~--In=
Pd
---In 111ax Pd
Index Ii = (max e - mine)/ mine was defined by Terzaghi [150] to describe compactibility of soils. In the laboratory, the densest packing, max Pd, of coarse-grained soils with particles larger than 0.06 mm (no silt fraction) can be achieved by a tapping method. After being flooded with water, the soil is compacted in layers by tapping the test mould with a tool as shown in Fig. 14. The details of the method are given in DIN 18126 [46h]. After the prescribed con1paction, the height of the specimen is n1easured to 0.1 mm and from the mass and volume of the soil inax Pd is calculated. For soils with a silt fraction of up to 15 °/o a vibrating device (shaking table, Fig. 15) can be used. The specimen is compacted in a cylindrical mould under prescribed vertical normal stress applied via a plunger (DIN 18126 [46hj).
}'ig.14. Maximum density
.Fig. 15. Maximum density test using shaking table
140
Paul von
Soo~
and Jan Boh
To determine 1nin pa, the same mould as used for the densest packing test is carefully filled
with oven-dried soil using a spoon or a funnel (DIN 18126 [46h]). Five relative compaction states of soils distinguished hy the value of the density index arc given in Table 7. Table 7. Relative compaction states
Density index ld = D [%I
0-15
15-30
30-50
50-80
>80
State of compaction
very loose
loose
medium
dense
very dense
5.5 Water content Water content w is defined as the ratio of the mass of water Mw to the mass of solids Md Mw Mct From the gravimetric \\1ater content \V, the volumetric \\1ater content Ow (Ow = nw, see Section 5.2) can be defined as the ratio of the volume of pore-water to the total volume of the soil. It is used mainly in pedology and hydrogeology and is also quite often used in the mechanics of unsaturated soils. The most common laboratory method for determining \\1ater content is oven-drying at 105°C until a constant mass of the specimen is achieved (e.g. DIN 18121-1 [46a, 83]). Adsorbed or chemically bound \\1ater may not evaporate at 105 <'C, and in evaluating its mass this is added to the mass of solid particles. Generally, quartz sands are almost completely dry after the procedure, v..·hile clayey soils containing significant amount of montmorillonite, gypsum or organic matter would lose more water in further drying at temperatures higher than 105 °C. More rapid methods of drying can he used, for example the sand hath method, infra-red or microv..·ave drying. Where appropriate, methods other than drying can be applied, e.g. density bottle, pycnomctcr, air-pycnomctcr, or the indirect method using the development of gas pressure after treating the specimen with calcium carbide (DIN 18121-2 [46b]). Hov..·evcr these methods all cxhihit different accuracies. When rapid drying or indirect methods are used, a check by the standard method of oven-drying is recommended [83]. W=-.
5.6
Limits of consistency- Atterberg limits
In the case of fine-grained soils, any change in water content brings about a change in engineering properties. With the decreasing v..·ater content, deformahility (plasticity) of clayey soils becomes lower, while strength increases. Four consistency states may be distinguished: solid, semi-plastic solid, plastic and liquid state. Soils of different mineralogy and grading pass from one of these states into the next one at different water contents, that arc typical for the particular soil. These \Vatcr contents can therefore he used for description and classification of soils. Although the transition between the states is gradual. three arhitrarily determined consistency liniits - Atterberg limits - arc used in practice: liquid limit \\'L at the transition from the liquid state to the plastic state, plastic limit Wp v..·hcrc soil becomes a semi-plastic solid, and shrinkage limit v..·s. which is the v..·ater content
1.4 Properties of soils and rocks and their laboratory determination
141
at which the soil ceases to shrink with further drying. The Atterberg limits are measured on remoulded soil [8]. Nevertheless, they are useful parameters of clayey soils. The liquid limit WL and plastic limit wp represent the upper and lower bounds of water content within which the soil is in the plastic state. This range of water content is called plasticity index lp = WL - wp. The plasticity index depends on the soil's grading and on its mineralogical composition and exhibits good correlation with some of the engineering properties of soils. Therefore it serves as an important classification attribute. The relationship hetween the natural water content of the soil and its consistency limits is expressed by its consistency index Ic WL-W
le=---WL-wr
or by its liquidity index IL W-Wp
IL=---WL-WP
where le+ IL = 1. Using the value of its consistency or liquidity index, the in-situ consistency of the soil can be classified according to l"able 8. Since the Atterberg limit tests are carried out on a remoulded soil. cementation and any other kind of bonding are lost. Liquidity or consistency indexes cannot therefore correlate with the peak strength of soils. Tue Atterberg limits represent the plasticity of the aggregate clay and silt particles of the soil and depend on both the type and the amount of the clay fraction. The dependence of the plasticity (Ip) of the aggregate on the amount of the clay fraction is represented by activity A (activity index IA) after Skempton [137]
A=
lp(%) % by weight of particles < 0.002 mm
A < 0.75 characterizes inactive clay minerals (e.g. kaolinite), 0. 75 < A < 1.25 suggests normal and A > 1.25 high activity (e.g. montmorillonite ). Typical values of activity A of some clay minerals are given in l"'able 9. The Atterberg limits WL, wp and ws are determined by laboratory test procedures, that are specified in detail by individual standards (e.g., DIN 18122 [46c]; BS 1377 [26]; AASHTO D T89-90, T90-87, T92-88 [1]; ISSMGE, 1998 [83], etc.). Table 8. Consistency of soil le
TL
Consistency; designation
Identification by observation in hand samples
<0 Oto 0.5 0.5 to 0.75 0.75 to 1
>1
1 to 0.5 0.5 to 0.25 0.25 to 0
Liquid; slurry Plastic: very soft Plastic; soft Plastic: firm
Cannot be \Vorked, flows as a slurry Extrudes betv,1een fingers Can be easily moulded Can be moulded with substantial pressure; threads 3 mm in diameter can be moulded 3 mm dia threads cannot be made. the soil cru1nbles: can be re-worked into lumps
>1;W<\VS
Semi-solid; stiff
142
Paul \'on Soos and Jan Bohac
Table9.Liquid limit, activity. and water absorption capability w,\ (Section 5.7) of typical minerals Minerals Quartz nour Kaolinite
Illite Ca-Montmorillonite Na-Montmorillonitc
WL
Activity A
o/o
WA
0 0.4 0.9 LS 7
60 100 500 700
'Yo
30 80 300 700
For liquid limit determination two different techniques are adopted. 'fhe first uses the Casagrande device described, e.g., in DIN 18122 r46c], AASHTO D T89-90 [1], etc
(Fig. 16). The second technique uses a fall-cone apparatus, either the British version (BS 1377 [26J), or the Swedish one (Swedish Standard SS 02 71 20; Fig. 17). The two
fall-cone apparatuses are based on the same principle, the only major difference bet\veen them being the geometry and mass of the cones.
Prior to determining the Atterberg limits the soil should not be dried, since the properties of clay minerals can change on drying. Wherever possible, the natural water content should be preserved in preparing the sample. If necessary, the sample may be air-dried prior sieving on a 0.4 (0.5) mm sieve.
~ 45 H=+==t'.l"iµ::)::I
" 40
robber base a)
b)
Z1
~'
ID 25 40 60 number of blows N -----
u::;io
grooving tool
Fig.16. Liquid limit - Casagrande method. a) Casagrande apparatus, b) data evaluation
adjustment
60g/60'
Fig.17. Fall cone
143
1.4 Properties of soils
In the Casa1?rande method according to DIN 18122 [46c], remoulded soil of an adequate \\'ater content is placed in the cup of the apparatus and a groove of standard shape is cut through it. It is then subjected to shocks by repeated lifting and dropping the cup. The sample has the \\'ater content at the liquid limit when the groove closes at a length of 10 mm (in accordance with DIN) after 25 shocks. Normally the liquid limit is determined by interpolating between at least 4 measurements made at different \\'ater contents (Fig. 16b ). When appropriate, a one-point method based on only one measurement and on an empirical relation may be adopted (DIN 18122-1 [46c]). In the fall-cone method, the sample is placed under the cone, which is then released and allowed to penetrate the soil for five seconds. The soil has the water content at the liquid limit if the depth of penetration is 10 mm or 20 mm in the case of the S\\ edish and British apparatus, respectively. Similarly, as with Casagrande device, the liquid limit is determined by interpolating bet\\ cen at least 4 measurements made at different water contents. A one-point method has also been developed (e.g., [83]). 1
1
In general, the Casagrande method is more operator sensitive, offers worse repeatability and comparability. The apparatus is also more difficult to standardize. Therefore, the fallcone method is preferred r83]- A comparative study by Farrell ct al. showed that there is no practical difference between the results obtained by Swedish and British fall-cone apparatus [58]. Prior to determining the plastic limit, the sieved soil is remoulded at an adequate \\'ater content. A small amount of the prepared soil is moulded by hand into a small ball, which is then rolled on a flat surface using one's fingers and a light pressure until the thread reaches 3 mm diameter. If the soil just crumbles at this diameter, it has the water content at the plastic limit. If not, the moisture content is lowered by re-working the sample by hand and repeating the procedure until crumbling at 3 mm diameter occurs and the water content of the sample at the plastic limit \\'pis determined. To avoid manual rolling of the soil threads and to minimize subjectivity of the method, a device has been developed by Kaiser and Gay [92]. The liquid and plastic limits correspond to characteristic strengths of the soils. At the liquid limit the undrained shear strength Su is about 2 kPa: values of Su = 2.3 kPa and 1 Su = 1.4 were found for soils at their liquid limit of 35 o/ci and 150 Yc1, respectively 61]. forrel/ et al. [58] arrived at Su = 1.6 for London Clay at its liquid limit, using both the British and Swedish fall-cone methods. Despite the plastic limit test being less consistent than the liquid limit, it is also a type of strength, test and the approximate average value of soil strength at the plastic limit is 200 kPa. It has been suggested that determination of the plastic limit as the state where the undrained strength occurs is lOUtimes the strength at the liquid limit p 60]. The relationship of the soil strengths at the liquid and plastic limits justifies, within the framework of the critical states, the correlations between Atterberg limits and soil strength and compressibility [159].
rl
When soil is slowly dried after being remoulded at high water content (at or above WL), its volume gradually decreases. The soil then has the water content at the shrinkage limit when a further decrease in water content docs not cause any substantial decrease of the soil volume (Flg.18). If the residual shrinkage between the shrinkage limit and full drying (oven-drying) is neglected, the shrinkage limit can be computed using the dry mass of the specimen, Ws = (Vd/md - 1/p,)pw (DIN 18122-2 [46d]). During this procedure, reaching the shrinkage limit is obvious, due to a change in the colour of the soil. The volume of the specimen
144
Paul von Soos and Jan Boha~
Fig.18. Shrinkage limit
after drying can then be determined by any appropriate method, for to DIN 18125. Part 1[46gl.
example~
according
As a rule, an open soil fabric produces a high shrinkage limit, while an oriented fabric
results in low ws. Due to remoulding at high water contents, samples prepared for shrinkage limit determination can be expected to exhibit a random, dispersed fabric. Therefore, natural soils with an undisturbed compression fabric can shrink at w < ws. Rolling the threads in the plastic limit test induces more aggregation of platy clayey particles and
produces a fabric which is again more oriented than in shrinkage limit specimens. For low plasticity soils the difference in the fabric can lead to Wp < ws as determined in the laboratory. The volume strain of a specimen due to drying from the natural state to the shrinkage limit, V" = (Vo - Vu) /Vo, is a measure of its susceptibility to shrinkage (Vo is the initial volume, V d is the final volume after drying). Linear (one-dime1nional) shrinkage is similarly defined as Ls= (Lo-Lci)/Lo, and Ls= !Vs. This linear shrinkage can be determined by direct measurement of a longitudinal specimen hcforc and after drying.
5. 7
Water adsorption
The water adsorption capahility WA indicates the nature of the clay minerals that arc present in the tested soil. It also provides a good correlation with some soil properties, especially with the swelling potential. Laboratory determination (DIN 18132 f46m, 117]) is carried out on particles smaller than 0.4 mm using the apparatus by Ens/in and Neff(Fig. 19). The apparatus is filled with water up to the filter platen and a small amount of the oven-dried ground soil is placed on the platen. At suitable time intervals, the amount of absorbed water is measured on the graduated capillary. The water adsorption capability WA is the ratio of the final mass of absorbed water to the dry mass of the specimen. Typical values for some minerals are shown in Table 9.
Fig. 19. Water adsorption after Enslin and Neff
1.4 Properties of soils and roL'.ks and their laboratory determination
5.8
145
Compaction; moisture - density relations
Compaction is a process associated with an increase in dry density and a decrease of voids, i.e. of soil porosity. It does not imply however any change in volume of soil water. Actually, compactibility of a given soil is geometrically limited by the volume of its pore water. Soils are usually compacted either by static compression, or by dynamic load -tamping or vibrations. Which of the method would be more effective depends on the type and nature of the soil. The dry density is a convenient measure of compaction. If the dependence of dry density Pd on water content w is plotted after the soil has been compacted at different water contents, a typical compaction curve Pd = f(w) as shown in Fig. 20 is obtained. In the figure, two such compaction curves are shown, the upper one being produced by a higher compaction effort. Obviously, the dry density depends also on the compaction effort E, Pd = f(w, E). The maximum dry density max Pd for the given compaction effort occurs at an optin1un1 water content \\'opt· For E2 > E1, max PU2 > max PUl, and w0 2 < w0 1. The maximum attainable dry density max Pd increases approximately with the logarithm of the compaction effort. In Fig. 20, the curve for full saturation (zero air voids) is also shown. Its equation can be readily obtained from the mass-volume relations of the phase diagram for saturated soil: Pd = Ps/(1 + WPs/P,v). The "wet of optilntun" sides of compaction curves run roughly parallel \\'ith the saturation line. The saturation line and compaction curves do not merge, since full saturation cannot be achieved by compaction. For unsaturated soil, the dry density is a function of water content and the degree of saturation (or air voids) Pd =
p,
p,(l - lla)
l+-SrP\v
l+-
-~w~p,-.
\\'Ps
Pw
Families of curves for a constant degree of saturation and constant air voids are plotted in Fig. 21.
t
0
0
w, Wa1 Wai water corrtem w -----
Fig. 20. C...ompaction L'.urves
-
constant saturation curves
--- contours of constant air porosity Fig. 21. Contours of Sr= Const and = const
Ila
146
Paul von Soos and Jan Boh
At w < Wopt ("dry of optirnun1''), con1paction is hindered by capillary effects. The attainable dry density decreases with decreasing water content. At w -+ 0 (dry soil) however, it can increase again, depending on the nature of the soil (Fig. 20). As shown in Fig. 20, for soils at w' < w < w 11 the required dry density Pd can he achieved with a con1paction effort A< A 1. For w < w' with an increased effort A> A 1, and for w > w" it cannot he achieved hy increasing the compaction effort. Fine-grained soils get a different fabric when con1pacted wet or dry of optimum. Generally, on the wet side, pores and solids arc more uniformly distributed and the soil exhibits lower pern1eability and higher shrinkage. Dry of optin1um, the pores hct\llccn larger lumps of solids are greater and therefore have higher permeability and also higher swelling potential. Soils compacted dry of optimum are prone to subsidence on saturating. Laboratory determination of con1paction behaviour of fine-grained soils was developed by Proctor [124]. DIN 18127 [46i] gives a procedure appropriate for all soils, including coarse-grained soils. Before the test, the soil is air-dried to a water content lower than wp, i. c., lower than Wopt· Coarse-grained soils without plastic constituents are oven-dried at 105 r.ic. The choice of the mould diameter depends on the largest particle in the specimen. The smallest eligible diameter should be used. In testing coarse-grained soils, grains larger than 31.5 nlnl, or 63 nlm, are removed. When large particles are excluded, a correction for them has to be applied in computing water contents and dry densities. Details of this procedure can be found in DIN 18127 [46i] or in, for example, AASHTO D T99-90 and TlS0-90 [1]. The soil is placed in the mould in layers and each layer is subjected to the specified compaction effort. Three layers are made in the standard test and five in the so called nlodified test, where a higher con1paction effort is applied. After con1pacting the last layer, the hulk density is determined, and samples arc taken to nleasure the water content from which the dry density Pd = p/(l + w) can be calculated. The soil is then removed from the mould, re-mixed with additional water added to increase the water content and the procedure is then repeated. Using at least five measurements a smooth compaction Wpr after DIN) curve can be plotted to detern1ine the optimum lvater content (wopt and 1naxifn111n dry densiry (termed the Proctor density PPr) for a compaction energy of A1 = 0.6MNm/m3, or the modified Proctor density mod pp, for Az = 2.7MNm/m3 (in DIN 18127 [46i]).
=
5. 9 Size of voids; filters The size of soil pores is important primarily for filter design and in soil capillarity. The design of such filters is based on several criteria. For soils with a unifor1niry coefficient Cu less than 2 the filter criterion by Tcrzaghi can be used. It requires D15 of the coarser soil to be smaller than 4D 85 of the finer soil, D 1s < 4D 85 (Fig. 22). For gap-graded soils it is possible to apply the filter criterion to the idealized grading curves of the two constituent parts of the original soil in estin1ating their susceptibility to piping. The requirements for D50 and Cu of a filter for soils of 2 < Cu < 20 (filter criterion by Cistin and Ziems) are shown in Fig. 23. For soils of Cu > 20, the filter criterion can he represented hy the ratio of the pcrmcahility coefficients of the soils, which should be less than 100.
147
1.4 Properties of soils and rocks and their laboratory determination 40 v
36 32
28
"'"'I~ 2<
20 0
20 53
0002
16 12
Fig. 22. Filter criterion by Terzaghi
8
~
I ,l I
/
J /
I
v
0 0
o- Dio -
-P....,F-. !'....
v
.
v
Ii---
~r--..
14 f----
I
,._ !--,,.
,
10
6'
-=:r,._
I
/
I
' /
,, 1--- r--
4 I Fig. 23. Filter criterion by Cistin and Ziems as reported in [158]
coarser sail U _D50_18
"-
/
--~
,..__
2
UD=J--
I
I
2
4
6
81D1214161820
finer sail Ud ~ dd60
------
'°
There is no routine method for determination of absolute size of soil pores, not even in the case of the most simple single-grain fabric. Pore size can be estimated, for example, from the soil grading, by mercury intrusion, or by the capillary condensation method [96, 113]. For uniform, poorly graded soils, the pore size can reach approximately to ~ of the particle size.
i
5.10 Capillarity In a glass capillary of diameter d, water can rise to a height of he above the water table due to surlace tension Ts in the contractile skin (air-water interface) and due to the tendency of water to wet the glass surface (solid-liquid interface). Therefore: 4 · T, he= - d · Yw
·COS
a
where a is the contact angle (Fig. 24). For Yw = lOkNm- 3 , surface tension of water Ts = 0.074Nm- 1, and for soils with an approximation of a = 0, the capillary height h,[mJ "" 3 x 10-5;ct[mJ. This approximate capillary rise however may not be achieved in real soils. In natural soils, the dimensions of pores vary similarly to the capillary action shown in Fig. 25. The non-uniform openings prevent the full development of the capillary height and the water can rise only to the active capillary height hea. goven1ed by the larger capillary diameter di. On lowering the water level (on drying) however, a passive (drying) capillary height hep will develop, which may correspond to the full capillary rise for diameter d2 shown in Fig. 25. The capillary rises hca and h(..11 depend primarily on the pore size distribution. Some typical values of capillary rise hep can be found in Table 10.
148
Paul von Soos and Jan Iloh<'iC
lensio'>f pressure
+
u,..=z,,,·r . . Fig. 24. Capillary rise
Fig. 25. Active and passive capillary rise
Tuble 10. Empirical values of capillary rise hep
'l}'pe of soil
Effective particle sized..,. [mm]
Sandy gravel Medium to coarse sand Fine to inediu1n sand Silty sand
0.7 111111 0.35 mm 0.10 mm 0.045 111111 0.01 mm
Silt Clay
0.001111111
hcp[tn]
0.08 111 0.20m O.SOm
1.0111 5.0m 50 111
The negative capillary pore pressure induced by a capillary height he creates the pore water capillary suction which corresponds with the tnatric suction Ua - Uw component of the total suction in soils: (ua - Uw) = Ywho = 4T,jd, where a= 0 for water (e.g., [61]). According to the principle of effective stress, the tensile stress of the pore water acts as a compressive stress in the soil structure. With a decrease of water content it causes shrinkage and contrihutes to the shear strength. The strength increase is independent of external load and may be dealt with as a form of cohesion - capillary cohesion. In Fig. 26, line ABCD shows a schematic representation of the dependence of the capillary height on the degree of saturation for a lowering of the water table (drying path). Capillary height hep corresponds to the rise in the saturated soil helow the capillary saturation level and is 1narkedly lower than 1naxi1num capillary rise 1nax he. The air entry value (u ... - Uv,)b, the matric suction which must be exceeded for air to start entering the soil pores, is reached at hep· Above 1nax he, when the residual degree of saturation Sru was reached, the remaining soil water is retained hy surface tension at the contacts of particles, and there is practically no change in saturation and no How of water. I,ine EFG in Fig. 26 represents the process of saturating a soil from its dry state. At equilibrium, the maximum saturation (Sr < 1) reaches the height min he above the free water surface. In the lahoratory, hep and the air entry value of sands and silts can be determined using the apparatus developed by Beskow (Fig. 27). Water saturated soil is placed in the container A, which is connected with the vessel B by a flexible tuhing. Initially, the sample is flooded by lifting the water level above the base of the sa1nple. Capillary suction is then produced in the sample by slowly lowering the vessel B. As the air entry value of the soil is exceeded, air bubbles can be observed in the sample. The corresponding height difference is equal to hep· The active capillary rise hea can be estimated by another simple laboratory test. The dried soil is placed in contact with water and the capillary rise can he determined as the height at which the colour of the soil changes.
149
1.4 Properties of soils and rocks and their laboratory determination
A
j "'3
-
"'~ ~ c
soil specimen llJi~f'-'-n'iffer platen
B
--sru
water-~
n,,
max he
1il ;,
·"'
I
cl
E
Jg
""
hep= (u,ruw)b lrw
8
ffexlble tube
D 1.0
0
s,----~
Fig. 26. Capillary height and saturation
Fig. 27. Determination of hep after Beskoi·v
The matric (capillary) suction of partially saturated soil samples can be studied in a pressure plate apparatus sho\\'n schematically in Fig. 28. A sample of unsaturated soil is placed on a high air entry ceramic disk, \\'hich acts as the interface, preventing pore air from entering the measuring system in the pore-water pressure. On setting up the sample the suction in the soil induces negative pore pressure readings of the pore-\\'ater pressure transducer. To avoid any increase of suction, \\-'hich could lead to cavitation in the pore water system, a zero \\'ater pressure reading is maintained by increasing the air pressure in the cell. Once equilibrium is achieved, the matric suction in the sample is determined as ua ~ Uw (or ua, in the case of Uw = 0). In the test, the reference value for the pore-\\'ater pressure is translated from the atmospheric state to the final pressure in the cell. The procedure is therefore called the axis translation technique [77]. It is commonly used in the mechanics of unsaturated soils to avoid cavitation. The apparatus shown in Fig. 28 can also be used to determine the soil water characteristic curve, the relation bet\\-·een suction and \\-'ater content (or saturation) of the sample. From the soil water characteristic curve engineering properties of unsaturated soils, for example their strength, water permeability etc. can be estimated [61]. air pressure Pa=Ua
r~------
...
expressed pore water
- specimen
..----~
--. high air-entry disk
Fig.28. Principle of pressure plate apparatus
150
Paul von Soos and Jan Bohat
5.11 Water permeability In saturated soils, the laminar flow of water is governed by Darcy's law q l>h =k·i = k A t>l
V= -
where q is the quantity of water flowing in unit time (flow rate), A is the area through which the flow occurs, v is the apparent flow velocity, i is the hydraulic gradient, .6.h is the difference in pressure head, .6.l is the flow path, and k is coefficient ofpernzeability. Laminar flow occurs, and Darcy's law is valid, if the hydraulic gradient is lower than a critical value i < 0.1/d;. = icr [119], where dw is the effective particle size, which generally lies between D10 and D25. For a hydraulic gradient of i > icr, k is not linear (post-linear range) and decreases with increasing i. However, before the initial gradient is overcon1e, there is also a pre-linear range of flow, where k increases with i (e.g., DIN 18130 (461]). Permeability is a highly anisotropic soil property. In the case of the oriented fabric of clayey soils, a higher k may be expected in the direction parallel to the platy particles. The ratio of horizontal to vertical permeability k11/kv of apparently ho111ogeneous soils commonly reaches 2 to IO. In stratified sediments, permeability parallel to stratification (assumed to be horizontal) kh = (k1d1 + k1d2 + ... + k 0 d 0 )/d is greater than in the perpendicular direction (serial drag) kh = d/(d1 /k1 + d1/k2 + ... + d 0 /k 0 ). For water flow through unsaturated soils Darcy's law also applies, the coefficient of per1neability however is not constant. It is a variable, depending on the water content, or the matric suction. Several expressions for prediction of the coefficient of permeability of unsaturated soils ku from the niatric suction have been proposed. For example, Brooks and Corey [28] (as reported in [61]) suggested that the coefficient of water permeability of saturated soil should be taken as ku = ksat for suctions below the air entry value of the soil. For higher suctions ku = ksat((Sr - Sru)/(l - Sru)r\ where Sru is the residual degree of saturation (see Section 5.10) and b is an etnpirical constant, related to the pore size distribution of the soil. According to Ktzdi 0 = 3 [96]. Coefficients of permeability of coarse-grained soil can be measured using a constant head pern1ean1eter as shown schen1atically in Fig. 29. The water is allowed to flow through the specimen from a tank where a constant water level is maintained. 'The flow rate is nieasured. The coefficient of per1neability is then given by: k=
__9__ ~ A· t .6.hw
where A is the specimen's cross section and tis the time interval. For testing fine-grained soils with k < io- 5 ms- 1 , a constant head apparatus as shown in Fig. 30 can be adopted. The cell is sitnilar to a triaxial apparatus, where a cell pressure, top and bottom drainage and axial load can be applied. 'fhe hydraulic gradient is given by the back pressure difference between the top and botto1n ends and is controlled by suitable instrumentation, for example by a ram connected to the drainage leads at the base of the specimen (e.g. [2]). Another way of measuring the coefficient of permeability of fine-grained soils is the falling head test arrange1nent, shown in Fig. 31. The quantity of water flowing through the
151
1.4 Properties of soils and rocks and their laboratory determination
back pressure
2!!J pressure
~=="
stepping
motor--
L:'.:S.fE'E:"l'=~
ram Fig. 29. Conslanl head pennea1neter
differential pressure transducer
Fig. 30. Pern1eability tesl in a lriaxial cell
speci1nen in tin1e tis 1neasured by a standpipe of diameter a, and a. J
h,
k= - - InA· l h1
The testing procedures arc dcscrihed in, for example, DIN 18130 [461or83]. The coefficient of permeability can also be evaluated from the consolidation data obtained hy a one-dimensional compression in an oedo1neter (sec Section 6.2). Determination of the coefficient of permeability in the laboratory results in a number of difficulties. Firstly, there is the problen1 of reliability of the laboratory equip1nent. For example variations of saturation and possible occurrence of air in the pore water, or fluctuation of te1nperature, which changes the viscosity of the per1neating fluid (water). Saturation of the pore-water systen1 can he facilitated hy the application of hack pressure. 1b limit changes in viscosity, the codes of practice require that the ambient temperature during lahoratory testing is constant ±2 c'C. Also, for comparahility of the results. measured coefficients of permeability are converted to the reference temperature of 10°C using a correction factor (e.g., DIN 18130 [461]). Similarly to other laboratory tests, any porous stones should be prevented from clogging, since this nlay cause erroneous data.
p
t I
i
standpipe of cross section a Fig. 31. Falling head permeameter
152
Paul von Soos and Jan Iloh
The possibility of seepage along the rigid side walls of the cell should he minin1ised. The best solution to this is to avoid rigid walls by using flexible membranes, and a cell pressure. In this case, an effective cell pressure (the difference hetween the cell pressure and the back pressure) of at least 30 kPa is required. For pertneability n1easure1nents to be reliable, the flow into the specitnen 1nust equal the outv. ard flow (steady state flow). This condition is fulfilled if the volume of voids is constant 1
at full saturation of the specin1en (Sr= 1 and~ Yp = 0). Volun1e changes of speci1nens, sv. clllng of an ovcrconsolidatcd soil or compression of a soft normally consolidated soil, can be effectively suppressed by controlling the cell pressure and the hydraulic gradient. Finally, the 1nain probletn in the laboratory detern1ination of the coefficient of permeability is the accurate reproduction of the in-situ conditions. Laboratory specimens can hardly ever preserve the in-situ nlicrostructure of the soil, i. e., porosity, grading, fabric, stress conditions etc. Disparity in porosity can be dealt Y.'ith hy a parametric study, if required. Unfortunately, the inadequacy of small laboratory san1ples in reproducing flow direction, hcdding, fissuration, anisotropy, etc., cannot be overcome. "Where appropriate, laboratory tests should therefore be supple1nented by field experin1ents. A number of empirical approximations of the coefficient of permeability have been proposed. For coarse-grained soils they are based on particle size and/or voids ratio, in the case of fine-grained soil on consistency limits, grading and voids ratio. Table 1 J shov. s typical values of k for some soils. 1
1
Table 11. Typical values of coefficient of permeability
In unsaturated soils, the air phase can either be continuous, approximately at Sr ~ 0.85, or in the fortn of occluded air hubbies at higher saturation. With occluded air, the flow of air is limited to a diffusion process through the pore-water. The flow of air in the continuous air phase is controlled by the pressure gradient, which is usually considered as the only driving potential, or by concentration. Both Fick's and Darcy's lav.'s can then be used to describe the flow. Steady state air flow is produced by applying an air pressure gradient hctwcen the two ends of a soil clement. The air coefficient of pcrmeahility is assumed constant throughout the specimen. The a1nount of air flowing through is 1neasured under constant pressure conditions, usually at 101.3 kPa absolute, or zero gauge pressure. An equation similar to Darcy's law can be derived: dh,
Va = k a -
dy
1.4 Properties of soils and rocks and lhcir laboralory determination
153
where ka is the air coefficient of perrneability, Va is the ft.ow rate, ha is the pore-air pressure head, and dh,/dy is the pore-air pressure head gradient in they direction [28, 6J], As the matric suction in the soil increases, or the degree of saturation decreases, the air coefficient of permeability ka increases. Similarly to the water coefficient of permeability, the prediction of ka can he hascd on the pore-size distrihution and the matric suction (sec Section 5.11). According to Brooks and Corey, ka = 0 for suctions below the air entry value (Ua - Uw)b of the soil. For higher suctions ka = kJ((l -Src) 2 x (1 -Src))(l+AJ/A, where Sre is the effective degree of saturation Sr~ = (Sr - Sru)/ (1 - Sru), Sru is residual degree of saturation,~- is a pore size distrihution index and kd is the air coefficient of pcrmcahility at zero saturation Sr= 0 [61]. The air coefficient of pcrmcahility is significantly higher than the water pcrmcahility value at any water content of a soil. The main reason for the difference is the variation in the viscosity, which is (at the reference temperature of 1o:oc) ahout 70 times higher for water than for air. Assuming that the volume-mass properties of soils do not change with a change in suction, the (saturated) water coefficient of pcrmcahility would he 70 times smaller than the air coefficient of permeability of the dry soil [61 ]. A laboratory instrument for the determination of air permeability is shown in I~'ig. 32 [13].
air
specimen P1-fb
.
P
w'
water Fig. 32. Air permeability after Bicz(Jk
6
Stress-strain behaviour
6.1 General considerations 6.1.1
Representation of stresses and stress changes
ln the laboratory, the stress-strain relationships of soils are often investigated on cylindrical specimens that are loaded hy normal stresses in the axial and radial directions. During such tests the principal stress directions do not change. Due to axial symmetry two of the principal stresses are equal, 02 = 03, and a Mohr circle for two-dimensional stresses in the t : a diagram (Fig. 33a) can be used to represent the stress state in the specimen. The stress state can also be represented by the coordinates t = (u1 - 03)/2 ands= (01+03)/2 of the apex of the Mohr stress circle. In this way it is possihlc to represent each Mohr circle of stress by a single point in the planes: t (Fig. 33b). sand t represent an appropriate set of stress variables for the plane stress conditions, which arc relevant in many geotechnical situations. For three-dimensional stress conditions though, the relevant stress variables arc octahedral normal and shear stresses a 0 ct and tact, which for a common situation of axial sy1nmetry can be reduced to the mean normal principal stress p = ~(ua + 2or) and
154
Paul -..·on Soos and Jan Boh
b
,..::,o
..-IN
"
3
'1,
'
b
6'
5 ~6 '
Fig. 33. Representation of plane stress a) in -.:/o-di<1gn1m b) in 0.5(u1 - u:\)/0.5(01 + 03)-diagn1m
~"'
"' "6'"
"~
~
D.Q"1
,,.~
6
"o'"o
1 6
= iiQ"3 > 0
"o
.~
8
~
"
6"
"'\-ob
4
8cr1 =-D.0"3<0
v
2
B
1II
"
!]
"'
8
7
Fig. 34. Stress paths int : s diagram
deviatoric stress q = (oa - or), v. here aa and Or are the axial and radial stresses, respectively. In the following text ho\J..·ever, for simplicity the variables t ands are also used for axial sy1n1netry. For a change of stresses u1 and/or 03, the Mohr circle of stresses and the position of its apex change. Correspondingly, the point representing the Mohr circle in the t : s plane moves into a new position. Its trajectory is called stress path, e.g., the stress path AB shov.1 n in Fig. 33. The direction of the stress paths gives information about the relation and sense of the change in principal stresses. This is demonstrated in Fig. 34, showing 8 basic directions of principal stress changes. They divide the stress plane t : s into eight regions in v»hich more general stress paths are situated. For example, in region 1 in Fig. 34, 6.a-i > tla3 > 0, or in region 2, tlu 1 > 0 > tlo3, etc. 1
Stress changes under three-dimensional stress conditions with 02 i- a3 can also be represented by stress paths in a space \Vith coordinates u-i, u2, u3 (Fig. 35). For u2 = 03 the representation reduces to the plane situation u1 : u3,.J2.
o; I!
P(a).
' a-,
"' "
~w
?----0 11
'
'
+
P(a)JTz=OJ/
' I I I I
rT,Yl
Fig. 35. Representati{)n of stresses under rr,V7~qV7
three-dimensional conditions
ffitl=a;Y7
6.1.2 Total and effective stresses; pore pressure If an external total stress increment 6.a is applied to a saturated soil, the immediate effect is an increase in the pore-\1.:ater pressure L\.u, which can be either negative or positive, depending on the load increment and the nature (state) of the soil. This excess pore-~vater pressure tlu produces a potential in the pore-v.1atcr to seep through the soil. The seepage enables tlu to dissipate while producing an increase in effective stress tlo 1 , according to
155
l.4 Properties of soils and rocks and their laboratory detern1ination
the principle of effective stresses Lla = Llo' + Ll u. As expressed by Terzaghi, the effective stress, which represents an excess over the neutral stress u, has its basis exclusively in the solid phase of the soil, and is controlling the strength and deformation behaviour of soils. For effective stresses in unsaturated soils Bi.shop [161 suggested the relation 0 1 = (cr-ua)+ x(ua - Uw). where Ua and Uw are pore-air and pore-water pressures, and xis a parameter related to the degree of saturation. Further extensions to this theory, including the solute suction component and discussions on the stress state variables for unsaturated soils are presented, for example, by Fredlund and Rahardjo in [61]. Both total and eff'ective stress paths can be plotted in one diagram t : s, t = (01 - 03)/2, s = (01 + 03)/2 (Fig. 36). The abscissas of the individual stress points then differ by the value of n1ean stress, since t' = t and s' = s - u. Therefore a simultaneous representation of total and effective stresses (stress paths) at a single plot depicts the development of pore pressures. Development of pore pressures in a saturated soil under undrained conditions can be quantified using pore pressure coefficients after Sketnpton [138]
The pore pressure coefficient B expresses the influence of the all-round (hydrostatic) pressure increment on the change of pore pressure Llu, Llu = BLlcr3. With the degree of saturation, B increases from zero to B = 1 for saturated soils (Fig. 37); ho\\.'ever if the soil skeleton is not relatively stiff, B can be less than 1 even at full saturation. This coefficient can be determined in a triaxial apparatus by varying the cell pressure and n1easuring the response in pore pressure under undrained conditions. The pore pressure coefficient A shows the influence of the increase in shear stress (principle stress difference Ao1 - ~a3), Au = BA(Lla1 - Llt1J): A < ~ indicates a tendency to dilate at undrained shear (~ V = 0); negative excess pore-\Vater pressures are built in the soil, Llu < 0; A > ~ indicates a tendency to con1press (negative dilation) at undrained shear (AV= O); positive excess pore-water pressures are built in the soil, Ll u > 0. l11e coefficient A varies with the state of the soil (OCR, relative density) and with the strain level. The value At at failure is considered a reference value for a soil (see Fig. 38). It can be n1easured in the triaxial test. For an ideally elastic material A=~, Au= 0. w~~~-~~-~-~
W1
aa effective sir= path PfUr'. uJJ
~
=17%
=?%
-~1 -+--~~r+----; {Japt=?.18tlm 3 i---+--l--+--j
Wopt=/'.'8% ,5[]
a?l=J>J44o55 Fig. 36. Total and effective stress paths
70
75
80
85
s,-
90
95% 100
Fig. 37. Dependence of pore pressure coefficient R on sa1ura1ion Sr (after Skemp ton)
156
Paul von Soos and Jan Boh
K:i =4Jr,, 4 ::2J%
•
~o 5,~1,__+-_,
Fig. 38. Dt:pt:ndt:nct: of port: prcssurt: parameter A (Ar at failure) on strain level
24875]2
c,
and OCR
OCR
For normal three-dimensional stress conditions, the change of the pore pressure in an undrained loading condition can he expressed according to Henkel [741 ~u=B
[
~cr1 +~a2+~cr3]
3
where a is the pore pressure coefficient after Henkel. For axial symmetry A = a../2 + ~. 6.1.3
Factors inHuencing stress-strain relations
Depending on the type of soil, the stress-strain equations depend on the stress history. stress level, mobilisation of the shear strength, and on the nature of the stress changes., i.e. the direction and form of the stress paths. Moreover, the soil hehaviour is timedependent. A soil is said to he normally consolidated, if its current level of stress has not been exceeded in its stress history, and overconsolidated if it has experienced higher loads, for example if the effective geological pre-loading o~ was greater than the current effective overburden pressure o~. 0 = yz. The ratio of the yield stress o~ and the current stress is called the overconsolidation ratio, OCR=
157
1.4 Properlies of soils and rocks and their laboratory determinalion
6.2
One-dimensional compression and consolidation (oedometer) test
6.2.1 Apparatus - oedometer A soil specimen in the form of a disc is contained in a stiff metal ring so that radial strains remain zero throughout the whole test. The specimen is sandwiched between two porous discs that act as drains. Static axial load is then applied from the top via a loading cap. The upper filter plate has a slightly smaller diameter than the ring to he allowed to penetrate into the ring and compress the specimen. The vertical load and vertical deformation are hoth recorded. In the fixed ring assembly, the ring does not move with respect to the bottom porous plate (Fig. 39a). In a floating ring cell, the diameter of the bottom porous plate is smaller than the ring and both plates can penetrate the ring. This latter arrangement limits the wall friction between the soil and the ring.
In addition to the wall friction, other sources of errors in the oedometer tests are caused by the seating and bedding effects of the top and bottom plates. A height to diameter ratio of the specimens h : d ~ 1 : 5 is recommended to minimize the errors. With higher h : d ratios, the wall friction should also be measured [121 ]. The usual diameters of oedometer rings are 50, 70 or 100 mm. Specimens of coarse-grained soils, which are generally reconstituted, should not contain particles larger than 1/8 of the ring diameter
and 115 of its height.
'·
strain gauge
standpipe
free water
level
l_ lO
r
j
100
~
I
P;z
P/2
If;z
I--
70
-I
Fig. 39. Oedometer a) with fixed ring, b) with floating ring
6.2.2
Incremental loading test
This is the standard test in which the vertical loading is normally applied through a lever system by adding or removing weights. The loading is therefore stress controlled and applied in increments. After each load increment is applied, the readings of vertical deformation are recorded until the specimen is fully consolidated, usually over24 hours. Further load increments are then applied, each heing double the previous one. The number and the size of the stress steps are specified according to the type of the soil tested. The duration of every load increment throughout the test should he the same, usually 24 hours. It is recommended that the maximum stress applied in a test is at least 1.5 times higher
158
Paul von Soos and Jan Roh.iC
e
w,.u
',I
J.Dr--~~---+----+--
O
e
e,. 2.Dt-'>-;;ct--__,,'C1'~":-I
overconso/idated
DI distlirbed soil
15
e ID s'
Fig. 40n. Effective stress versus void ratio diagra1n
ti'ig. 40b. Positions of con1pression lines in e : logo' plot
than the operating stress in-situ, and in detern1ining the compression index Cc is at least 800 kPa. After the final loading inc..Tcmcnt, the specimen is unloaded and allowed to swell, again preferably in several steps. Then the cell is dismantled and the final water content of the specimen is determined. The test procedure is described in DIN 18135 [46n] and in, for example, [83 or 73 J. The 1nain purpose of this test is to detern1ine the para1neters of the soil cotnpressibility and the primary consolidation. Also, secondary compression, swelling charactcristic-s and, \\/here appropriate, preconsolidation pressure can be esti111ated. According to Fig. 40a a' + a'. e = co - Cc log -,--' o 0 + o~
and
Jcr' Jc = -C, · 0.434-o'
+ o'c
u;
where is the distance in the straight line of the gradient C, from the stress point aO on the onc·Jimensional compression line anJ s' is the relative vertical settlement during the oedometer test, wheres'= L:.H/H; = -(e - e;)/(l + e,) (Fig. 41) and the subscript i means initial. The tangent oedometer modulus is Eocd = Jcr' /de = -(1 + c)dcr' /Jc = (1- s')/ds'"' do' /ds' for smalls' (Fig. 41).
Fig. 41. Deformations in oedometer
159
1.4 Properties of soils and rocks and their laboratory determination
From the experimental test data the following secant parameters for compressibility may be determined: .8.r.v
Hi - Ht
00 v
Hi
coefficient of volume compressibility mv = -;:-; or mv = and/or (secant) oedometer niodulus
where Cv is the vertical strain, a~. is the effective vertical stress. Hi and Hr are the initial and final height of the specimen in the loading increment, o~ 1 and o~ 2 arc the effective stresses applied to the specimen in the previous and in the current increment. e is the voids ratio and ~refers to an increment along a chosen section of the experimental compressibility curve. During the first loading after deposition, without any cementation or swelling, the state of the soil travels down the normal compression line NCL fOr one-dimensional loading. In the laboratory, linear NCL is obtained for soils reconstituted at w > WL. In Fig. 40b NCL lines are shown for soils of different liquid limits [136, 141]. The compression line of a cemented soil lies above, while the line for an overconsolidated soil lies below the corresponding NCL of the reconstituted soil (sedimentation line). Both the compression lines join the NCL at high enough stresses [32, 152]. Unloading (swelling) from the normal compression line brings the specimen into an ovcrconsolidated state. On rccomprcssion, the yield point can be identified as the intersection of the swelling line with the NCL (o~ in Fig. 40a). The yield stress may correspond to the maximum past stress - preconsolidation pressure. As a result of hysteresis, there is no unique relationship between e and 0 1 during unloading and reloading from the NCL (Fig. 40a). However there is a unique stress o~ on the NCL corresponding to lhc stales, which was called lhc equivalent stress by Terzaghi [150]. The equivalent stress o~ is used as a reference point for normalising test data. Knowledge of the yield (preconsolidation) pressure is needed for predictions of settlement and the interpretation of geologic history. However, since the present voids ratio can be reached by unloading and/or by creep in the e : o' diagram, the yield point observed in the oedometer test may represent an apparent preconsolidation pressure rather than the real maximum pressure experienced by the soil in the past. There are several methods for estimating the yield point from oedometer test results (e.g., (83]). In testing overconsolidated specimens, the yield stress appears as a kink on the compression curve, which can also be identified on the compression line of cemented soils (Fig. 40b ). The development of settlement and the oedometer modulus with the effective stress for normally consolidated and overconsolidated soil is shown in Fig. 42. Passing the yield point at Bon the normal compression line is associated with a decrease in Eocd·
160
Paul von Soos and Jan Iloh
Table 12. Values of coefficienls for estimaling Eaed
of norn1ally consolidated soils after Ohde [119] Soil
v,
w,
Organic soil Clay
0.85 to l.O
Silt
3to15 5 to 20 20 to 80
Sand to sand with gravel
100 to 750
0.55 to 0.70
0.85 to 1.0
0.80 to 0.95
The oedometer modulus of normally consolidated soil increases with increasing effective vertical stress. Ohde [118] proposed an empirical relation: Eoed = VeOat
(o,,')w' _.'.:.__
the values of v, and w, after Ohde [119] are shown in Table 12.
During the one-dimensional compression in the oedometer, for any change of o~, the horizontal stress o~ changes by .6.o~ = Ko · .6.o~., where Ko is the coefficient of earth pressure at rest. 1berefore, in the oedometer generally a~ #a~, which means that there are shear stresses in the specimens and individual soil elements undergo shear distortion. On the NCL where OCR = 1, Ko for normally consolidated soils Kone can be approximated by an empirical expression Kone = 1 - sin q:/, where 1fi 1 is the angle of friction of the soil. Tlle stress path for one-dimensional compression can be seen in Fig. 43. A load increment applied at the stress point E, brings the oedometer specimen into state F and after consolidation and dissipation of excess pore pressures the stress point G on the Kone line is reached. On the Kone line, the mobilized friction angle is mob 1fi1 = arcsin(sin q//(2-sin q:/)). The ratio tan(mobcp')/ tan
Fig. 42. Oedometer modulus Eaed AD or A'D': overconsolidated CD or EF: normally consolidated
1.4 Properties of soils and rocks and their laboratory determination
161
a;-a:i -?-
1 Fig. 43. Stress paths in oedometer test
During unloading under Ko conditions, the value of Ko increases with overconsolidation since a~. decreases nlore quickly than a~. For overconsolidated soils Ko can exceed I as a~ exceeds a~. and its value can be estimated by Ko = OCR 11 • According to Mayne and Kulhawy [llOJ n = sin
Eoed = k · Yw
To estimate Cv fro111 the oedo111eter data it is convenient to plot the settlement of a single load step against square root of time (Fig. 44b) or against log t (Fig. 44c). The experimental curve is then fitted to the theoretical relationship between the degree of consolidation U1 = s/sr and the tinie factor 1' = Cvt/H~r' where Hdr is the flow path (Hdr = ~ H for a specimen of height H allov,1ed to drain from both sides). Taylor's y'(time) method makes use of the initial straight portion of the experimental curve. First the ti111e for Ut = 0.6 is deter111ined as the point \Vhere the experimental curve begins to depart from the initial straight line. From the theory of one-dimensional consolidation it can then be deduced that the gradient for U 1 = 0.9 would be 1.15 times higher than
Fig. 44. Taylor's and Casagrande's methods for determining the coefficient of consolidation, a) Theoretical curve of degree of consolidation versus logarithm of time factor b) Taylor's method c) Casagrande's method
s cl
for Ut = 0.6. From this the experimental point of U 1 = 0.9 and the corresponding time Jf90 can be identified on the experimental curve (sec Fig. 44b). Then the coefficient of consolidation Cv = T90H~r/t90 = 0.848 · H~r/t90. In the log 10 (1ime) method suggested by Casagrande, the theoretical and experimental curves arc fitted at U 1 = 0.5, when half of the consolidation is completed. In the semilogarithmic plot the theoretical initial point of U1 = 0 may be identified assuming that the first part of the curve is a parabola (sec Fig. 44c). The point representing Ut = 1, the end of primary compression, is located at the intersection of the tangent at the inflection point and at the final part of the curve (Fig. 44c). Then tso is obtained and the value of cv calculated using c,, = TsoHJr/tso = 0.197 · H~r/t50. The secondary compression index (coefficient) is the gradient of the straight line in the semilogarithmic plot of voids ratio versus the logarithm of time after the end of consolidation, Ca= 6.e/ 6. log t (Fig. 45). Ca serves as a parameter in analysing creep settlement. With decreasing OCR, Cu increases reaching its final value at OCR = 1. A plot of Ca against log OCR can be used in estimating the preconsolidation stress [116].
6.2.4
Continuous loading oedomete.r test
An oedometer test i,vith continuous loading requires the measurement or controlling of the pore pressures in the sample. The specimen is therefore drained on one side only and the pore pressures arc measured on the opposite side.
1-4 Properties of soils and rocks and their laboratory determination
e
163
Ca=-~ /ogf7lf,
log 1211, c--------'
.----"i-...
I
dei_ _ _ f-----"'"1-..
Fig. 45. Definition of secondary compression index and its dependence on OCR
Several testing techniques for continuous loading in the oedometer have hcen developed. In the constant rate of strain (CRS) and constant rate of loading (CRL) tests, the rate of strain or load increase must he low enough to produce allowable pore pressures that can be measured at the impermeable base of the specimen (e.g., [31]; ASTM D 4186 [4]). During the constant gradient (CG) test the loading rate is controlled to maintain a constant value of pore pressure. This gives a constant gradient of pore pressures since the pore pressure at the opposite side of the specimen is also constant due to the applied back pressure. In the constant pore press1ire ratio (C'PR) test the ratio of the pore pressure to the total vertical stress u/av is kept constant ("continuous loading test" after Janbu et al. [87]). The two latter techniques ((."(;and (."PR tests) require computer control during the test, for the first two (CRS and CRL tests) computer data logging is sufficient. The main advantage of continuous loading tests is the continuous definition of compressibility and consolidation parameters as functions of effective stress. Also, a shorter time is required for a test to be completed than with the incremental loading apparatus. On the other hand, it is not possible to look at secondary compression effects. The effective vertical stress u~ is not constant along the height of the specimen. Therefore the relationship a~, against c (or strain) may not be reproduced correctly, for example its non-linearity may be misrepresented.
6.2.5
Determination of swelling behaviour
The swelling of soils is an increase of volume as a result of a decrease in effective stress caused by unloading, and/or of the addition of water. The swelling of fine-grained soils due to the addition of water is controlled by physicochemical interactions between particles, which depend on particle surface forces, or by chemical changes (e.g., anhydrite - gypsum). Soils containing clay minerals that cxhihit a specific surface sufficient for the existence of unsatisfied water adsorption forces, e.g., smectite or vermiculite, are highly expansive. The effects of crystal lattice configuration, osmotic pressure and water adsorption are dealt with by Mitchell [113]. In an oedomctcr, swelling pressure can he measured on specimens that arc re-compressed to their in-situ total overburden stress and then flooded. 111e swelling pressure is deter-
164
Paul van Soos and Jan Boh
mined from the load that is necessary to maintain zero axial strains after flooding the specimen (DIN 18135 [46n, 73], DGGT E 11 [531]). Alternatively a swelling test can be performed, in which the specimen is step-wise unloaded after free access to '~:ater is allowed. The only difference fron1 the standard oedometer test is that the initial height of the specimen must be lower than the height of the ocdomctcr ring to 1naintain the Ko-condition of zero lateral strain in unloading and swelling [73].
6.3
Triaxial compression test
The triaxial compression test is the standard and most common version of all tests carried out in the triaxial apparatus. Generally, the advantage of triaxial tests is that all stresses in the specimen are kno¥.·n, and two of the principal normal stresses are equal. The stress conditions can therefore be represented by a two-di1nensional Mohr stress circle. The triaxial apparatus offers a great flexibility v.:ith respect to possible stress changes and drainage conditions. The 1nost com1non undrained and drained triaxial testing of soil strength is described in Section 7 .2. The basic features of the conventional triaxial test arc shown in Fig. 46. A cylindrical specimen (1) is enclosed in a rubber sheath (2), which is sealed against the bottom and top cap. The speci1nen is contained in a cell (3) filled with water, through which all-round stress a3 is applied. A frictionless ram passing through a bushing in the top of the cell applies a force P = A(o1 - 03), where A is the speci1nen cross section area and 01 is the vertical principal stress. A duct leading through the base pedestal allows water to drain fro111 the voids of the speci1nen and per1nits 1neasuring its volun1e by a suitable device (4 in Fig. 46). In this ¥.'ay the pore pressure can be controlled. Alternatively, drainage can be prevented and the pore pressure nleasured. The cell with the specin1en is placed inside a loading frame, which typically allows both strain and stress controlled axial loading. The change of specitnen height and the axial force are conventionally 1neasured outside the cell. A full account of conventional triaxial tests is given by Bishop and Henkel [19]. For stress-strain 1neasuren1ents however, strains and the axial force should be nleasured internally in the cc IL Further, strains should be homogeneous, i. c. friction between the soil and the top and bottom platens should be eliminated (see, e.g., [9], or DIN 18137-2 [46r]). Radial strains are typically computed from the axial and volume change. Volume change of saturated specimens is measured by the amount of water flowing in or out of the specimen.
-1.4 Properties of soils and rocks and their laboratory determination
165
B
Fig. 47. Stress path in triaxial shear tests
For direct measurement of radial strains, strain hells with displacement transducers, or proximity transducers can be used. Volume changes of partially saturated soils can be measured using local strain gauges, proximity transducers or determined from the volume change of the cell fluid (water) using double cells, etc. (for example, [6, 61, 73]). Tn a typical drained triaxial compression test the specimen is first isotropically consolidated, as shown by path OA in Fig. 47, and then brought to failure along the total stress path AB. If the loading is strain controlled, with 03 kept constant throughout the shearing stage of the test, the constant increase in axial deformation should he slow enough to enable the full drainage. The recommended strain rates can be found, for example, in DIN 18137-2 [46r] (see Table 13) or in [19, 83] etc. The triaxial apparatus enables the specimen to follow any stress path in the o1 : o3 ("triaxial") plane. Tn the stress controlled test the rate of loading should not cause any excessive increase of pore pressures either. Because of the development of excess pore pressures and differences in the effective stress paths, step-wise loading in undrained increments, shown schematically hy the effective stress path COE in Fig. 47, should be avoided. Continuous loading and computer control is therefore required in stress path testing. Full saturation, if required, can be achieved by applying a hack pressure, i.e., hy increasing the pore pressure within the soil, simultaneously balancing the cell pressure to maintain the effective stress. A continuous increase of pressures is preferred to a step-wise application. The required back pressure can be estimated by Osat = ua(l - Sr)/0.02Sr, where Ua is atmospheric pressure [19]. If the isotropic consolidation stage of the test is carried out as a continuous compression, compression parameters may be derived, e.g., the bulk modulus K, K = E/(3(1 - 2v)) = Table 13. Recommended maximum strain rates for a drained test on fine-grained soil specimens of diameter 36 mm and height 72 mm (DIN 18137, Part2 [46r])
Index of plasticity Ip [ 0/o l
Maximum rate of axial displacement [mm per minute]
up to JO 10 to 25 25 to 50 over 50
O.oJ 0.005 0.002 0.001
166
Paul von Soos snd Jsn Boh3C
loose or OCR « 1
- ¥(volume increase)
+
;v (volume decrease)
iSV-0 fecrl
Fig. 48. 'fypical stress-strain curves and volume changes in standard triaxial test (u3 = const.)
6-p' / 6.cv = 6.a?,/ 6.cv, where 6-cv is the volume strain, determined conventionally from the change of volume of pore-water. On the other hand, step-wise (incremental) isotropic loading makes it possible to look at the time-strain consolidation behaviour.
In Fig. 48 there are typical stress-strain curves of soils tested under constant cell pressure 03. The full line represents the behaviour of overconsolidated fine-grained or dense coarse-grained soil, the dashed line is for normally consolidated or loose soils. Normally consolidated and loose soils contract (compress), while overconsolidated or dense soils dilate (expand) after only a small compression (Fig. 48, bottom). In Fig. 47 there are two examples of other stress paths. The path AF in Fig. 47 is for a test with 01 constant and LTJ decreasing, in which the stress-strain diagram is qualitatively similar to Fig. 48. AG in Fig. 47 is an example of the triaxial extension test, with 01 constant and a3 increasing. In undrained tests, saturated specimens deform under constant volume (.6. V = 0, w = canst.). The effective stress paths deviate from the total paths by the value of the pore pressures which develop during the test. The stress paths of undrained tests represent contours of constant void ratio. Henkel [74], following the work by Rend11/ic [125], showed the uniqueness of the state boundary surface for fine-grained soils by plotting the contours of constant voids ratio for both drained and undrained tests, and showing that they are of the same shape (Fig. 49). With sands such correspondence between drained and undrained tests has not been observed. (~oop and Lee [39] showed that undrained tests on reconstituted specimens of quartz sand exhibited a very brittle response on shearing at elevated stresses, with a rapid unloading to lower p' than expected from drained tests. It was suggested that this feature is caused by differences in the structure of the specimens which is destroyed during the drained loading which requires large shear and volumetric strains to reach the critical state. Experimental data of high pressure tests suggest that even the behaviour of structured coarse-grained soils can be normalised and fitted into the framework of critical states, if the soil fabric and structure are accounted for [42].
167
1.4 Properties of soils and rocks and their lahoratory determination
14
14
12
12
10
10 \
08
\
I
I
05
04
04
0.2
02
0
0
0.2
a)
04
0.6 08 1.0 12
14
00
b)
as a-,·tf in
08
MN/mt
10
1.2
1.1
---
Fig. 49. Contours of constant w for drained and undrained tests on a) normally consolidated, b) overconsolidated clay
In the standard conventional triaxial cell •.vithout local measurement of strain, secant Young's modulus Eso = qso/sso, can be determined. However, the strain sso at 50 °/o of the maximum deviatoric stress q should be corrected for possible false displacements caused by bedding, seating and compliance errors. There is plenty of experimental evidence that the "initial" Young's or shear moduli measured in the standard triaxial apparatus misrepresent the soil stiffness (e.g., [41, 146, 147]). Soils have been shown to be highly non-linear and the initial (elastic) stiffness can only be measured in the so called (very) small strain range of the order of 10- 7 to10- 4 , depending on the nature of the soil (Fig. 50; after Atkinson and Sallfors [7]). However, very precise local instrumentation resolving axial strains down to 10- 6 is available even for static, continuous loading, tests. Clayron and Heyman [37] reported an accuracy of ±0.027 µm of the submergible LVDTs which were used as local gauges mounted on the triaxial specimens.
very small
small
large
Fig. 50. Schematic representation of nonlinearity of soils {shear modulus versus shear strain)
168
Paul vnn Soos and Jan Boh:iC
Bender elements f54] can be used for propagating shear waves through the specimen and determining elastic stiffnesses at very small strains.. The main limitation of the bender elements technique, the accuracy of determining arrival times of the shear waves travelling through the soil specimen, seems to have been solved [91]. The non-linearity of the stress-strain behaviour and the elastic stiffness of clays have been studied by Viggiani and Atkinson [154]. and the coarse-grained soils - quartz and biogenic carbonate sands and decomposed granite - have been dealt with by JoviCiC and Coop [90].
Georgiannou et al. [63] suggested that stiffnesses measured at static loading in the triaxial apparatus at strains down to 10- 4 are of the same order as stiffnesses measured by dynamic methods (resonant column). Further comparisons of measurements by dynamic methods (both in the laboratory and in-situ) and by local gauges at very small strains during continuous loading showed that the difference between the "static" and "dynamic" stiffnesses of soils is primarily due to the magnitude of strain and not due to any rate effects. It has been demonstrated that in the triaxial apparatus elastic moduli can only be measured with special instrumentation. Standard triaxial apparatus serves primarily for determining shear strength parameters.
6.4 Unconfined compression test According to ETCS of the ISSMGE [83], the unconfined compression test is not suitable for stiffness measurements.. Without the instrumentation referred to in Section 6.3 the non-linearity of stiffness cannot be accounted for and the test can serve only for a rough estimate of stiffness. With rocks however, it can be used for the determination of Young's modulus (DGGT El [53a]). An approximate value of unconfined compressive strength qu and undrained shear strength Cu (su) can be determined under the condition a2 = a3 = 0. The cylindrical or square specimen is placed between the platens of the load frame and axial load and axial deformations are recorded. Since no provisions are taken to prevent drainage, the test is suited only for soils of very low permeability. The axial strain rate is usually 0.5 °/u to 1 % per minute. Details can be found in DIN 18136 [460] and in [83].
6.5 Tests with the general state of stress - true triaxial test and biaxial test Due to the dependence of the mechanical behaviour of soils on stress paths, attempts have been made to develop devices with a general three-dimensional state of stress in the specimens. To date however, because of the complexity and cost of testing, their use has been limited to basic research into pre-yield behaviour and the limit states of soils. There are two kinds of apparatuses capable of producing general state of stress in soil specimens. In the torsional apparatus the specimen has the shape of a hollow cylinder and the larger principal stress rotates [76]. In the "true triaxial" devices for testing prismatic samples, the directions of the principal axes of the stress and strain tensors are constant throughout the test. Several modifications have been designed, the main difference being the method of transmitting stresses into the sample. This can be done either by flexible membranes, with homogeneous stresses assumed [98], by stiff platens producing homogeneous deformations (Fig. Sla) [69, 123], or by a combination of ftexible and rigid platens within one device [65].
169
1.4 Properties of soils Lind rocks Lind their lahoratory determination '-A
i
cross section A·A
p
• -- specimen
"
----TN,_,.,,, '""'-
l""W.f
>•o;
""x zi"
a)
b)
Fig. 51. a) An example of true trillxial apparatus; 01 > 02 > 03 h) An exllmple of plane strllin apparatus; rr1 > 02 > rr3; e2 = 0
A simplified version of the true triaxial device is represented by the bi-axial apparatus. A prismatic specimen is confined in one direction while loaded and allowed to deform in the remaining two directions ([40, 45, 70] or [64]: Fig. 51h).
6,6
Measurement of time dependent deformation
Soils exhibit time-dependent stress and strain behaviour, represented by srress relaxation and creep. The extent of the rheological hehaviour depends primarily on plasticity, activity, v..·ater content and organic content of the soil. (:Jcncrally, volumetric and distortional creep can be investigated by all of the usual devices in a soil mechanics laboratory, i.e. in the true triaxial apparatus, triaxial apparatus, oedometer, or shear hox. The device used should allow long-term measurements while maintaining constant stress, or constant strain (strain rate). Secondary compression, as measured in the oedometer, represents creep deformation under the condition of free change in the specimen volume. The process is characterized by the secondary compression index Ca (see Section 6.2 and Fig. 45). If the influence of stress anisotropy on creep behaviour is investigated, the oedometer v..·ill not he sufficient due to the constant stress ratio. Apparatuses allowing more general stress conditions will have to be used. The relationship between the secondary compression index under triaxial and the oedomcter stress conditions is (Jl - 03
Ca·triax = 2Ca-~---
,
max(a1 - a3)
where max(a1 - u3) is the stress difference at failure [101]. Creep at constant volume can be investigated in undrained tests of saturated specimens in the triaxial or in the hiaxial apparatus. The possible test arrangements are listed in Table 14.
170
Paul von Soos and Jan Bohac
Table 14. Possible test arrange1nents
Type of test
Variable held constant
Creep test Constant strain rate test Relaxation test
After Leinenkugel [105] (as reported in [66]) the influence of the strain rate on the undrained strength of normally consolidated soils can be characterized by the viscosity index Iva t'B
~Tu
Iva=~ Inc·~ En
Ivu is detern1ined in a triaxial test with a step-wise change of the axial strain rate between stages a and B, usually by a factor of 10 (Fig, 52). Oe = (01 + OJ)e is the equivalent stress corresponding to the specimen void ratio. Iva is independent of shear strength mobilisation. The strain rate dependence of the shear strength can be expressed as
i
t
=
ta ( l
+ l,
0
·
In
y:)
where ta is the shear resistance at strain rate '(a (about 1 °/o per hour in the laboratory test). It has been suggested that both undrained and drained strength can be expressed in this way. Gudehus and Leinenkugel [67], found a good correlation between Iva and the liquid li1nit, Ivu ~ 2.59(ln WL - 2.7), where WL is in o/o. In rocks, logarithmic creep E = eo +A· Jn t may be adopted with A=
(i)" = (01 2~03)"
where the exponent n is dependent on the stress level (Fig. 53) (57]. Reconunendations on creep testing of rocks. including te1nperature influence can be found in DGGT E 16 [53i]. Relaxation tests are dealt with in DGGT E 17 [53j].
I< z t tog sea.le to:= v-
1
Cp =ll° /%/hi
Fig. 52. lfiaxial tests with a stepwise change of strain rate to determine the viscosity index
0 o~~~~~~zo~~~JOHN!m
1
rrFig. 53. Dependence of the creep exponent n on normal stress
17!
-1.4 Propenies of soils and rocks and their laboratory determination
7 Determination of shear strength parameters 7.1
General aspects of strength testing
Fig. 48 shows stress-strain diagrams obtained in a shearing test. Dense coarse-grained and overconsolidated fine-grained soils dilate (expand) after a small initial con1pression and exhibit peak shear strength at relatively lov..· strains. With further straining the shear stress decreases (softening) until it reaches the critical state (ultimate state, steady state), in which it shears at constant volume and stresses - critical strength lcr at the critical voids ratio ecr [35, 133]. These soils are said to be initially on the dry side of critical (the dilative zone [113 J). Normally consolidated fine-grained soils and very loose coarse-grained soils compress (contract) during shear since they arc on the wet side of critical (the contractive zone). T11ey harden during shear and reach the critical state without any previous softening (without peak shear stress). Shear failure can take place in a relatively broad plastic zone (zone rupture), or it can be concentrated in a thin shear band v..·ith a width of about 16 dso, as shown by both experimental and numerical studies [115]. At very large displacements, residual strength r, develops in the shear planes of fine grained plastic soils [140, 151]. The drop of strength from the peak to the residual value may be quantified by the brittleness index In = (tp - i:,)/
For a lin1it state of zone rupture (hon1ogeneous strain), the principal stresses 01 and 03 can be represented by a Mohr circle in the t : o diagram (Fig. 54). The envelope to the Mohr circles drawn for the limit states under different stresses 03 can be expressed by the Mohr failure criterion. If the failure envelope is a straight line, it is called Mohr-Coulornb failure envelope 2c cos cp . ---+s1ncp 01
+ 03
The shear strength parameters cp and c arc the slope and intercept of the Coulomb (MohrCoulomb) envelope, respectively. In the t : s diagram (see Section 6.1.1) the limiting Mohr r=c+o.tanip
o;-Oj
2
a;
maxaJ =maxfu;-OjJ
.--'----
b)
Kt
,a
c.cos 'f sir11f ~tan a O'f+O'j
-2-
Fig. 54. Representation of failure points and envelopes; a) in 't: a diagram, b) int: s diagram, drained test, e) undrained test
10
172
Paul von Soos and Jan Bohil:
J,rt:0/~
l
~h
Fig. 55. Criteria for failure
-c=71
circles are represented by single points that lie on the failure line Kr. For the failure line on the t: s plot it can be easily shown that the slope a= arctan(sin q:) and the intercept is c cos 'I' (Fig. 54 ). According to the Mohr failure hypothesis the intermediate principal stress docs not have any influence on the limit state. Laboratory tests have shown that this assumption is plausible for fine-grained soils [75, 153), for sands however its validity is limited. The differences bet\\'een strength measured under axial symmetry and at 02 > 03 have been studied, c. g., by Lade and Duncan [102]. The influence of the intermediate principal stress 02 on the strength increases with the initial density of sand specimens (Fig. 56) [40]. For saturated soihi under drained loading, the shear parameters arc the effective angle of friction q/ and the effective cohesion c'. 1. The parameters depend on the stress history or density of the soil, i.e. on its current state with respect to the critical state line. The strength of soils initially on the wet side of critical is represented by the critical state friction angle cr~r (c~r = 0). On the dry side it is represented by the peak strength parameters (p~ and c~ > 0. Fig. 57 shows that on unloading the peak friction angle is lower then on reloading (cp~ 1 < q:i~2 < cp~r) \\'hile the peak cohesion intercept is greater on unloading (c~ 1 > c~ 2 > 0).
45~--.~--,~.--~~---,
44 42
f----'!.~
t~
~f--~b-+u..---±:-.......r----1
~f--""""""-+---+""-~cc--1
32
0.34 0.35 0.38
0.40 O.t.2
0.44
initial porosfty ~ -
Fig. 56. Angle of friction of a sand from
triaxial and plane strain tests
Fig. 57. Definition of 1p1 and fP~·
r
173
1.4 Properties of soils and rocks and their laboratory determination
,//
/'"~ _."
,,"
'· ~"-·~f-'~~~~~~ff
marrr
Fig. 58. Definition of peak secant friction angle rp~
Fig. 59. Definition of q::'u and Cu
2. For coarse-grained soils on the dry side of critical. the strength can be approximated by the peak secant friction angle ~~ defined according to Fig. 58. 3. The residual strength is given by the residual friction angle
r:p~ (c~
= 0).
4. According to Hvorslev [82 J. for soils on the dry side of critical, shear parameters at identical water contents my be expressed by ci:<.,. and c~ as shown in Fig. 57. Since c~ is proportional to the equivalent stress o~, it is also a function of the water content w (DIN 18137, Part I). Fig. 57 shows that 'f;_. < cp~ and c'., > c~. According to, for example [66], or DIN 18137 [46p], the strength of normally consolidated fine-grained (cohesive) soil may be expressed as:
In the DIN code, an English equivalent "angle of total shear strength" is suggested for q.:~ ("Winkel der Gesamtscherfestigkeit .. ). By definition, r:p~ corresponds rather to the critical state friction angle 0. In determining Cu (q.: 0 = 0) it is essential to carry out the test with the water content identical to the in-situ value.
80
16
,...., '
10
M
I
160
!
12
c'
/{}
' 50 ~
~t.O
~s JO ~
?O
10
8
wt =35%
~
'I
I
I
0
I
__L_L
00
6
;,, =15% <0002mm:46% IA = JJ% e, 0965 CT;,,.= t.OOkN/ml
?
J
'
c1 in%-
_L
5
J ~
·,,.
2
l Fig. 60. Mobilization of r.p' and c' in shear [131]
174
Paul von Soos and Jan Bohac
From consolidated undrained tests on saturated specimens with n1easure1nent of pore pressures (CU tests), the failure envelope can be derived in terms of both effective and total stresses. The total stress envelope depends on the soil state and on the conditions of the tests, cp 11 and Cu are not true soil parameters. Therefore cpu and Cu values cannot be used in design. The closely related ratio c 0 /o' however is useful in estimating the increase of undrained shear strength if consolidation takes place due to step-wise loading, e.g., in constructing a fill in stages. With fine-grained soils 'l!~r depends prin1arily on tnineralogy and peak values q:~ on overconsolidation. The residual friction angle q::~ depends on the percentage of bulky and of platy particles, i.e. on the contribution of particle rearrangement and rolling shear to the sliding shear of platy particles [108], and on the amount of active clay minerals It has been reported that this decreases with an increasing stress level [145] and a decreasing rate of shear [142]. The critical friction angle of coarse-grained soils q:~r ~ c.p~ depends on the shape and roughness of the particles.The peak friction angle of coarse-grained soils depends (subject to the shape and roughness of the particles) on the density, gradation and confining pressure. The influence of soil structure and creep on drained strength is discussed, e.g., by Mitchell [113]. Some typical values and ranges of the effective friction angle are given in Table 15. Table 15. The shear parameter of son1e minerals
Mineral Monttnorillonite Hite Kaolinite Mica Quartz
er~
(after Olson (120]) and {p~
7.5' to 16" 16.5' to 25.5° 24.S'~ to 31 ~ 30" to 45''
q:i~
(after Kenney (94]) {p~
4" to 10"
15" 16" to 26' 30c to 35°
The cohesion intercept c1 is primarily the result of a linear approximation of the curved failure envelopes, the 1nain reason of the curvature being overconsolidation or density, with subsidiary possible effects due to the soil structure (gradation, rearrangement, etc). True (effective) cohesion is, according to the Coulomb failure criterion, the strength at zero effective normal stressi which is very difficult to measure. The possible sources of the true cohesion may be cernentation. or electrostatic and electromagnetic attractions and primary valence bonding [113]. Cementation plays a decisive role in the true cohesion. In the absence of ce1nentation only very s1nall cohesions have been 1neasured to date- up to a few kPa, i.e., at the limit of the accuracy of soil mechanics laboratory equipment. Failure envelopes are therefore likely to be curved as the effective nor1nal stress decreases down to zero value. The apparent cohesion, which does not depend on particle ce1nentation or bonding, can be attributed to capillary stresses or to apparent mechanical forces (particle geometry and packing, e.g., locking [113]). Schmertmann and Osterberg [131] have demonstrated the difference in the mobilization of c' and q::'. Cohesion c' is usually mobilized earlier than q:i', and decreases soon after its mobilization (Fig. 60). According to DIN 18137 Part 1 [46p], the increase of effective cohesion with the increase of consolidation pressure {tnax 0 1, or tnax o~.) can be expressed by a "cohesion constant":
1.4 Properties of soils and rocks an
175
c' = max o')·cs and c' =max o~"J...c where Ac or Acs are cohesion constants for zone failure and shear band failure, respectively. The strength of unsaturated soils may be expressed according to Bishop [16] as "tf =
c'
+ ((o -
Ua)
+ ;((Ua -
Uw)) tancp'
where u3 and Uw are pore-air and pore-water pressures, (ua - uw) is matric suction, xis a parameter related to the degree of saturation. Khalili and Khabaz (97], proposed that x is given by the relationship x = tan q:{1 / tan r.p', where x = ( (ua - Uw) / (u 3 - Uw )b) 0·55 . 'lhe influence of matric suction on the shear strength within the concept of effective stress for unsaturated soils is shown in Fig. 61.
Fig. 61. Influence of suction (ua - Uw) on shear strength (after [971)
{ll,-u ... ) - - - -
Alternatively, the stength of unsaturated soils can be expressed after Fredlund and Rahardjo [61] ""Cf=
c'
+ (o-
u3 )tanq/
+ (ua -
u.,.,.)tanq:ib
1,vhere q:it, is the angle indicating the increase of shear strength with respect to matric suction. When testing the strength of unsaturated soils, the suction in specimens must be either measured or controlled during the tests, see, for example, (61 ]. For determining peak strength, failure is usually taken as the point on a plot of deviator stressq against axial strain ca giving the maximumdeviator stress, or the point of maximum stress ratio oifa3 (Figs. 54b, c and 55). DIN 18137, Part 1 [46p] gives the following criteria for determining failure in individual types of standard triaxial tests (Sections 7.2.l to 7.2.4): l. In drained tests and undrained tests evaluated in total stresses at maximum deviatoric stress q = 01 - a3. 2. In undrained tests evaluated using effective stresses
- on normally consolidated specimens at maximum stress ratio a~/a3 > 1, - on overconsolidatedsoils 1,vhen the effective stress path reaches the Mohr-Coulomb envelope. 3. In tests 1,vith shear bands when "t reaches a maximum value of ""Cf or a minimum value of "tr.
176
Paul von Soos and Jan Bohac
Measuring strength parameters of natural soils in the lahoratory is not an easy task hecause
of the general problems of reliability and representativeness of specimens and the influence of, for example, anisotropy, fissuration, scale factor etc. (see, e.g., [109]).
Generally, two basic types of tests for soil strength can be distinguished: a) Test procedures allowing free development of shear zones or shear hands. The specimen is loaded on its boundary by principal stresses that are increased up to the limit value while their directions are constant. The shear planes develop at the angle 45~' - i:p' /2 to the direction of the major principal stress. These conditions are applied in triaxial tests and in bi-axial tests with 02 > 03, etc.
a =
b) Tests with the predetermined position of the shear plane or direction of shear. The induced kinematic conditions cause rotation of the principal stress directions-the state of stress changes to suit the deformations. Typical examples are direct (translational) and torsion shear box test. and the simple shear test.
7.2
Triaxial compression test
Tests for strength without frictionless top and bottom platens (see also Section 6.3) are carried out on specimens with a height to diameter ratio h/d = 2 to 2.5. The typical
diameter of specimens of fine-grained soils is 38 mm (36 mm according to DIN 18137-2), for coarse-grained soils at least 8 times the largest particle size. In the case of using frictionless platens h/d ~ 1. Depending on the consolidation and drainage conditions, three hasic tests arrangements may he distinguished: consolidated drained test, consolidated
undrained test and unconsolidated undrained test (DIN 18137-2 [46rj). If appropriate (for example if there are not enough specimens available) multistage tests can be carried out. In such tests one specimen undergoes several stages of consolidation and shearing, the consolidation stress heing increased prior to each stage. An important issue of multistage testing is the selection of an appropriate criterion of failure. Generally, the stages should he stopped just at the verge of failure. Three stages are usually performed,
the third one with a possibility of continuing after the peak failure is reached (e.g., [73] or [83 J). Two types of loading were proposed hy Ho and Fredlund, as reported in [61 ]. In the cyclic loading procedure the specimen is unloaded to the zero deviator stress prior
to the next stage. while in the sustained loading method the deviator stress is kept at its maximum value when starting the next stage. The technique is applicahle for hoth drained and undrained tests of saturated and partially
saturated specimens. Tests with many types of soils yielded drained (effective) strength parameters that were practically identical with the results of a standard procedure using a similar set of specimens. However, when distinct shear surfaces develop in the specimen, the strength may be reduced from its peak value. Other properties, for example compressibility, dilatancy etc. may not compare as well with the results of standard tests
[73 J. 7.2.1
Consolidated drained test (CD test)
Typically, three or more specimens are initially suhjected to different all round stresses and allowed to consolidate (isotropic consolidation). After consolidation they are subjected to drained shear, during which the volume changes are measured. In the standard test
1.4 Properties of soils and rocks and their laboratory determination
177
the loading is applied by a constant rate of axial deformation. However, stress controlled loading may also be use
7.2.2
Consolidated undrained test (CU test)
After consolidation (usually isotropic), the specimens are sheared undrained, while pore pressures arc measured. With the exception of testing stiff, fissured clays, the rate of shear can be higher than in drained tests (see, e.g., [83]). For pore pressure measurement full saturation of the soil is important, which is usually achieved by using an appropriate back pressure (see Section 6.3). T11e effective stress paths JcpenJ on the OCR (Fig. 62). Both effective stress strength parameters (r' and c1 , and total stress para1neters can be obtained from ClJ tests. Although the total stress strength parameters cru and Cu have not much practical use, the test yields the undrained strength Cu at different void ratios (i.e at different normal stress levels).
a;-o;
-2-
a
0 a)
O~+cr;
u,·+u,
-2-
b)
-2-
Fig. 62. Stress paths and evaluation of shear parameters in CU tests a) normally consolidated b) overconsolidated soil
178
Paul von Soos and Jan Boh
7.2.3
Consolidated test at constant volume and pore pressure (CCV test)
This test is carried out basically as a ClJ test on a saturated specimen, the only difference being that the ce11 pressure u3 is adjusted throughout the shearing to prevent any change in the pore pressure (i.e. in the initial back pressure). Therefore, despite the drainage leads being closed, the test can be viewed as a type of a drained test (L'>u = 0). with the volun1e of the specin1en maintained at a constant value(~ V = 0). Due to the change in u3 the stress path deviates fron1 the drained stress path to follow the effective stress path of an undrained test. The effective shear parameters
7.2.4
Saturated specimens (usually three "identical"specimens, i.e., of the same \Vater content) arc set up in the cell and sheared to failure without consolidation. The drainage valve is closed and pore pressures are not measured. A high rate of axial strain can therefore be adopted, typically in the range 0.5 °/o to 2 o/o per minute. Jn this test, any increment of the all-round pressure is taken hy an increase in the pore-water pressure of the saturated specimen and the effective stress remains equal to the initial negative pore pressure. Since the strength depends only on effective stress. Mohr stress circles of the same diameter are obtained for all the s.pecimens regardless of the applied cell pressure (Fig. 59). Without cavitation therefore 41u = O and Cu = nlax(u1 - a3). For sensitive soils Bjerrum [21] suggested that anisotropic consolidation by in-situ vertical stresses ovo and Kouvo should be performed prior to lJlJ tests.
!
7.2.5
Triaxial compression tests on rock
Depending on the type of rock to he tested, cell pressures of 03 up to 60 MPa and axial stresses 01 up to 600 MPa are necessary. The minimum diameter of the specin1en should be 30 nlnl. or 10 times the size of the largest particle. Jn the standard test, after isotropic consolidation, the specimen is sheared by strain-controlled loading. The maximum rate of axial strain loading should correspond to a strain rate that \\'ould bring an identical specimen to failure in an unconfined compression test in not less than 5 minutes (DGGT E2 [53bj). Data of at least three tests should be evaluated in lolal stresses, the failure criterion being the nlaxin1un1 deviator stress 01 - 03. To detern1ine the residual strength or shear resistance on slant joints, the top and bottom platens should he free to nlove sideways to ensure kinen1atic compatibility. Jn order to gain more information from an individual specimen, Koviiri and Tisa [100] suggested a modified test procedure. Shortly before reaching failure the cell pressure 03 should be continuously increased by such a difference that the stress-strain line changes its direction from a1 to a2. az < a1 (points A an B in Fig. 63, left). The stress path therefore runs from A to B close to the failure line (Fig. 63, right). With 03 kept constant the specimen is then sheared farther, up to the residual state - point C in Fig. 63) and finally unloaded along the residual strength envelope by decreasing 03. From Fig. 63 qi= arcsin((m-1)/(m + 1)) and c = k/(2,/iii), where m = L'>ui/ L'.u3 and k is the intercept on the u1 axis. By testing large diameter specimens (600 mm) of densely fissured rock in the triaxial apparatus. the strength of the rock mass can also be determined. Multistage testing technique
1.4 Properties of soils and rocks snd their laboratory detern1ination
8
'a,
A
a,
J{}[)
c
' \ , __
179
8
2(f) l(f)
D
5 2 J axial strain c, {%}
'
-5-
00 20 40 50 cell pressure 17i{/'1N!m 1J
Fig. 63. TTiaxial tests on rock (sfter [1001)
described in Section 7.2.2 can be used (see DGGT E 3 [53cj). The main condition for the use of this technique is that the rock docs not exhibit hrittlc failure (DGGT E 12 [53gl).
7.3
Detennination of' unconfined compressive strength and sensitivity
The unconfined con1pressive strength Qu is the maximum stress 01 sustained by an axially compressed cylindrical specimen at the radial stress of 03 = 0. To n1inin1ize the influence of end friction, the height to diameter ratio of cylindrical specimens should be 1.8 to 2.5 for square specimens 2 to 2.8. The results nlay be used to derive the undrained shear strength Cu of soils. However. during the test no provision is made to prevent drainage and therefore the derived value of the unconfined strength is only valid for soils of low permeability that can be considered to hchavc as undrained throughout the test. The specimens must be tested at the in-situ \Vater content, other\vise the results are not relevant to the in-situ strength. Therefore, the method should not he applied for fissured or varvcd clays, silts and peats [83]. The test also serves for detern1ining the sensitivity St of clays: firstly Qu is found for undisturbed specimen, then the soil is remoulded without any change of the original water content, and the test is repeated to obtain the strength of the remoulded soil Qur· The soil sensitivity is then given by S1 = q 11 /Qur· The sensitivity can also be nleasured using the lJlJ test, giving nlore reliable results due to provisions against drainage, or by the fall-cone apparatus (see Section 5.6 and Fig. 17). The sensitivity of fresh \Vater sediments is usually small (1 to 2). intermediate (2 to 4) and seldom high (4 to 8). For leached marine sediments h(J\Vever it can exceed lCXl (Fig. 64) {20, 1431. As a result of thixotropy, soils with sn1all to intermediate sensitivity regain a substantial part of their strength \Vilh time. The unconfined compression test on rock yields Ou, the unconfined strength (sec DGGT E 1 [53al). Slender specimens should be tested, h/d = 2 to 2.5. otherwise the measured strength must he reduced by multiplying hy 8/(7 + 2d/h). The unconfined strength of rock au can be also estimated from the strength index ls derived from the point-load test, au = 24 T5 • In this test, rock core specimens of at least 50 nlm diameter are loaded axially or radially to failure between two conical platens (Fig. 65). Irregular lumps can also be tested. The failure load F and the initial distance between the platens a are measured and the strength index Is = F /a 2 calculated. Correction factors arc applied to account for specimen size and shape ([27]. DGGT E 5 [53d]).
180
Paul von Soos and Jan Bohac F
s,
U=50° r =5mm
10'
~
' "\
+-
I
-lo'
'.,, .
\~ ~
=,
~
-JD
'"'
111
~-
-
-20 -10 1,
D
Skempton 7952 Bjerrum 1954 \\ Sr = sensitivdy 0
"WO
10
+
20
JD
salt concentratioo in kg/mJ
Fig. 64. Dependence of sensitivity on ~alt concentration
7.4 7.4.1
Fig. 65. Point load test
Shear box test Direct (translational) shear box test
A cylindrical or square specimen of 60 mm minimum diameter (or 60 x 60 mm) is contained in a split box, sandwiched hct\vccn two porous filter plates (Fig. 66). The specimen is firstly consolidated one-dimensionally at zero horizontal strain and then sheared along a horizontal plane defined by the halves of the hox. The normal stress is applied by weights (load Pin Fig. 66) and the shear force Sis usually produced by a constant rate of horizontal displacement. During the test the hox shaJl he submerged in water to prevent capillary forces T11e design of the top platen and of the movable half of the box should prevent any effect of tilting or rotation of the top platen and any \Vall friction on the vertical face (DIN 18137-3 [46s, 83] or [156]). To measure the drained strength the rate of shear displacement must be slow enough to prevent any huild up of excess pore pressures in the specimen [82). The n1easurcd vertical and horizontal stresses arc then taken to be effective. From the volume changes, the angle of dilation can he estimated [156]. Tests under constant volume are possible by changing the vertical load to n1aintain zero vertical displacement during the shearing stage [148],
1.4 Properties of soils and rocks and their laboratory detern1ination
181
(231- The failure states of individual specimens represented by normal stress a' and shear stress tf arc then plotted in a t: : a' diagram. The best fit straight line yields the shear strength parameters qi' and c'. The recognised problem of the actual direction of a slip plane in the direct shear test, and its influence on the experimental results \Vas dealt \Vith by Hansen f72] and Rowe f129]. This sho\\'ed that the direct shear strength of loose quartz sand was conservative, hile there \\'as practically no significant difference between triaxial and direct shear box strength of dense quartz sand. 1 \\
The residual strength can be measured in the translational direct shear box test by reversing the direction of shearing (23]. The residual state is achieved when there is no drop in shear resistance after repeated reversals (DIN 18137-3 (46s, 73]). The main disadvantages of the direct shear box are that there is no possibility of controlling the drainage and that the principal stresses arc not kno\\'n (Mohr circles). At every point in the specimen and along the shear plane there is a different state of stress with different principal stresses and principal planes. On applying the shear force, principal stresses rotate. In evaluating the peak strength, the soil is assumed to have reached the condition of plasticity and is obeying Coulomb's lav·.: at all points of the shear plane. This requiren1ent is obviously not fulfilled. The stress strain curves from direct shear box tests are also not reliable [130]. Despite the v-.·eaknesses the test is \\ idely used in practice. 1
In rock mechanics the shear box is used mainly to determine the friction in cracks and joints, often using the multi-stage lest technique (DGGT E 12 and E 13 [53g. 53h, 1261).
7.4.2
Ring shear test
In this test an annular specimen is placed and consolidated in a split box similarly as in the translational direct shear apparatus. However the shear force is produced by torsion (Fig. 67) (82, 151]. The main advantage of this test against the translational direct shear box is the constant size of the contact (shear) surface bet\\'een the upper and lo\\'er half of the specimen at any value of shear displacement. This makes the apparatus expedient in determining the residual strength (18]. Data processing is similar to the direct (translational) shear box test.
A cylindrical or square specimen is loaded by a vertical normal stress and sheared horizontally \Vhilst the shear strain is uniforn1 along the height of the specin1en (see Fig. 68) [128]. A different, easier to use design of the simple shear apparatus enables the sides of the speci1nen to rotate by sealing the soil in a rubber sheath reinforced by a spiral wire [22]. With this apparatus, pore pressure measurement is also possible and the tests can be either drained or undrained. p
piston
Fig. 68. Simple shear
During the test, as in the direct shear box, the volume of the specimen is allowed to change. Due to the plane strain conditions, the angle of dilation can be computed from the volumetric and shear strains ~ITIV
LI. V = --- = Ll.·1
(e1
+ e3)
(£1 - £3)
The specin1en is supposed to fail in simple shear along a horizontal plane (parallel to the top and botton1 platens). The shear and normal stresses on the plane are then measured while principal stresses are unknown. The shear parameters er~ and c~ arc relevant to plane strain conditions.
8
Determination of tensile strength
1'ensile strength is associated \Vith cohesion regardless of its physical nature. Soils with capillary forces, like stiff clays and marls, or compacted soils, can apparently exhibit tensile strength \Vhen simple tests, like unconfined co1npression, are evaluated in total stress tern1s. Generally, tensile strength of rock or ce1nented soils (including stabilised soils) can be tested by the Brazilian strength test ([95], DGGT E 10 [53e ]. This test belongs to the indirect tensile strength tests. Combined tension and compression is generated using loading applied by rigid platens diamctrally to a core specimen of h : d ~ 1 (Fig. 69).
Fig. 69. Brazilian test
1.4 Properties of soils and rocks and their lahoratnry determination
183
Strain rates arc si111ilar to the unconfined compression test (see Section 7.3). The Brazilian tensile strength is then given by Oz = 2Pq/(rrhd). Tt approximates well the unconfined tensile strength of rocks [84).
9
Detennination of slake durability of rock
All rocks are to so1ne extent affected by wetting and drying, and rocks containing clay minerals, e.g., shales and some igneous rocks, can disintegrate \\'hen exposed to wetting and drying cycles. Slake durability, defined as the resistance of a rock to ,,,,etting and drying cycles, can be tested by a standardized test [60). Ten lumps of rock, each weighing from 40 g to 60 g, arc oven-dried at 105° (1nass tnu 0), placed in a sieve mesh drum and in1n1ersed in water (Fig. 70). During 10 minutes of slow rotation the specitnens are allowed to disintegrate while fragments leave the drun1 through the mesh of 2 mm aperture. After re-drying (1nass nldt) the second cycle of slaking is applied and the final 1nass determined (ma2). The second cycle slake durahility index Ia2 = md2/mdo serves as a reference for rock classification. For lct2 < 0.1, the results of the first cycle nlay be considered, for Ia2 > 0.6 it is useful to carry our additional slaking cycles. Low values of Ia2 indicate a material (soil) highly susceptible to slaking, for more rocklike materials lct2 approaches 1 (100 1X, ).
r-
100 mm
1
water -
~level
sieve m~ drum
=
IC ---------
10
Fig. 70. Slake durability test
Correlations
Reliably measuring parameters for the stress-strain behaviour and strength of soils often requires expensive laboratory tests on good 4uality, usually undisturbed. satnples. This is generally more difficult and 1nore costly than carrying simple descriptive tests, for example grain size distribution, Atterbcrg litnits, \\ ater content etc. To facilitate cost effective 1nanagement of site investigation and sampling, and to estimate soil and rock properties in preliminary investigations, correlations tnay therefore be used between the simple tests and the para1neters of soil behaviour. 1
10.1
Proctor density and optimum water content of 6ne-grained soils
a) pp,= f(WL, lp); Wopt
= Wp,
= f(pp,)
See Fig. 71: these are shown by the arrow in the figure [55).
184
Paul von Soos and Jan Boh
50 F.F~;;f"-~
! ,or--+"'
''£ 30~....t=c-
Fig. 71. Proctor n1axin1un1 density and optin1um 0\ 13
I
water content as a function of Atterberg lirnits
'
I'
15
16
17 18 19 max gl"r (proctor standard) [t Im'}---
20
b) pp,= 2.29 - 0.887WL - l.165wp + 1.360wL. Wp -0.1441'>a,;[l Wop! = \Vpr = 0.0763 + 0.237\VL, or Wopt =° Wp, = 0.0446 + 0.62wp WL, Wp and the silt fraction Llasilt are sho\vn as decimal fractions, PPr in t/m 3 and \Vopt arc decimals [106]_
10.2 Water pcnncability 10.2.1
Coarse-grained soils
For 0.06 < d10 < 0.6 mm and I < C 0 = doo/d10 < 20 k = [A/(Cu
+ B) + CldTo
where for d10 in cm, k is in m/s and the coefficients depend on density (data from Beyer [12J):
A~ B~
c~
Loose
Medium
Dense
3.49 4.40 0.80
2.68 3.40 0.55
2.34 3.40 0.39
10.2.2 Fine-grained soils a) k = 0.0174 · {[c - 0.027 x (wp - 0.242 x Ip )J/lp )4·20 /(1
+ e)
For wp and Ir in °/ci and for soils of e > 0.8, the equation gives kin mis l33 J. b) k = 6.5 x 10 11 . [c/a x (IA+ 1)] 4 For a (the percentage of day particles < 0.002 mm) and IA (the index of activity) expressed as a decimal fraction the equation gives kin n1is for soils of e > 0.8 [112].
1.4 Properties of soils and rocks
185
10.3 Stress-strain relations for soils 10.3.1
Compression index
a) C, = N · (WL - 10) For WL in% and N = 0.007 for disturbed soil and N
= 0.009 for undisturbed soil [136].
b) C, = 0.5 ·Ip · p, For Ip as a decimal fraction and p, in g/cm 3 fl 60]. c) C,
= f(w).
sec Fig. 72 [99, 103]
d) C, = 0.0126 · Wn - 0.162 For Wn in% [106]. e) C, = 0.496. c; - 0.195 Where e; = initial voids ratio [106]. 0.60 0.50
2
~
~ ~
0.40
0.30
·""
~- 0.20 l---+-+---1~~
! tJ>
jf
0 10
0
e-~---10
••
20
Fig. 72. Compn:ssion index Cc as a function of liquid limit wL J{J
40 50
5[} 1(} /J[)
100
lOfJ
JOO 400
water content % ---
10.3.2
Coefficient of' consolidation
c, = f(wr.) Sec Fig. 73 after U.S. Navy 1962, as reported in [103].
Fig. 73. Coefficient of consolidation c,. as a function of liquid lin1it WL
186
Paul von Soos and Jan BohiiC
10.3.3
Index of secondary compression
a) Ca= 0.00018. w For win % [135]. b) Ca= a· C, where a= 0.04 ± 0.01 for inorganic clays., a = 0.05 ± 0.01 for organic clays [ 111 J.
10.3.4
Parameters of the equation f'or oedometer modulus
Eol!d
=Ve Oat
(~, )'" Oa1
a) For coarse-grained (cohesionless) soils at 0.1 < def [in mm] <9 nin1, 1 < Cu < 8 and 0.66 < r < 1.0 (r is roughness coefficient, see Section 4.4 and [89]). • Normally consolidated soil 390r'd0.23 Ve = C 0. 896 e;t_ 55 for def < 0.6 mm; and Ve u
2.8r4.8Jd0.87 /r1.1 = 1ef
V.'c
c~·64/r2 el.15/r2.76
=
JOOr'·sd-0.345 for def > 0.6 mm (' r0_81 ;;_ 55 ·u
731 272
0.9r 736 d- 01 · - ,r for def< 0.3 mm(ll; and
V.'c
= 1-
ef
2 ~6
C8-74r 158 ·· el.15/r ·'
for def > 0.3 mm(lJ (lJ
for roughness coefficient r > 0.85 the correlation is valid for Cu> 1.9 and e > 0. 7 only.
• Overconsolidated soil v _ s -
lOOOr-o. 95 do ..47ro.r-.::i ct C 0.545r0.54 el.68ro.ss
450r-l.54ct-.0.125r11.11
for def < 0.33 mm: and vt> =
u
ct C 3.sel.68r0.68 for def > 0.33 mm
Wt,=
1-
0 16r0.43d0)58rl.05
, et C 0.837r0.64 e2.48 ·u
for der < 0.7mm,, an d \Vs= ·1 -
0·.68 0 ·15r 0 .4ctef
C ~-111r0446 e2.48
for def > 0.7 nim b) for fine-grained soils 758 · v - A/IB vc-- ... 5 5/1°· p 'sp where A= 11: 14: or 20, and B = 0.88: 0.96: or 1.04, for stiff, firm or soft consistency, respectively (Jiinke in [89] after Ohde [119]).
c) for any soil w, = f(n). see Fig. 74 Ve= f(n). see Fig. 75 v, = f(w), see Fig. 76 [86].
187
1.4 Properties of soils and rocks and their laboratory determination
a rock ~ morain 10 1\----'l------+----j o sand ::=.~.::i T Silt •clay IO'l---\--'\----+--~'~--~--j 5
=
sand (Moussa 1961)
o~~.,.-,--.~~~~-,-~~~
\ \
~
\
~ 05>------' ~
I
10
\
silt
'
\
(Kotzias 1963)
\
clays
so Fig. 74. Exponent
v'1---~~'l
morains ~ sand o silts +
\
rock
,.
crushed rock
We
10
100%
as a function of porosity n
50
100%
Fig. 75. Compressibility coefficient function of porosity n
JIJ ,-"'"'~---,--m-e-an-"~ac-/ue • maximum__, T minimum r-- l
HOO
Ve
as a
j'.:::2,'.:j:~-k-+-+-+-++-l-W
1200 k-'..L+;c'.A-c:,-'.~--l--l1000
I
'>+'<"-"'. >d,,"~20"- 50 kN!m 2
o~ 800
-C, w
600
f"'-c-:+L.c:f".-"c'?fcLc'oiS>,1-
i.:.-:.-1-..:.lb..:..-l-"-,L+4''.l>l-ct--l-l--I
4006~m
200
0
JO
,g
50
60
70%
15
2
3
~
5
6 7 8 9 10
w
OCR
Fig. 76. Compressibility coefficient Ve as a function of water content w
t'ig. 77. Dcpcn
10.3.5
Undrained stiffness Eu
Eu/Cu dependent on the overconsolidation ratio (OCR) and the plasticity index (Ir) is shown in Fig. 77 (after Duncan and Buchignani, as reported in [38]).
10.4 10.4..1
Parameters of shear strength Peak ellcctivc angle of friction of coarse-grained soils
+ 0.005 where ei is the initial voids ratio [134]
a) cot qi' = 3.36 e;
q:i'
188
Paul von Soos and Jan Roh
b) cotq/=a ·ei+b where a= 2. Hl5 + 0.097ds5/d15 b = 0.845 - 0.398a [149]. c) sinqi'=k/(l+c) where k = 1.5 ~ 0.475 ·log Cu for crushed soils. k = 1.1 - 0.35 ·log Cu for natural soils [114]. 1 08 -0.0697/r1UJ45
d)
'
tan cp =
0.65r . dd
+ 0.6r-0.225(1.2 - c)
C0.157/r.3.JS
"
This relation is only valid for the limits given in 10.3.4a [89]. 10.4.2
Effective ang]e of friction of nonnally consolidated fine-grained soils
lfl~r
qi;,= f(lp) sec Fig. 78 [101]. q{, = 0.81 - 0.223 log Ir where the plasticity index Ir is in o/o [93].
Naniocks OM-7 !197fJ: mean value± 1 standard dev.
•O
--L:: --=.,,_ -
o..
",
I '
Kenney (7959) o Bjerrum and Simons (1960) o Bferrum (7 963) o
10 -
5
10.4.3
00
_____q__
10
20
Ip fin%)
50
}'ig. 78. Dependence of cp~ (= i:p~r) on plasticity index (after [101])
700
Effective engle of' friction of overconsoJidated fine-grained soils cp'
tanqi' = [(tanqi;,)/0.8611. 608 [56] 10.4.4
Effective angle of friction of fine-grained soi1s after Hvorslev qi~
a) tan qi~= tanqi;,(1.07 -0.04651A) where IA is expressed as a decimal fraction. This is only valid for 0.2
qi~=
11.76 -14.7 log Ip± 3.37 where the plasticity index Ip is expressed as a decimal fraction [801.
189
1.4 Propenies of soils and rocks and their laboratory determinaLion
10.4.5 a)
Strength of undrained soils Cu
(0.11+0.37 lp) · o~ For norn1ally consolidated soils, with
Cu=
b) log Cu = 3.23 - 3.44w where w is a decimal fraction and
c) Cu= CTat
[
0.163 +
Cu=
e)
Cu=
as a deci1nal fraction [139].
is in kPa [134]. 6.33
0.166 37.le- w p lp(4.14 + l/lA)
For soils of c > 0.8, Wp, Ip in cyo,
d)
Cu
Ip
IA
l
as a decimal fraction and
Cu
in the units of Oat [33].
M exp[-4.6(lc -1)] For normally consolidated young (not aged) sediments M = 170. For fully remoulded sediments M = 52.6. For le as a decimal fraction: Cu in kPa [160]yz(0.23 ± 0.04)0CR"· 8 For soils of Ip < 60 °/o and relatively low overconsolidation 185].
11 Classification 11.1
Soil classification
Soil classification aims to categorise soils according to their probable engineering behaviour into classes or groups. The divisions arc carried out on the hasis of standard index tests and each group is defined by given limits of appropriate characteristics. Besides offering access to previous experience, classification should also provide a common language for communication. Many soil classifications exist that are problem-oriented and designed for different fields of activities. 111e early syste1ns that were developed for road, highway and aviation gcotcchnical purposes focused mainly on pavement design.The classification attributes were particle size distribution (disregarding cobbles and boulders), Attcrhcrg limits and partly organic content. The density of coarse-grained soils and consistency of fine-grained soils were neglected. A soil classification developed for the lJS Bureau of Public Works by Hugentugler and "/"erzaghi [78] was later modified and has been in standard use as an ASTM Designation [3]. By using sieves No JO (2.0mm), No 40 (0.42mm) and No 200 (0.074mm) coarse-grained particles arc divided into three main groups Al to A3. Fine-grained soils, containing inore than 36 o/o particles s1naller than 0.074 mm, arc divided into four main groups A4 to A 7. The groups Al, A2 and A7 arc further subdivided into subgroups. The ·'A-classification" is used inainly in the USA in road engineering, and more especially in the field of soil stahilisation. 111e ·'Airfield classification system" formulated by Casagrande [36] was adopted hy the lJ.S. Corps of Engineers. lt introduced 15 soil groups, each of which was assigned a na1ne consisting of two capital letters. 111e first letter expresses the type of soil, G (gravel),
190
Paul von Soos and Jan Boh
50
40
I
30
·=....,.
20
...,_
10 7
4
a
0
10
20
30
40
50
60
70
80
w,in% Fig. 79. Plasticity chart afler Casa.grande, wilh group sy1nbols after DIN
S (sand). M (silt). C (clay), 0 (organic soil, Pl for peat). the second capital letter represents the grading of coarse-grained soils (W for well graded, P for poorly graded) or an admixture of clay (C) or of fines (F). The groups of fine-grained soils where then designated according to their position in the Plasticity chart (Fig. 79). Soils that plot above the A-line are clays (C), while silts (M) plot below the A-line, H (the second letter) expresses high soil plasticity if WL > 50 o/o, L denotes low plasticity for WL < 50 o/ci. The system was further developed into the "Unified Soil Classification System" (USCS) in 1952 by A. Casagrande and the U.S. Bureau of Reclamation ([30, 155]; Table 16. after [79]). The four major groups are coarse-grained, fine-grained, organic soils and peat. The classification is carried out on soil passing a 75 mm sieve. The coarser fraction is documented and if appropriate subdivided into boulders (>300 mm) and cohhlcs (75 to 300 mm). Coarse-grained soils arc those having 50 o/o or 1nore 1naterial greater than 0.075 mm (No 200 sieve). They are then subdivided into sand and gravel, the limit hcing 4.75 mm. Fine-grained soils have more than 50 % passing the No 200 sieve (0.075 mm). Gradation symhols are W for well graded soil (sand with uniformity coefficient Cu > 6 or gravel at C.u > 4, and with the coefficient of curvature Cc fron1 1to3) and P for poorly graded when not n1eeting the criteria for W. The coarse-grained soils arc called clean at fines < 5 Lo 12 °/c, require dual sy1nbols, for example GW-GC, SP-SM etc. The German standard has been based on the USCS (DIN 18196, sec Tahlc 17). Since the limits hctwccn fine-grained and coarse-grained soils, and between sand and gravel were changed to the metric scale (0.06 and 2.0 mm, respectively), the limit for distinguishing hctwccn fine-grained and coarse-grained soils has changed to 40 1Yo. This yields almost identical grading in both systems. Some symbols for soil groups were 1nodified to correspond to the German words (C was replaced by T, M by U, P in poorly graded by E). Low (L), intermediate (M) and high (A) plasticity of fine-grained soils is distinguished (Fig. 79). The term "tnixed grained soil" for soils with 5 °/o to 40 °/o of fines was introduced,
1.4 Properties of soils and rocks
~nd
191
their laboratory determination
Table 16. Unified soil classification system (USA)
Soil characteristics
Laboratory classification (based on grain-size curve lo quantify fractions of soil)
Co&1rse to composite soils with calcareous or silleeous componern:s
OK
Not decomposed to moderately deo;omposed pel!IH
HN
Decomposed peats
HZ
' 0' 0
~·
"
Clays with organic
- - ~~ ,-
f-
Silts with organic componenu, and organo99nic3) fl.lits
7%,
aod below
W[__
~
~
"
0
~
-
-
£
•'e 0
-
~
~
•
~
~
27
~
E 0 u
~
, ea'
28
-
-
1i• ,. "
Mud, eollactlll'a term comprising putrid mud, peaty mud, QVttJa, dy and sapropel
F
Mede ground of natural materlal (with relewnt group symbol given .In square brackets)
[]
Made ground of altered or artlflclal msterlal
A
-
11 For guldence only, column• 10 to 21 provide a general information on engineering properties and the 8UltabllltY of soils lwlth examples gl'llllln In column 9). 21 The nroke above the ...,mboll U and T, or the asterisk may be omlned. 3j Soils formed wlth the cooperation of orgenlsms. Kev to symbols used In columns 10 to 21 Column 11
Column 10
--
I
Very low
-
I
Low
--
I Very low
-
I Low
Columns 12 to 15
--
I very high
I High
Columns 16 'to 21
--
1 Unsultabla
I Let1 1ultabla
Bohii~
195
1.4 Properties of soils and rocks and their laboratory determination
Containing plant materiel, mostly dark colour, musty Odour, loss on ignhlon up to about 20% by m . .
~11
Containing mineral components, mostly light colour, of low density and high porosity
C11lcareou• sand, tufa sand, bog lime
Topsoil, fostll
o:\~fono~ ~~c~~Organic materiel grown at site
fibrous( rich in wood, lght brown to brown
~~~1~n°~ ~~fQ."
dark brown to black
Depotlted by weter, conslnJng of plant remains, faecal pellet$, mlcroorganisms, oft9T'I contalnlng &and, clay, llme; blueish black or greenish to yellowillh brown; elastic, soft/spongy
-o
++
-0
-0
0
-- -- --
++
0
+0
-0
-0
+O
-0
--
~e
++
-0
-0
-- -- --
--
22
0
--
23
-0
--
"
-- -- --
--
25
0
-0
-0
0
+O
-0
+
0
-0
-0
0
+O
0
+O
Low moor peat, rail8d bog peat, forestswamp peat
Peaty mud, putrld mud
co.
+O
-0
-0
"
"
15
+O
H
20
1B
14
-
--
--
--
-0
0
+O
-- -- --
--
--
26
+O
-- -- --
--
--
27
2B
..
-
Domestic waste, slag, rubble, Industrial
wotte Key to aymbo1s ueed in columns 10 to 21
Column 11
Column 10
Columns 12 to 15
-0
Moderate
-o
Modereite
-0
0
Medium
0
Medium
0
High to medium
+O
Fair to medium
+O +
++
Hl;h Very high
+ ++
Fair Excellent
+O + ++
Medium
Columns 16 to 21 Moderately suitable Fairly sulteible 0
Low to med lum
+O
H lgh to medlu m
Vary low Extremely low
-o
+ ++
Suitable Very suitable HlghlV 1ult8ble
!96
Paul von Soos Hnd Jan Boh3.C
with syn1bols GU, GT, SU or ST. The content of fines over 15 o/o is indicated by a prime or star (Table 17). This made it possible to avoid the dual symbols or the USCS. Coarsegraincd and mixed-grained soils qualify as well graded (W) if Cu > 6 and I < Cc < 3, and uniformly graded (E) for Cu _::: 6 regardless of C.:. A gradation class "intertnittent (or poorly) graded'' with its symbol Twas added for soils of Cu 2: 6 and Ci.; either <1 or >3. Organogenic and organic soils are distinguished and subdivided into 4 and 6 subgroups, respectively. Fills are introduced as a separate group in which a distinction is made between fills or natural soils (with the corresponding symbol in square brackets, for example [CT]), and other fills (designation A). For individual soi I groups, engineering properties and the possibility of the use in practice arc also included in Table 17 (columns 10 to 20). The classification in DIN 18196 [51] is therefore frequently utilised by other German codes of practice_ For exan1ple the Classification offrost smceptibility (ZTVE-StB-94 [29]) distinguishes the following classes on the basis of the DTNl8196 designations: • Fl - not susceptible; all coarse-grained soils according to DIN 18196. • F2 - low to medium frost susceptibility; TA, OT, OH, OK. Mixed-grained soil groups with a sn1all percentage of fines belong to the F2 classification at 5 o/o of fines if Cu > 6, or at 15 o/o of fines if Cu> 15, otherwise they belong to Fl. • F3 - high frost susceptibility; all other soil groups according to DIN 18196. Jn Britain, intermediate plasticity (I) for 35 % < WL < 50 o/c has been added to the original Casagrandc's Plasticity Chart ([127]), and they are other similarities with the DIN standard for soil classification. One of the main differences however is the designation of composite soils. Tn the British code the classes of fine-grained soils begin on exceeding 35 o/o by the mass of particles smaller than 0.06 mm; coarse grained soils nlay have up to 35 90 %, respectively. Recently the need to internationally unify the soil classification systems has led to formulating TSO 14688-2 "Geo technical Engineering - Identification and classification of soils - Part 2 Classification". Many specialised standards that are widely used in practice use further attributes in their classification systems. For example DIN 18300 "Earthworks" l52a] makes use of soil workability to designate 7 classes: Class 1: topsoil Class 2: slurries and soils of very low strength Class 3: soils of high workability Class 4: soils of medium workability Class 5: soils of low workability Class 6: rocks of high workability Class 7: rocks of low workability The main attribute is grading. especially the amount of boulders up to 0.1 m 3 • for fine grained soils plasticity and consistency arc also in1portant. Hard and cemented soils are rated as easily workable rocks (Class 6). However, since there are no criteria or guide-
197
1.4 Properties of soils and rocks and their laboratory determination
lines defining hard soil, the shrinkage litnit, which is hardly appropriate, is often used. In DIN 18319 [52c] relative density is also considered in the classification of soils for driving pipes.
11.2
Rock classification
Due to the important role of joints (discontinuities) in the rock 1nass behaviour, rocks are more difficult to classify than soils. A classification based only on the properties of the rock material is insufficient and it is necessary to introduce characteristics that evaluate joints and their influence.
Spacing of joints is defined as the average distance between adjacent joints measured normal to the joint plane. As a rule, drill cores are available for rock characterisation, and the spacing of joints is usually measured along the core axis. The orientation, dip and strike however cannot be detennined from the cores. Often, only segments of the core of various lengths, together with hroken material, are recovered and docu1nented as a result of exploratory core drilling. The state and quality of the core obtained, as well as the lengths of individual core segments are affected hy the quality of the rock and the quality of the drilling, i.e. the technique used and the experience of the drill operators (see Chapter 1.3). l~herefore, rock characteristics hased on core recovery can be too conservative. The main rock characteristics from this kind of investigation are: 1. Core recoi.·ery R which is the ratio of the length of recovered core to the total length of the drillhole (DIN 4022-2 [49]). The characteristic depends on the length of the minimum piece of core considered. Wittke [157] suggested the limit of 50 mm. 2. Rock quality designation RQD is the sum of the lengths of the pieces of the core that
are longer than 100 mm, expressed as the percentage of the total length of the drillhole
[431. 3. The nutnber of discontinuities per 1netre, i.e. frequency (or intensity) of jointing F [49], or joint spacing.
The effect of discontinuities on the mechanical properties of rocks is often manifested by a substantial difference in the defor1nation moduli of rock material and rock mass, or in their wave propagation velocities. The differences can he used in characterising the soundness of the rock. For example rock mass factor (soundness index) can be defined as the ratio of Young's moduli determined in the field (EF) and in the laboratory (E1, "intact" rock material of core specimens), I = (Ep)/(E1). Alternatively, comparison of compressive wave velocities measured in situ by cross hole or up hole tests vp with laboratory velocities vi yields the velocity ratio vp/VI, which can be used in rock quality description. No single rock 1nass classification is generally accepted. In practice, the important classification systems are based on the behaviour of underground openings. In the stand-up tirne classification syste1n developed for tunnelling, rock is ranked into seven rock classes A to G (Fig. 80) according to the stand-up time and the rock span of unsupported tunnel excavation [104]. Rock class Rock behaviour
A
B
c
D
E
F
G
stable
fragile
very fragile
friable
very friahle
compressed
very compressed
Paul von Soos and Jan Bohac
198 10"
10m 1"
1'
10Vr
100Vr
B 6
t c
4 2
~
= 00
c e
1,0 0,8 0,6 0,4 0,2 0,1
stand-up time
Fig. 80. Span versus
~tand-up
Lime diagram with classification (after Lauffer)
The first attempt to base a rock classification on characteristics was made by Deere and Miller [44], by defining five rock classes according to ROD. By introducing further rock properties, other classification systems were developed. For example rock nzass rating (RMR) system by Bieniawski [15], or Q system (rock mass quality index developed in NG!) by Barton et al. [11]. The rock niass rating system, RMR is a sum of the influences of the following paran1eters: • • • • •
uniaxial strength or point-load test ROD value spacing of join ts condition of joints (filling and coarseness) groundwater condition
0 to 30 pts 3 to20pts 5 to 30 pts 0 to25 pts 0 to 10 pts
Allowance for the orientation of the joints was introduced by Bieniawski [141 forindividual types of geotechnical structures: for tunnels down to -12, for foundations down to -25, for slopes down to -60 points. According to the sum of points a rating is designated: ROD RMR designation
0 to 0.25 () to 2() very poor
0.25 to 0.50 21 to 40 poor
0.50 to Cl. 75 41 to 60 fair
0. 75 to Cl.90 61 to 80 good
0.90 to 1.00 81 to 10() very good
The NGI Q system considers six parameters. Their values are based on a large number of tunnelling case histories. The range of the values (shown from the best to the worst rock quality) is as follows: ROD
J., J, Ia
Iw SRF
100 to 0 nun1ber of joint sets roughness of the most unfavourable joint degree of alteration or filling of the weakest joint water inflow stress conditions - stress reduction factor
Cl.5 to 2()
4 to 0.5 Cl.75 to 20
1.0 to 0.05 0.5 to 4()()
199
1.4 Properties of soils and rocks and their laboratory determination
The rock mass quality Q is then expressed as a combination of the six parameters O= ROD:'._<~ J,, J, SRF The firs[ ratio in the equation is related to the block size, the second and the third factors reflect the strength of joints (interblock shear strength) and "active" stress, respectively. According to the values of the index Q, nine classes of rocks are distinguished, from exceptionally poor to exceptionally good rock. Q
0.00!-(J.O!
0.01--0.1
0.1-1
1-4
4-10
!Cl-40
40-100
10Cl-4(Xl
valuati on
exceptionally poor
extretnely poor
very poor
poor
fair
good
very good
extremely good
-
~
- - -
--
400-1000
I
exception-1 ally good
-~
Comparative analyses between RMR and Q systems have also been carried out. Rieniawski [15] reports an analysis of 111 case histories which yielded a relationship between the two systems: RMR=9xln0+44 On the basis of case histories, the classes of the RMR system were incorporated into a stand-up time vs. span diagran1, and can therefore be compared with classifications based on such diagrams (Fig. 81). Barton and Grimstad [10] suggested that the design of supports can be based on the values of 0 using factors (ESR) that depend on the type of structure: 2 to 5 for temporary openings down to 0.5 to 0.8 for the excavations required for underground nuclear power plants.
40.0
I E
·" ~
"'
~ ~
~
10'
1'
10 1D
1Wk 1Mo3Mo 1Yr
10Yr
20.0 10.0 8.0 6.0 4.0
0
~
2.0 1.0 0.8 0.6 0.4 stand-up time
1'1g. 81. Roof span versus stand-up time diagram (after Bieniawski)
Correlations have been developed between some rock characteristics and classifications, which can be used as estimates in comparisons and preliminary design [81 ]: Deformation modulus
Ep = exp((RMR - 40)/10)
Rock quality (soundness) indexes
I= EF/E1 = 0.5 x (ROD/ 100) 2 I= EF/Er = 0.72 x (RMR/100) 2
in 1000MNm- 2
200
Paul von Soos and Jan Boh.3.C
For rocks without unfavourable joints: Lower limit of the friction angle Lower limit of the cohesion Compressive strength of the rock material for RMR < 42 for RMR > 42 According to Barton and Grimstad [10]
min 'PI = 0.50 x RMR + 7.5 min q = 0.020 x RMR
in grad in MNn1- 2 2
Oct! =
in MNrn-
Oct!
in MNrn- 2
0.07 x (RMR - 11) = 0.17 x (RMR - 30)
Oct! =
7yQ'I'
in MNm- 2
Some classification systems consider and evaluate further features of the rock mass, such as stratification and tectonics (c. g., classification by Brii.utigarn and }Jesse [24]). The above classification systems aim to predict the rock behaviour for planning and preliminary design. However the rock classes according to DIN 18319 [52c] and DIN 18312 [52b] serve as a basis for pricing. 1hc first document distinguishes rock classes according to one-dimensional strength (4 grades) and discontinuities (2 grades). the latter n1akes use of the rock behaviour in underground openings and the support needed.
12
References
[11 AASHTO (American Association of State Highway and Transportation Officials): Standard specification for transportations materials and methods of sampling and testing, Part 2. 1990, ISBN 1-56051-005-6. [21 Aiban, S.A., Znidarcic, D.: Evaluation of the fl.ow pun1p and constant head techniques for permeability measuren1ents. Geotechnique, 39 (4). 1989, 655--666. [31 ASTM Designation: D 2487-98: Standard Practice for Classification of Soils for Engineering Purposes (Unified Soil Classification System). [41 ASTM Designation: D 4186-82: One-Din1ensional Consolidation Properties of Soils Using Controlled-Strain Loading. (5] ASTM D 4254-91: Standard test method for n1inimum index density and unit weight of soils and calculation of relative density. [6] ASTM STP 977: Advanced triaxial testing of soil and rock, ASTM STP 977, R.T. Donaghe. R.C. Chaney and M.L. Silver (eds), American Society for Testing and Materials, Philadelphia, 1988. [71 Atkinson, ]. H., Siillfors, G.: Experimental determination of stress-strain-ti1ne characteristics in laboratory and in situ tests. Proc. I I th ECSMFE, Florence, 1991. Balken1a, 9 I 5--956. (8] Atterberg, A.: Die Plastizittit der Tone. Int. Mitt. filr Bodenkunde l, 19 I I, 10 pp. [9] Baldi, G., Hight, D. W., Thornas, G. E.: A reevaluation of conventional triaxial test method. Advanced triaxial testing of soil and rock, ASTM STP 977, R. T. Donaghe, R. C. Chaney and M. L. Silver (eds), American Society for Testing and Materials, Philadelphia. 1988, 219-263. [ 10] Barton, N., Grimstad, E.: The Q-System Following Twenty Years of Application inNMT Supports Selection. Felsbau 12, 1994, 428-436. [11] Barton, lV.., Lien, R., Lunde, 1: Engineering classification of rock masses for the design of tunnel support. Rock Mechanics, Vol. 6, 1974, 189-236. l12l Beyer, W.: Zur Bestimmung der Wasserdurch!assigkeit von Kiesen und Sanden aus der Korn\1erteilungskurve. Wasserwirtschaft u. -technik 14, 1964. (13] Bicz6k, E.: Untersuchung der Luftdurchltissigkeit \1on Kornhaufen. Proc. 3rd Budapest Conf. on Soil Mechanics and Found. Engng., I 968, 11-22. [14] Bieniawski, Z. T: The Geomechanics Classification in Rock Engineering Applications. Proc. 4th Conf. ISRM Montreaux, 1979, Vol. 2, 41--48.
1.4 Proper1ies of soils and rocks and their laboratory de1errnination
201
[15] Bieniawski, 7. T.: Engineering rock mass classifications, 1989, J. Wiley & Sons, 251 pp. [16] Bishop, A. W: 1he principle ot effective stress. Tecknish llkeblad, 106, 1959, 859--863. [17] Bishop, A. W: Progressive Failure -with Special Reterence to the Mechanism Causing i1. Proc. Geot. Conf. Oslo 1967, Vol. 2, 142-150. [18] Bishop, A. W, Green, G.E., Garga, V.K., Andresen, A., Brott:n, J.D.: A new Ring-Shear Apparatus and its Application 10 1he Measurement of Residual Strength. Geotechnique 21, 1971, 273-328. [19] Bishop, A. W, Henkel, D.J.: The Measurement of Soil Proper1ies in the Triaxial Test. 2nd ed. Edward Arnold, London, 1962. [20] Bjerrum, /,.: Geotechnical Properties ot Norwegian Marine Clays. Geotechnique 4, 1954, 49-69. [21] Bjerrum, L.: Problems ot Soil Mechanics and Construction on Son Clays and Structurally Unstable Soils (collapsible expansive and o1hers). Proc. 8th ICSMFE ~losco'Yl-· 1973, Vol. 3, 111-153. [22] Bjerru1n, /,., Landva, A.: Direc1 simple shear tests on a Norwegian quick clay. Gt:o1echnique 16 (1), l 966, 1-20. [23] Borovicka, TT.: Der Wiener Rou1inescherversuch. Mitt. d. Inst. t Grundhau und Bodenrnech., T. H. Wien, No. 5, 1963. [24] Briiutigant, F., Hesse, K.H.: Ingenieurgeologische Gebirgs1ypisierung Hir Tunnelbauten an der DB-Neuhaustrecke Hannover-Wtirzhurg in Osthessen. Ber. 4. Nat. Tag. Ing. Geo\., Goslar 1983, 151-165. [25] British Standard Institution: Code of Practice for Site Investigation. HS 5930, London 1981. [26] British Standard Institution: Methods ot test tor soils tor civil engineering purposes.. HS 1377, 1975. [27] Broch, E., Franklin, I. A.: The point-load s1rength lest. Int. J. Rock Mech. Min. Sci .. 9, 1972, 669-697. [28] Brooks, R.H., Corey, A. T: Hydraulic properties of porous media. Colorado State llniv, Hydro!. Paper. No. 3, Mar. 1964, 27 pp. [29] Hundesrninis1eriun1 fiir Verkehr: Zusatzliche Technische Vertragsbedingungen und Richtlinien fur Erdarbeiten im Stra/3enbau ZTVE-St B 94. [30] Bureau of Reclamation: Earth Manual, Denver, 1960. [3l] Burghignoli, A.: An experimental study ot the structural viscosity of sof1 clays by means ot continuous consolidatin tests. Proc. 7th ECSMFE. Brighton 1979, Vol. 2, 23-28. [32] Burland, JB.: On the compressibility and shear strength ot natural clays.. Geotechnique 40, 1990, 329-378. [33] Carrier, W.F., Beckman, J.F: Correlation be1ween index tests and the properties of remoulded clays. Geotechnique 34, 1984, 211-228. [34] Casagrande, A.: Die Araorneterrnethode zur Bestirnrnung der Kornverleilung. Springer-Verlag, Berlin, 1934. [35] Casagrande, A.: Characteristics ot cohesionless soils affecting the stability of slopes and earth fills. J. Boston Soc. Civ. Eng., 1936. [36] Casagrande, A.: Classification and iden1iflca1ion ot soils. Proc. Arner. Soc. Civ. Engrs., 1947, Vol. 73, 783-810. [37] Clayton, C.R. I., He:ymann, G.: Stiffness of geornaterials at very sn1all strains. Geotechnique 51, 2001, No. 3, 245-255. [38] Clough, G. W., Schlnidt, B.: Design and Performance of excava1ions and 1unnels in son clay. In: "SoU Clay Engineering", Elsevier 1981, 569-634. [39] Coop, M. R., f,ee, I. K.: The behaviour ot granular soils at elevated stresses. Proc of C. P. Wroth Memorial Symposium, Oxtord, T. TeUord, 1993, 186-198. [40] Cornforth, D. H.: Some experiments on 1he influence ot strain conditions on strength of sand. Geotechnique 14, 1964, 143-167. [41 J Cuccovillo, T, Coop, M. R.: The measurement of local axial strains in triaxial tests using LVDTs. Geotechnique 47, No. I, 1997, 167-171. [42] Cuccovillo, T., Coop, M. R.: On the mechanics of structured sands. Geotechnique 49, No. 6. 1999, 741-760.
202
Paul von Soos and Jan Boh
[43]
Deere, D. U.: Technical description of rock cores for engineering purposes. Felsmechanik und Ingenieurgeologie. 1963. Vol. 1/1, 16-22. Deere, D. U., Miller, R.P: Engineering classification properties for intact rock. Technical Report No. AFNL-TR-65-116, Aii- Force Weapon~ Laboratory, New Mexico, 1966. Dickey, J. lV, Ladd, C. C., Rixner, J. J: A plane strain shear device for testing clays. Res. Rep. R68-3. Soils publicatioo 237, Mass. Inst. of Technology, 1968, 158 pp. DIN-Normen des NABau; Baugrund - Versuche und Versuchsger8.te. "lJntersuchung von Bodenprohen" == UvB. DIN 18121-1; UvB; Wassergehalt - Bestimmung durch Ofentrocknung, .A.pr. 1998. DIN 18121-2: UvB; Wassergehalt - Bestimmung .ttai 1998. DIN 18132: UvB; Bestiinmung des Was,.<;eraufnahrnevern10gens, Dez.1995. DIN 18135: UvB; Eindimensionaler Kompressionsversuch, 2002. DlN 18136: lJvB; Bestimmung der einaxialen Druckfestigkeit, Aug. 1996. DIN 18137-1: UvB; Bestimmung der Scherfestigkeit. Begriffe, Aug. 1990. DIN 18137-2: UvB; Dreiaxialversuch, Dez. 1990. Dl.N l-8137~3: lJvB; Direkter Scherversuch, Entwurf 10.1997. DIN 4020: Geotechnische Untersuchungen fUr bautechnische z,vecke, Okt. 1990. DIN 4022-1: Benennen und Beschreiben von Boden und Fels, Sept.1987. DIN 4022-2: Benennen und Beschreiben von Boden und Fels, Schichtenverzeichnis fjjr Bohrungen im Fels, Marz 1981. DIN 4023: Baugrund- und Wasserbohrungen -zeichnerische DarsteUung der Ergebnisse, Marz 1984. DIN 18196: Erd-und Grundbau; Bodenklassifikation flir bautechniscne Zwecke, Okt. 1988. DIN 18299 and DIN 18300 to DIN 18384: Verdingungsordnung fi.ir Bauleistungen VOB Teil C. 1996. DIN 18300: Erdarbeiten, Juni 1996. DIN 18312: Untertagebauarbeiten. Dez. 1992. DIN 18319: Rohrvortriebsarbeiten,Juni 1996. Deutsche Gesellschaft flir Geotechnik (DGGT): Empfehlungen filr die Versuchstechnik im Fels: E 1: Einaxiale Druckversuche an Gesteinsproben, 1979/1986. E 2: Dreiaxiale Druckversuche an Gesteinsproben, 1979/1986. E 3: Dreiaxiale Druckversuche an gekllifteten GroBbohrkernen im Labor, 197911986. E 5: Punktlastversuche an Gesteinsprohen, 198211986. E 10: Indirekter Zugversuch an Gesteinsproben -Spaltzugversuch, 1985/1986. E 11: Quellversuche an Gesteinsproben, 1986. E 12: Mehrstufentechnik bei dreiaxialen Druckversuchen und direkten Schervcrsuchen. 1987. E 13: Laborscherversuch in Felstrennflachen, 1988.
1.4 Properties of soils and rocks and their lahoratory detennination [53i] [53j] [54]
203
E 16: Ein- und dreiaxiale Kriechversuche an Gesteinsprohen, 1994. E 17: Einaxiale Relaxa1ionsversuche an Gesteinsproben, 1994. Dyvik, R., lY!adshus, C: Laboratory measure1nents of Gmax using bender ele1nents. Proc. ASCE Conven1ion '"Advances in the ar1 of testing soils under cyclic conditions'', Detroit, Michigan, 1985. [55] Eberle, H.: Zusa1nmenhang zwischen PPr und wrr 1nit den Plastizitiitsgrenzen. Diplomarbeit TU Mlinchen, 1966. [56] Engel, J., Franke, D.: Improved methods for the calculation of geotechnical properties from the results of classification tests. Proc. 15th lCSMFE, Hamburg 1997, Vol. 1, 283-285. [57] Farmer, J. W.: Engineering properties of rock. E & F. N. Spoon Ltd, London, 1968, p. 180. [58] Farrell, E., Larsson, R. and Schuppener, B.: Detennination of Atterberg li1nits, ECSMFE, Copenhagen, 1995, Workshop 2. [59] Forschungsgesellschaft fOr StraBen- und Verkehrswesen: Technische Prtifvorschrif1en flir Boden und Fels im StraBenhau (TPBF-S1B). [59a] Teil B3.2: Bestimmung der Korndichte mit dem THuchgef313 nach Dr. Ing. HaHs, 1988. [59b] Teil B5.2: Bestim1nung von Feinkornanteilen in grobkOrnigen BOden und Baustotlen mit dem Absetzgera1 nach Dr. Ing. Haas, 1988. [60] Franklin, J.A., Chandra, R.: The Slake Durability Test. Int. J. Rock Mech. ~lin. Sci. 9, 1972, 325-341. [61] Fredlund, D. G., Rahardjo, H.: Soil mechanics for unsaturated soils. J. Wiley & Sons, 1993. [62] Friedrnan, (l. AJ., Sanders,./. F..: Principles of sedimentology. J. Wiley & Sons, 1978. [63] Georgiannou, V N, Rampe!lo, S., Silvestri, F: Sta1ic and dynamic measurements of undrained stiffness on natural overconsolidated clays. Proc. 11th ECSMFE, Florence, Balkema, 1991. 91-95. [64] Goldscheider, M., Vardoulakis, /.: Biaxialgerat zur Untersuchung der Festigkeit und Dilatanz von Scherfugen in BOden. Geotechnik 3, 1980, 19-31. [65] Green, G. !'..: Strength and defonna1ion of sand n1easured in an independent stress control cell. Proc. Roscoe Mem. Symp. "Stress-Strain Behaviour of Soils", G.T. Foulis & Co, 1972, 285-323. [66] Gudehus, G.: Bodenmechanik. F. Enke Verlag, Stuttgart, 1981, 268 pp. [67] Gudehus, G., Leinenkugel, H. J.: Fliel3druck und Fliel3bewegung in bindigen Bliden: neue Methoden. Vortriige der Baugrund1agung 1978 in Berlin, DGEG, 411-429. [68] Haas, H.: Verbesserungen des Analysenverfahrens zur Ermittlung der Korngrlll3enverteilung feinkOrniger BOden und Bodenanteile nach DTN 18123. Forschungsarbeiten aus den1 Stral3enund Verkehrswesen, Hef1107, Kirschbaum VerlHg, Bonn, 1997. [69] Harn!Jly, E. (~.:A new triaxial apparatus. G~otechnique, 19 (2), 1969, 307-309. [70] lfa,nbly, !'.. C.: Plane strain hehaviour of remoulded normally consolidated kaolin. Geotechnique. 12 (2).1972, 301-317. [71] Ha1notJche, H.K., Laroueil, S., Roy, M., Lutenegger, A.J.: ln situ evaluation of Ko in eastern Canada clays. Can. Geotech. Journal, Vol. J2, 1995, 677--688. [72] Hansen, B.: Shear-Box-Tests on Sand. Proc. 5th lCSMF 1961, Vol. 1, 127-132. [73] !lead, K.H.: Manual of Soil Laboratory Testing, Pentech Press, London, Vol. 1, 1980; Vol. 2, 1982, Vol. 3, 1986. [74] Henkel, D .. T.: The shear strength of saturated re1noulded clays. Res. Cont on Shear Strength of Cohesive Soils, Boulder, CO, 1960, 533. [75] Henkel, D.J., Wade, N.H.: Plane strain tests on a saturated remoulded clay. JS~IFD, ASCE, Vol. 92, 1966, No. SM 6. 67-80. [76] Hight, D. W., Gens, A., Synies, M. J.: The development of a new hollow cylinder appara1us for investigating the effects of principal stress rotation in soil. Geotechnique 33, No. 4, 1983, 35.1-383. [77] Hilf, J. W.: An investigation ')f pore-water pressure in compac1ed cohesive soils. PhD Diss, Tech Men10 654, Bureau of Reclan1ation, Denver, CO, 1956.
204
Paul von Soos and Jan Doh::lC
[78] lfogentogler, C. A., Terzaghi, C.: Interrelationship of road, load and subgrade, Vol. 3, Public Roads, Washington 1929, 37-64. [79] Holtz, R. D., Kovacs, W D.: An introduction to geotechnical engineering. Prentice-Hall, 1981. [80] !Torn, A.: Die Scherfestigkeit von Schluff. forschungsberichte des Landes NordrheinWestfalen. Westdeutscher Verlag, Kolt1, 1964. [81] Hdnisch, K.: Zur Strategie der Gestaltung grofier Krafthauskavernen. Schriftenreihe Lehrst. und Prlifamt filr GB. BM un
1.4 Properties of soils and rocks and their laboratory determination
205
[107] Lu"1h, P: Application of Statistics in Soil Mechanics. In: Lee, J.K: Soil Mechanics-Nev,• Horizons. Newnes. -Butterworts, London 1974, 44-111. [108] Lupini, J.F, Skinner, A.E., Vaughan, PR.: 111e drained residual strengths of cohesive soils. G6otechnique31, 1981, 181-213. [109] A.farsland, A.: The shear strength of stiff fissured clays. Stress-strain behaviour of soils, R.H.G. Parry (ed.).1972. G. T. Foulis & Co, 59-68. [110] A.fayne, P.W and Kulhavy, FJI.: Ko-OCR-Relationships in Soil. Journal Geotech. Eng. Div. ASCE, Vol. 108, 1982, 851-872. [111] Afesri, G., c:astro, A.: CICC concept and Ko during secondary con1pression. J. Geotech. Engng. Div. ASCE, Vol. l IJ, 1987, 230-249. [112] Afesri, G., Feng, T.W., Ali S., Hayat, T.M.: Pern1eability Characteristics of soft Clays. Proc. 13th ICSMFE, ~ew Delhi 1994, Vol. I, 187-192. [113] .Mitchell, J. K.: Fundamentals of Soil Behaviour, 2nd ed. J. Wiley & Sons, 1993. [I 14] Mogarni, !., Yoshikoshi, JI.: On the Angle of Internal Friction of Coarse l\.faterials. Proc. 3rd Budapest Conf. on Soil Mech. and Found. Engng. 1968, 19<}..-196. [115] .Miihlhaus, H.B., Vardoulakis, /.: The thickness of shear bands in granular n1aterials, G6otechnique 37, 1987, 271-283. [116] A-furayama, S., Shibata, T.: On the secondary consolidation of clay. Proc. 2nd Japan Congr. Test. Mat., Non-Metallic Mat., Kyoto, 1958. [ 117] ,\lejj,' JI.: Der Wasseraufnahme-Versuch in der bodenphysikalischen Prilfung und geotechnische Erfahrungswerte. Bautechnik 65, 1988, 153-163. [118] Ohde, .T.: Bauingenieur 1939, 454 pp. [119] Ohde, J.: Grundbaumechanik, Htitte, Vol. III, 27. edition, 1956, 886--902. [120] Olson, R. E.: Shearing Strength of Kaolinite, Illite, Montmorillonite. ASCE, Vol. LI, 1974, 121.1-1229. [121] Ostermayer, H.: Die Zusamn1endrilckbarkeit gen1ischtkOrniger Boden. Bauingenieur 52, 1977, Heft 7, 269-276. [122] Parry, R.H.G.: Triaxial Compression and Extension Tests on Remoulded Saturated Clay. Geotechnique 10, 1960, 166-180. [123] Pearce, .T.A.: A new true triaxial apparatus. Proc. Roscoe l\.fen1. Symp. ';Stress strain behaviour of soils", R.H. G. Parry (ed.), 1972, G. T. Foulis & Co, 330-339. [124] Proctor R.R.: The design and construction of rolled earth dams. Engng. News Rec. 111, 1933, (9) 245-248, (10) 216-219, (12) 348-351, (13) 372-376. [125] Rendulic, L.: Relation betv.-·een void ratio and effective principal stresses for a remoulded, silty day. Proc 1st ICSMFE, Cambridge (l\.fass), 1936, Vol. 3, 48-51. [126] Rengers: Unebenheit und Reibungswiderstand von Gesteinsfttichen. VerOf. des Inst. for Bodenmech. und Fels mech, lJni Karlsruhe, No. 47, 1971. [127] Road Research Laboratory: Soil Mechanics for Road Engineers. London 1952. [128] Roscoe, K.H.: An apparatus for the application of simple shear to soil samples. Proc. 3rd. ICSMFE, Ztirich, 1953, Vol. 1, 186-191. [129] Rowe, PW.: The Relation betv.-·een the Shear Strength of Sands in Triaxial Compression, Plane Strain and Direct Shear. Cieotechnique 19, 1969, 75-86. [130] Saada, A. S., Townsend, F C.: State of the Art: Laboratory strength testing of soils. Laboratory shear strength of soils. ASTM STP740, R. N. Yong and F. C. Townsend (eds.),An1erican Society for Testing and Materials, 1981, 7-77. [131] Schmertmann, J.N., Osterberg, J.-0.: An experin1ental study of the development of cohesion and friction with axial strain in saturated cohesive soils. ASCE Research Conference on Shear Strength of Cohesive Soils, Boulder (CO), 1960, 643-694. [132] Schn1idt, 1V. 0.: Suggested method of test for organic carbon content of soil by wet combustion. ASTM STP 479 ';Special procedures for testing soil and rock for eng. purposes", 1970, 217-278. [133] Schofield, A. /\'., Wroth,(.'. P: Critical state soil mechanics. l\.fcGraw-Hill, 1968, London.
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[ 134] Schultze, E.: Somt: Aspects Conc.:t:rning tht: Application of Statistics and Probability to Foundation Structures. Proc. 2nd Int. Cont Application Statistics and Probability in Soil. Aachen, 1975. 457-494. [135] Sbnons, 1V.E.: Normally consolidalt:d and lightly over-consolidated cohesive 1naterials. General Report Session 2, Conf. on Settlement of Structure, London 1975, 500-530. [136] Skernpton, A. W: Notes on the Compressibility of Clays. Quart. J. Gt:ol. Soc. London, Vol. C. 1944.119-135. [137] Ske"1pton, A. W.: The Collodial "Activity" of Clays, Vol. l, Proc. 3rd ICSMFE, l 953, 57-6"1. [138] Skempton, A. l-V.: The Pore-Pressure-Cot:fficient A and B. Geott:chnique 4, 1954, 143-147. [139] Ske1npton, A. l-V.: Disc. on Paper Grace and Henry in Proc. Inst. Civil. Eng. 7. 1957. [140] Skempton, A. l-V.: Long-term stability of clay slopes. Gt!olt:chnique 14, 1964. 77-101. [141] Skempton. A. W: The consolidation of clays by gravitational co1npaction. Quart. J. Geol. Soc. Loodoo, 125, 1970, 373-411. [142] Skenipton, A. l-V.: Rt:sidual strength of clays in landslides. foldt:d strata and the laboratory. Geotechnique 35, 1'::185. 3-18. [143] Skernpton, A. iv., IVorthey, R. D.: The Sensitivity of Clays. Geotec.:hnique 3, 1952, 30-53. l144J Soos v., P: Zur Ermittlung dt:r Bodenkennwerte mil BerOcksichtigung von Streuung und Korrelation. Vortr. Baugrundtagung 1982. 83-103. p45] Stark. T.D., Ei£L Tl. T.: Draine
1.5
Constitutive laws for soils from a physical viewpoint Gerd Gudehus
1 Introduction l.l
Motive and objective
When a block foundation \Vas pushed into the ground by a skew load near Karlsruhe 1990, the participating experts were disappointed that none of their predictions agreed satisfactorily \Vith the ohscrvations. The ground, with horizontal layers of silt, fine and gravelly sand, had been meticulously investigated, hut apparently its hchaviour could not he adequately described hy the different constitutive relations and material parameters. Nobody had anticipated the pre-stress in the grain skeleton due to capillarity in the unsaturated region \Vhich only made some of the calculations realistic afterwards. A sheet-pile wall was later driven into the same ground and a pit was excavated down to groundwater next to it. Again predicted and observed displacements deviated substantially from each other, hut some predictors could at least calculate reasonably the limit state caused hy yielding struts. The results of the calculations and ohscrvations could only be brought close to each other afterwards by adaption of the soil parameters. One must therefore ask - what is the use of constitutive laws and material parameters, which apparently play a key role, if even in such simple cases big differences occur between different predictions and reality? The prohlcm is not new, hut has been aggravated by the increasing po-..ver of computers and correspondingly increased expectations. The user would like to know - which constitutive relation is applicable, how can the niaterial parameters be determined practicably and reliably, which initial and boundary conditions are cornpatible with the latter and which nurnerical discretization and iteration is appropriate for solving the proble1n. Objectivity and robustness of mechanical models depend mainly on the constitutive relation and the associated data. The attempts at prediction mentioned above were not the first efforts towards clarification and agreement. Constitutive la-..vs have heen compared with each other and -..vith laboratory data in several -..vorkshops. Nc-..v and modified constitutive laws arc continuously published in periodicals and conference proceedings. Many of these papers are not even tractahle, others are only availahlc to a fc-..v insiders and the determination of the parameters is rarely outlined. On the other hand, over-simplified and often unverified constitutive relations arc often used in practice. At the Xlth International Conference for Soil Mechanics and Foundation Engineering, a Japanese working group produced the report 'Constitutive Laws of Soil' fl 9]. Micromechanical aspects, elastoplastic constitutive relations, viscous effects and dynamical problems were clearly outlined and have remained mostly valid until now. A survey was missing ho-..vever, and little -..vas said about the application of the laws. My own contribution on constitutive laws in the previous Grundbau-Taschenbuch gives a more concise overview, but even this is not enough for the user.
208
Gere! Gu
The present contrihution is therefore the result of a thorough review of the situation and should better meet the wishes of users. It is confined to soils with unccmcntcd grain skeletons and therefore already covers a very wide range. The approach is physical and therefore uncommon to many experts on constitutive relations. In my opinion, constitutive laws have to relate to physical state variables and their changes realistically. In the first main section soil states and their changes are therefore covered in detail. Constitutive laws in a narrower sense follow in the second main section and some extensions of the theory are included at the end. The selection, organisation and representation of the material is unusual. This contribution is mainly directed at users, hut also to developers of constitutive equations. Some knowledge of physics and continuum mechanics is assumed, as a handbook cannot give such a detailed explanation as a textbook. Above all the reader should keep in mind the characteristics of the real soil and its change by geotechnical actions, and should critically review constitutive laws he fore working with them.
1.2 Contents The description of the soil state (Section 2.1.1) starts from the fundamental difference between molecules and grains. The latter arc changed by encounters and therefore are never all equal, can not be precisely described and their interaction is dissipative. The position of fictitious marker grains is appropriate as a state variable, whereas the deformation does not say anything about the soil state. 'The state of simple grain skeletons is widely described by the solid fraction and the grain pressure tensor. Pressure and mass fractions of pore water have to be included too and exceptionally gas pressure and ion fractions. Instead of the symbols 0 1 and u, which arc reserved for tensile stress and displacement in mechanics, p 5 and Pw are used for the partial pressures of the solid and the water and Pw is not reduced hy the atmospheric pressure. Velocities of the solid and the pore water normally have to be added. Further state quantities for describing fluctuations and localization zones can also be introduced for a better description of the material. Relations hctwccn state quantities are denoted as 'state conditions' in which deformations do not occur (Section 2.1.2). These are of a thermodynamic nature for pore liquid and gas. The well-known separation into skeleton and pore water pressure is justified by the pressure-independency of both fractions and can be extended to unsaturated grain skeletons. Pressure-dependent void ratios are presented in a granular phase diagram as asymptotic values for proportional compression, constant volume monotonous deformation and isoharic cyclic deformation of simple grain skeletons, as long as these are granulometrically permanent. These asymptotic states substitute conventional limit states and serve as restraints for constitutive relations. Boundary conditions are briefly indicated, as they have to be compatible with the material properties. The role of initial conditions is emphasized for fixing initial states, although they are rarely distinguished in a physical sense. The section on fahric (Section 2.1.3) illustrates that mean values do not generally suffice for describing the state of soil. Systematic irregularities in the arrangement and the forces of particles can he described in principle, but they are rarely allowed for in constitutive laws. Different length scales from particle size to layer thickness are neglected when restricting the assessment to mean values, although they can play a hig role.
209
1.5 Constitutive laws for soils from a physical viewpoint
Changes of state of simple grain skeletons are described \\ ith the aid of so-called soil clements (Section 2.2.1). Shortening is used for deformation related to pressure to precisely define some material properties with the aid of familiar diagrams, in particular friction and dilatancy. The grain skeleton behaviour under alternating loading patterns, particularly stress or strain cycles, is represented with the aid of associated paths and the differential stiffness is represented by so-called response polars. The thermal activation is treated in detail (Section 2.2.2), as it is decisive for the viscoplastic grain skeleton behaviour and the fluid transport in the pore space. The associated mechanical activation is only briefly indicated, as it is physically less understood and can therefore only be covered heuristically by constitutive relations. Inevitably hcuristical relationships are also stability considerations, with soil elements following molecular dynamics. This illustrates that conventional rupture and failure criteria for describing soil behaviour arc not adequate from a physical vic\\'point, and also not required. Some geotechnically relevant mechanisms in earth bodies and on structural parts in contact with them are briefly introduced (Sections 2.3.2 and 2.3.3). Features of stabilization and destabilization, including rearrangements and pressure changes for the grain skeletons and also thermal activation in the grain skeleton and the pore fluid, are emphasized. This leads to hints on constitutive laws. 'The next main section starts with elasticity allo\\ ing for skeleton pressure and the degree of saturation (Section 3.1.1). Pscudoclastic finite constitutive relations are only briefly mentioned, as they can at best only be used for crude estimates (Section 3.1.2). How the accumulation of anelastic changes of shape and stress under alternating loadings can be approximated by analytical expressions (Section 3.1.3) is also briefly covered. Flow and creep laws related to thermal activation arc treated \\ ith similar brevity. Starting from perfect solids, elastoplasticity with simple flow condition and associated flow rule is outlined \\ ithout special representations of functions (Section 3.2.1). Volumetric hardening or softening is covered by means of an equivalent skeleton pressure, which covers several of the usual constitutive relations. Extensions to the relations are briefly introduced by diagrams. and criteria arc proposed for selection from a wide choice of elastoplastic constitutive relations (Section 3.2.2). The extension into viscoplasticity then is simple, due to the uniform effect of thermal activation (Section 3.2.3). For describing changes of state due to the deformation of grain skeletons by rearrangements, hypoplasticity has become equivalent to clastoplasticity in many applications (Sec· tion 3.3.1). lt is also more easily outlined and used. The few easily determinable material parameters are valid for a wide range of density and pressure. The default of too rapid an incTcasc of changes of pressure and shape under alternating loading with small deformation amplitudes and a number of cycles, is rcn1ovcd by the so-called intergranular strain. The evolution of this and the differential stiffness related to it are briefly outlined (Section 3.3.2). An extension of the theory for viscous effects is very briefly outlined (Section 3.3.3). 1
1
1
1
Physico-chcmical and granulometric changes ares only mentioned in passing as they arc. as yet, scarcely allowed by constitutive relations (Section 4.1). Transport laws for the pore fluid arc related to thermal activation and the mechanically activated transport of grains is briefly indicated (Section 4.2). Some 1proaches for dealing with shear bands arc introduced as they can be used for interfact.. ~iements. The cracking of grain skeletons is only briefly referred to as physically justified interface clements are scarcely available for it yet (Section 4.3).
210
Gerd Gudehus
The list of references has hccn kept small with only a few books and papers given as examples. This is not to emphasize or set aside any of the numerous contributions on constitutive relations. I asked 13 experts to give detailed information on the mathematical, physical and experimental aspects as well as on the range of applications and verification. I would like to thank Profs Arslan, Hashiguchi, Molenkamp, Nova and Vogt for their valuable contributions. However, the fact that even they could not answer some questions shows the size of the prohlcm. My proposal to release some of the procedures for constitutive relations on the Internet has only hccn accepted until now (July 2000) by followers of hypoplasticity.
2 2.1
States and changes of state States
2.1.1 Description of state 111e state of a system of permanent atoms or molecules is fully described by the positions and velocities of all the particles l4J. Changes of state result from conservation laws and interaction potentials, generally including external gravitational and electromagnetic fields. The forces among the particles (called conservative) result from their relative position via the spatial derivative of the interaction potentials, in general, and also from relative velocities. For the macroscopic description of the c4uilihrium states of gases and fluids, temperature and density suffice, from which the pressure is uniquely obtained by equations of state. For ideal solids, i.e. homogeneous crystals, the stress tensor is ohtaincd from the stretching of the crystal lattice compared to the stress-free state. In colloid physics, identical permanent particles and interaction forces determined hy their mutual position (and in general velocity) are assumed, and equations of state are obtained for the mean values. For soils, however, the solid particles (grains) are never all equal and are also not permanent, and the non-conservative interaction forces do not follow from the mutual position as the particles arc changed hy encounters. The methods of thermodynamics can only be applied to the pore fluid and to interfaces between solid, li4uid and gaseous constituents. The numerical following-up of all grains (granular dynamics) can, as yet, yield only qualitative results because of inevitable simplifications. Atlcmpts arc niadc to catch size, shape and hardness of grains statistically. Such granulometric quantities cannot precisely be defined and are therefore fuzzy and partially suhjcctivc. The boundaries of granulometrically hornogeneous regions as layers, inclusions and faults therefore have to he suitahly simplified. The state of a sirnple grain skeleton is sufficiently described by mean values. The required state variables are, as a minimum, the solid volume fraction ns and the symmetric partial stress tensor Ts of the skeleton or an equivalent (e.g. the void ratio e = n 5 /(l - n 5 ), or the density index Id as outlined in Section 2.1.2 and the grain pressure tensor where Ps = -T5 ). The position is dcscrihcd hy the coordinate x5 of fictitious marker grains. Other than for an elastic solid, the deformation is not a state variable of the soil. Cross sections with isolines, grey or colour tints (Fig. 1) arc appropriate for graphical representation of the field of solid fractions. A loosened shear zone appears as hand or
1.5 Constituth·e laws for soils from a physical viewpoint
211
Fig. 1. Rt:prt:sentation of the field of pressure-corrected density index (Eq. 2.20) using grey tints
line. If the picture is not the same for parallel planes (plane-parallel or plane state) several cross sections are required. With the grain pressure P.i· = trP 5 averaged over a ll directions (p' in soil mechanics convention) the deviator -
!
T; = Ts + p., l
(2.1)
is obtained. From the stress vector ts = T 5n, related to the direction n, the normal pressure p 511 =
n T P, n
(2.2)
(conventionally a' ) and the shear stress tn
=Ji P,
n 12
- p;n
(2.3)
are obtained . The prinLipal stress or pressure components p, 1(= - '/ ~ 1 ), P .12 and P.13 with t,, = 0 act in the three principal directions. Useful invariants are, apart from p_,, the spatially averaged shear stress tm
= J trT; 2 /5
(2.4)
and the so-called L ode-parameter (Fig. 3c) cos 38s = ,./6
tr'r.. 3 / (tr T"s 2 ) 3/ 2
(2.5)
For graphical representation of the skeleton stress field, sectio ns with principal stress crosses can be used (Fig. 2a). Two principal stress components Psi and p, 2 and their directions are thu s represented for plane-parallel states, but not the third principal stress component Ps3 (P.d ~ (p 5 1+p 5 2)/3 nearly holds). The directional field leads to trajectories which can be seen as pressure-spreading lines, for curvatures with only o ne sign and also as arching lines. Non-plane-parallel states can be represented by projection into two orthogonal families of parallel section planes. Except for symmetry planes, there are no trajectories and also no lines of arching or pressure spreadi ng. "lhe vectors of pressure
212
Gerd Gudehus
1~
I
+
+
I~
1..-
+
r--r-1--r1
a
b
Fig. 2. Representation of grain pressure fields \vi th stress crosses (a) or vectors (b)
and shear stress can be represented along section lines (Fig. 2b)- Spatial states require two orthogonal families of parallel section planes. One can represent simplified pressure fields with sectionally assumed distributions and resultants and can also allow for equilibrium conditions (including uplift and seepage force from pore water). Different diagrams are suitable for representing local stresses (Fig. 3). Stresses and principal components arc represented hy stress circles, where the pole also shows directions (a). Principal pressure components appear as points on the planes in a stress space. The plane p,2 = p,3 (b) is frequently used, and also the central projection top, = 1 (c). Several points of an earth body can thus be represented in one figure, however this representation is not complete. The state of the grain skeleton is widely represented by the field of solid fractions and grain pressure tensors with marked positions, but not completely. Tf the grain properties are substantially changed hy technical actions - e.g. under pile footings or in a hallast track - one has to introduce granulomctric variables. The mechanical behaviour is also substantially changed by irregularities of the mutual positions and forces of the solid particles (Section 2.1.3). The state of pore water and gas is also extensively c.lescrihed by mass and pressure fractions, but not completely. Volume fraction nw and pressure Pw are taken for the pore water (or the water content tv = Pwnw/(p 5 n 8 ), or the degree of saturation Sr = nw/(l - n 5 ) and
Ps1·Ps2
.
B
c
.c •B
A
•
•A
p,
e, Pa3·Psm13
a
b
c
Fig. 3. Representation of stress states by stress circles (a), two principal components (b) and pressure-normalised deviators (c)
1.5 Constitutive laws for soils from a physical vie,,.ypoint
213
the pore pressure u = Pu, - Pa referred to the atmospheric pressure Pa}. The gas fraction 1 - ns - nw, the gas pressure Pg, and the molar fractions 1n1 and nig of ions and gas molecules dissolved in the pore water also have to be added. With gas pockets or channels, such mean values do not suffice (Section 2.1.3}. ng =
The description of state is completed hy including velocities. The velocity Vs = X.,. and the stretching rate D = (\7vs - Vs V)/2 belong to the grain skeleton. The velocity Vw of the pore water is frequently used as state variahle and is objectively replaced by Vu, - v.1. in the case of grain skeleton rearrangements. Partial velocities of other fluid fractions rarely have to be added. The absolute temperature Tis seldom explicitly allowed for, although it is at least effective via thermal activation (Sections 2.2.2 and 2.3.1 }. The state of the pore water can be graphically represented by the potential and the velocity fields. The potential, which can he represented by the hydraulic head, can also imply suction in unsaturated regions. The representation with streamlines is only adequate in case of stationarity, otherwise the paths of water deviate from them. Two families of parallel section planes are required for spatial states. Simplified representations are obtained with sectionally assumed distrihutions and resultants of water pressure and flow, which can he used for estimating forces at the grain skeleton and for taking into account the mass balance. 2.1.2
Conditions of stale
Apart from extremely rapid changes of state, which arc only encountered with blasting, local (i.e. in regions covering a few grain sizes} pore liquid and gas are practically in thermodynamic equilibrium [1]. The ideal gas equatinn
(2.6) in which the gas constant is R can be used if the gas quantity n 1Yf. expressed in mol, of the gas volume VI{ is known. The Vg of a closed soil sample can then he determined (e.g. by the pyknometer test}. One can also calculate the volume or the pressure of gas huhhlcs with Eq. 2.6 if these remain closed and the gas quantity is known, i.e. if also the fraction of gas dissolved in water is allowed for. Pore water is incompressable in the pressure range of soil mechanics. Its pressure pl)! is bounded from below: At the interfaces hctwecn water and air the equations of Kelvin 1 Pg - Pw = RTv;;; ln(l/1p)
(2.7)
and of Laplace (2.8) hold. Where 1p = Pvl Pv.1 is defined by the vapor pressure Pv and its value P~s for vapor saturation as relative humidity, Vw is the molar volume of water, aw the surface energy between water and air, and ru, is the resultant curvature of the interface, given by
2/rw = 1/Ri
+ 1/R2
(2.9)
with the principal curvature radii Ri und R2. lf gas penetrates into the pore space hetwcen grains of the mean size ds (capillar.v entry) the curvature has the order of magnitude
214
Gerd Gudehus
l/rw ::::: 10/d.n which with aw ::::: 7 · 10-5 kNin- 1 leads to the estimate Pg - Pw ::::: io- 3 kN n1- 1/ds. The grain pressure of a humid soil at a free surface under attnospheric pressure (p = Pa) is therefore (2.10)
For instance, fine sand with ds ::::: 10-4 tn has a Ps ::::: 10 kN m- 2 at capillary entry, which is visible from the change of colour. ln the soil interior an increase of water underpressure leads to the appearance and growth of gas bubbles. Depending on the grain size, negative Pw - Pa values (and corresponding grain skeleton pressures without total pressure) can reach between ahout 1 kPa for coarse sand and l MPa for silty clay. Even higgcr underpressures can act near the contacts of extre1nely fine soil particles. Eq. 2.7 gives the humidity 1.f' at which as many water 1nolecules evaporate and condensate at the same ti1ne. With RT /vw = l, 3 · 105 kPa and using the nun1bers given above for p 8 - Pw, humidities ovcr99 (Yu arc obtained for grain sizes over io- 6 m. Soil surfaces exposed to the air are therefore seldon1 saturated. The grain stress tensor T.1. of a 1vater saturated simple grain skeleton, which may also contain gas bubbles between the grains, is related to the total stress tensor T, the water pressure Pw and the unit tensor 1 via 1
(2.11a)
Ts =T+ Pw1 With Ps = -Ts and P = - T this can he replaced by P 1 =P-pwl
(2.llb)
(or conventionelly a' = r1 - u with the effective pressure a' and the total pressure r1 ). Ps is therefore a partial pressure tensor. Eq. 2.11 holds as long as the nlechanical behaviour of the grains is independent of the pressure of the surrounding water (the frequently given explanation, with a wavey intersecting surface passing through grain contacts. is not necessary).
For an unsaturatcJ simple grain skeleton with gas channels through the soil one can write instead of Eq. 2.llb P.1 = P
+ Pc.1·1
(2.12)
with the capillary grain pressure Pcs- As geometrically sin1ilar sitnple grain skeletons have geometrically sin1ilar distributions of pore water for the same degree of saturation, Pcs by Eq. 2.8 has the order of magnitude aw/ds. Where Pcsd~/aw depends on the degree of saturation Sr and the void ratio e as shown in Fig.4. It therefore reaches a maximum at
1.5 Constitutive laws for soils fro1n a physical vicv-.:point
215
Sr :=:::: 0.5 and vanishes for S, = 0 and Sr = Srg. The upper bound Srg for a1most no throughgoing gas channels, is not sharply determinable. The assumed uniform distribution of pore \Vatcr can at best appear in simple grain skeletons after a long \Vaiting time or \Vi th higher te111perature. Gas pockets appear in the pore space (Section 2.1.3) after rapid v.'etting or drying. The 111ole fraction mg of dissolved gas in the v. atcr increases with the gas pressure p8 approximately by Henry's law, 1
(2.13)
mg= Pg/Kg
v. ith a factor Kg depending on the kind of gas and proportional to T. Eq. 2.13 holds also for partial fractions and pressures of a gas mixture, but it has to be replaced by a non-linear relation for high pressures. One can then estimate ho\1.-' the gas quantity is changed v. ith the change of pressure in gas bubbles fixed by the grain skeleton if equilibrium is achieved between the bubbles by diffusion in the water. Eqs. 2.7 and 2.8 also hold true al the same time. Salls arc dissociated into ions after solution in the pore \1.-'ater. With the 111olc fraction rn; of one species of ions the osrnotic pressure al an impermeable membrane is obtained by van 't Hoff's lav.' as 1
1
Pm = RTm;
(2.14)
For example, Pos :=:::: 2 MPa holds for a solution of 3 o/o by weight of con1n1on sail. As the interface between water and gas is impervious for ions one has instead of Eq. 2.7 Pg - Pw =RT
[v~ 1 ln(INJ - m;]
(2.15)
v.'ith ions, whereas Eq. 2.8 rcn1ains unchanged. Osmotic pressure and underpressure due lo evaporation are con1pensated for at a plane interface in thermodynamic equilibrium. The determination of the porev, ater underpressure with Eq. 2.7 lcads to an overestimation if the ion concentration is neglected. 1
The difference limi replaces rni in the case of different ion concentrations on both sides of a 111e111brane which is impervious for ions. As n11 is mostly bigger at solid surfaces, for compensation of their charge deficit. than in free pore \Valer, an intcrgranular osmotic pressure Pp arises analogously v. ith the capillary grain pressure, which has to be added in Eqs. 2.11 or 2.12. lhis is the intergranular fraction of the Slvelling pressure v.'hich can be detennined experimentally v. ith a fixed saturated grain skeleton by changing the ion fraction in the mobile pore \1.-'ater, i.e. with unchanged bonded ion fraction. A second osmotic pressure acts inside the grains and can lead to the decay of clay particles after a reduction of the ion fraction in the pore water. We cannot deal here with the difficult and partially unclarified details of the ion layers of minutest particles. (Apart from the osn1otic swelling, there is an elastic one, due to rebound of the grain skeleton and a hygroscopic one, due to reduction of the capillary grain pressure. The swelling due to expansion of the gas bubbles has to be added.) The stress cotnponents of a sin1ple grain skeleton arc restricted by requirements in certain cases, depending on the void ratio. The reduction of the void ratio \1.-'ith the mean grain pressure can then be represented by a granular phase diagram (Pig. 5) with at least three curves [10]. p,1 is normalized by the so-called granulate hardness hs. 111e biggest void ratio 1
1
216
Gerd Gudehus
Fi~.
S. Pressure-dependent special void ratios of a simple grain skeleton (granular phase diagran1)
ln(pJhJ
ei
holds for an isotropic compression in the case of loosest packing. It can be approximated
hy (2.16) in the geotechnically relevant pressure range. Therein, eio is the ei for Ps = 0. Because of the intcrgranular osmotic pressure eio for clayey soils cannot be determined for Ps = 0, but only hy extrapolation from the anticipated pressure range. The lower bound e; = 0 for Ps ----+ oo makes sense for elastoplastic grains, hut due to grain crashing there is an upper bound roughly at Ps = 10- 3 h 1.• Instead of Eq. 2.16 one can use the conventional relation
(2.17)
for the pressure range wherein thee -ln(ps/ hs) curve is nearly a straight line. ei then has the value eif for the arhitrarily chosen reference pressure Pr, and Cc is the slope of the straight line. Eqs. 2.16 and 2.17 hold, with e0 instead of e; for uniaxial compression, as in the oedometer \Vi th e0 < ei. The ratio of principal pressure components is then a material constant Ps2/P.d =Ko
(2.18)
called coefficient of earth pressure at rest. The relation, \Vhich can be approximated hy Eq. 2.16 or 2.17, holds also for other proportional compressions and the stress components have constant ratios as given by Eq_. 2.18. These conditions of state hold asymptotically for sufficiently large deformation. The asyn1ptote can be unattainable however for too low an initial e due to grain crashing. The curve ec holds for critical states and can be approximated by Eq_. 2.16 or 2.17, replacing the subscript i by c. These states arc achieved hy constant-volume stretching with constant rate. They are also characterized by a fixed ratio of principal stress components which can be expressed by (2.19)
with the critical friction angle f./Jc for cylindrical shortening. Other stress ratios depending on
1.5 Constitutive laws for soils from a physical viewpoint
217
arc nearly the same. If pressures and not stretching rates arc prescribed, critical states are achieved asymptotically, without shear localisation, only if the void ratio is above the critical value related to the given pressure so that no dilatancy occurs (Section 2.1.3). The lower curve ed in Fig. 5 represents the void ratio for maximum densification by rearrangement under constant grain pressure p 1.• This can also he approximated by Eq. 2.16 when replacing the subscript i by d. The value edo comes close to the conventional densest packing for coarse-grained soil, whereas er:O comes close to the void ratio of the loosest packing. One can replace e by the pressure-corrected density index (2.20)
IJ = (e,. - e)/(e,. - CJ)
wherein ec anded depend on p.1.• The quantities explained in Fig. 5 arc appropriate for coarse- to fine-grained soils as long as Ps substantially exceeds the intergranular osmotic pressure (and therefore not for colloid mud). One can define the relations given above with the granular phase diagram as granular equations of state, hut they arc essentially different from thermodynamic phases and equations of state. Whereas thermodynamic equilibria in closed systems arise spontaneously, simple grain skeletons can reach the na1ned states only after certain deformations, and waiting times are less important (Section 3.2.3). A pressure-free grain skeleton can have void ratios between eio and edo, depending on the previous deformation. For small pressures Eq. 2.16 can he approximated hy (2.21)
\Vhere the subscript i has to he replaced by c or d respectively. This po\ver la\v is typical for a percolation that occurs due to the flattening of the contacts with increasing pressure. The exponent 1/n can be empirically related to the grain roughness. Other than a solid, a pressure-free simple grain skeleton is at its existence limit. For acertain p 5 , void ratios outside the range between ei anded cannot be reached by rearrangements of simple grain skeletons. Very loose skeletons with e > ei, which can have a metastable existence with gas bubbles and channels, are densified by any rearrangetnent and therefore tend towards total liquefaction. Grain skeletons with e < ed can arise hy
cracking or rock or by compression beyond
ed
and decompression thereafter and are
extre1nely dilatant like dry masonry. They cannot undergo the uniform changes of stress and shape assumed for sitnple grain skeletons. The pore water restrains the states that are covered by the granular phase diagram. For full saturation, and also with gas bubbles enclosed between the grains, one can work with the grain pressure given by Eq. 2.llb. The relations are more complicated for saturated clayey soils in the range of low grain pressures hccausc of the osmotic pressure. As a result of the low granulate hardness (hs < I MPa) the grains can substantially change \Vithin the geotcchnical pressure range. For low degrees of saturation, which cnahlc through-going gas channels, one can at best assume sitnple grain skeletons for grain sizes above about 10- 4 m (Section 2.1.3). Boundary conditions may be briefly touched here~ as they have to be cotnpatible with material properties. The grain skeleton pressure vanishes at the transition from soil to free water, and the water pressure or the entrance velocity of water is prescribed. At a soil surface exposed to air the pressure relations explained by Eq. 2.10 hold. The rate of
218
Gerd Gudchus
evaporation~ which is equal to the filtration velocity nu:(Vw - ti 5 ) in the normal direction, can then be estimated. 111e normal pressure is equal to the gas or fluid pressure outside the soil at an impervious membrane. The same holds true for a membrane supporting the grain skeleton but not water, c. g. a filter-cake, and the grain skeleton pressure is ohtaincd from the water pressure. Inside the membrane the seepage force provides for the increase in the grain pressure. Complete expressions for pressures and shear stresses are not possible for interfaces between soil and solid structural parts and forces in the latter can at hcst he related to the resultant soil forces. If an interface hct\vccn a structural part and soil acts as shear zone its state can be estimated (Sections 2.1.3 and 4.3).
1l1e initial slate for the calculation of deformations and stahility cannot he completely determined from all the named conditions and good measurement results, let alone special cases such as at rest pressure and statically determined limit stress fields. A purely statical construction by means of assumed distributions, stress ratios and principal stress directions, is justified in simple cases but in more difficult ones is unclear and too cumbersome, as it could scarcely be computerised. One is therefore left with the construction of initial states \Vith the aid of conservation laws and constitutive laws and with simplified initial and boundary conditions assumed to be valid prior to the start time. A fictitious initial state is frcqcntly assumed \Vithout gravity and the dead weight is then imposed step by step. A younger section of the mechanical history can be followed up with differential constitutive lav-.·s, where the erasion of traces of earlier processes may be presumed. l11is makes no sense \Vithout measurement values (at least of the void ratio and the degree of saturation).
2.1.3
Fabric
The granulornetric fine structure can substantially deviate from a simple grain skeleton. A sand\vich soil can be dcscrihcd as a composite of simple grain skeletons to realistically model mechanical properties in large regions. 1l1e order of layers is topologically lost by intensive rooting and fahrics \Vith lumps and macroporcs can arise instead of simple grain skeletons. Shear bands in simple grain skeletons arc recognized from a higher void ratio than the surroundings in a zone of about 3-30 grain diameters thickness. Inside the shear bands the grain stress tensor is no longer symmetric and an additional couple stress occurs.
The fluctuations of grain contact forces are substantially bigger than outside the shear band. Macroscopically measurable and relevant state quantities of the earth body arc still only position, mean void ratio and normal and shear stress along the shear band. Special consideration is required for shear zones with a granulomctric fine structure as, c. g., faults with smeared clay seams or bentonite Illms in slurry trenches. l11e state of simple grain skeletons can be more precisely dcscrihcd v-.·ith the aid of fabric tensors than with only the grain skeleton tensor and the void ratio. On the one hand one can describe the orientation of particles or contacts. The prohahility density for a direction n of principal grain axes or contact normals can then be approximated by p(n) = n'Sn
(2.22)
where S is normalized by
f
p(n)dn = 1
(2.23)
1.5 Constitutive lai,\'S for soils from a physical viewpoint
219
as any direction will certainly appear. Alternatively one can try to describe the directiondependent irregularity of the distribution of contact normal forces by statistical measures. Approaches for internal variables within elastoplasticity (Section 3.2.2) and hypoplasticity (Section 3.3.3) arc as yet heuristic i.e. such tensors are not yet well settled in a micromcchanical-statistical sense. The question is therefore open on how far force fluctuations in all length scales appear and are fractal. The usual separation of fabric-dependent or inherent and stress-dependent or induced anisotropy, points to two different fabric tensors. The pore water in unsaturated grain skeletons is rarely distributed so regularly as assumed in Fig. 4. Gas pockets or channels arise with the rapid entry or departure of water, especially in the case of fluctuating void ratios, so that the soil is inhomogeneous in respect of the degree of saturation. The spreading of a saturation or desiccation front is mechanically instable, whereas a non-homogeneous pore water distribution can be thermodynamically metastable [lJ and remain so for a long time. A spatial mean value of the capillary grain pressure Pei follows with Fig. 4 from the non-uniform distribution of the degree of saturation. With coarse-grained soils down to silt one can get along with empirical spatial mean values Pcs· Soils with very fine grains and crack systems can be approximated as composites. Two pressures arc often used in the mechanics of unsaturated soils: the net pressure p - PR and the suction p(J - Pw· This is empirically, but not physically justified. Only the grain pressure is relevant for stiffness and strength but it can be as irrcgulary distributed as the pore gas. Soils with lumps and rnacropores can be considered as composites of simple grain skeletons with capillary action. Lumps result from mechanical partition of capillary bonded soils and arc often nearly saturated. One can consider a fabric of lumps to be a simple grain skeleton with a very low granulate hardness determined by capillary suction. If the capillary suction gets lost by drying or cracking, or by access of water, a fabric of the lumps becomes a collapsible structure. Flooding of the lump fabric or gas production can lead to gasfillcd macropores which are larger than single grains. Such a soil can be metastabile by capillary action or weak cementa ti on, but can be transformed into a suspension by a small disturbance and then flow catastrophically.
Polydispersed soils can also be considered as composites of simple grain skeletons. ln the case of very big differences of grain sizes one has to differentiate between large grains floating in a skeleton of finer grains or forming a skeleton themselves, for which the granular phase diagram is useful. Gas-filled macropores have to be allowed for: like floating coarse grains they can be understood as highly compressible inclusions without strength. Cemented grain skeletons have solid condensate bridges at the grain contacts. Simple grain skeletons of this type obtain a prestress comparable to a capillary grain pressure, which gets lost irreversibly during rearrangements. Grain fabrics with lumps or macropores can become stable by ccmcntation, but they are brittle. The cementation often plays only a minor role for soils, whereas it is decisive for rock and concrete. Cracks appear in soils due to extension. A single crack, e.g. at a structural part or behind the shoulder of a slope, appears as free soil surface in the description of state, in general with water pressure. Crack patterns are of a fractal nature: a primary pattern first arises, with big distances in between, then a secondary one appears etc. One can assume simple grain skeletons only in regions which are not yet cracked; a composite of the same type appears to be a more suitable model because of the variable crack width.
220
2.2
Gerd Gudehus
Changes of state
2.2.1 Changes of simple grain skeletons The velocity of fictitious marker grains is described by v5 the velocity the stretching rate [15] is given by D = (Vv,
= dxs/dt.
From the gradient of
+ v, \7)/2
(2.24)
and the rate of rotation W = (Vv, - v_, \7)/2
(2.25)
are derived. (For the propagation of transversal elastic sinusoidal \\'aves, the shear strain y = Vs/Cs is obtained from Vs and the propagation velocity Cs and from it the change of shear stress L\i: = Gy (Section 3.1.1). This formula suffices for estimates even in the case of slightly anelastic behaviour.) The displacement or change ofposition is Us=
I
(2.26)
Vsdt
(')
where the symbol (s) denotes a material point which is unchanged during integration. Geotechnically relevant displacement paths are often markedly different from straight lines, and their gradients cannot always be determined. ff the displacements are small they can be described with the tensor E = (Vu,
+ u, \7) /2
(2.27)
and with D
=
(2.28)
dE/dt
In relation for the grain pressure tensor the shortening tensor and its rate (S = -E, ~ D) are used. 'fensors for large deformations are at most required for the evaluation of tests \Vithout rotation of principal axes; then Eq. 2.28 holds for the logarithmic stretching tensor. The coherence of a grain skeleton can get lost at a gap close to a structure or in the soil interior. Deformation tensors become insufficient if different granulon1etric fractions are mixed or segregated. The temporal change of grain stress is called grain stress rate 'i's = dTs/dt. For rotation rates \Vhich are large compared to the stretching rate and in the vicinity of limit states \\'ith principal axis rotations, one has to replace Ts by an objective, i.e. a frame-indifferent co-rotated stress rate, e. g. the one by Jaumann,
S=
(2.29)
f.=t+WT-TW
Alternatively one can use the grain pressure rate Ps = -T.1.• A simple grain skeleton remains homogeneous in a so-called elen1ent test \\ hen its state is altered by changes of stresses or displacements at the boundary. Strictly speaking this is impossible even \\'ith initial homogeneity: pressure can only be imposed via membranes, displacement via plates, so samples become inhomogeneous at the boundary and along the edges. For an ideal element test grain displacements and contact forces would have the same fluctuations at the boundary as in the interior. This can at best be achieved 1
221
1.5 Constitutive lav,.·s for soils from a physical viev.-·point
by granular dynamic calculations, but these can only give qualitative results because of inevitable simplifications. Even then the homogeneity is not always preserved, as fluctuations can spontaneously grow locally. With such localisations, leading to a new fabric, the grain skeleton cannot remain simple (i.e. mean values do not suffice), but it can become simple again under suitable boundary conditions (i.e. inhomogeneities arc 'ironed out'). Sections of the grain skeleton behaviour can be represented by conventional diagrams (Fig. 6). In the case of axi-symmetric deformation, i.e. s2 = s3 without principal axis rotation. and with Ps2 = Ps3 for axi-symmetric grain skeletons, Ps2 is often kept constant (conventional drained triaxial test). The axial pressure Psl and the volume change~ V/Vo can then be plotted against the axial shortcnings1 (a and b), or the pressure difference Psl P~·2, the pressure ratio p 5 1/Ps2 and the void ratio e. For constant-volume axi-symmetric deformation (undrained triaxial test with saturated sample) another evolution of Psl is obtained with the same initial state of the grain skeleton and instead of the now constant e, the pore pressure Pw is plotted (c and d). ror a uniaxial deformation, as provided by an ocdomctcr, the axial pressure Psl is plotted against the axial shortening s1, ore versus ln(p5 1/Pr) with a reference pressure Pr (e and f). For shearing. with constant vertical pressure or constant void ratio, the shear stress L and the volume change, or the pore pressure is plotted against the shearing angle y. The diagrams are similar to the ones for axial syn1mctry. The influence of the deformation rate or the waiting time under constant pressure is thus not allo\\'ed for. Definitions and material properties arc often derived from such diagrams, following concepts used for solids, \\'hich can lead to misunderstandings. dps1/ds1 or ~Psi/ ~s1 is defined as the deformation modulus Ev for triaxial and as the stiffness modulus Es for ocdomctcr tests. Both arc so variable that they are of use only in special cases (Section 3.1.2). The notions loading, un- and reloading arc linked with special processes and can only be generalized to arbitrary changes of state \\'ith certain constitutive relations (Section 3.2.1).
Psi
p,,
Psi
R.
p,,
p,, p,2
a
s,
c
s,
t.V
·~
v 0
Pa
s, b
e
ln(p,lp,)
d
.Fig. 6. Working lines tor monotonous cylindrical detorn1ation with conshint lateral pressure (a and b ), constant volume (c and d) and constant ni
222
Ger
The so-called preload pressure, i.e. the biggest pressure ever itnposed, cannot objectively be generalized and is not a state quantity. The lateral pressures Ps2 in the ocdomctcr test and p 1.2 and Ps3 in the shear test are difficult to measure and are usually not considered. Strength quantities arc nonnally derived fron1 stress components for limit states defined hy dpl'I /d.~1 = 0 or dT/dy = 0. lbe so-called drained and undrained shear strength parameters, q, 1 and c' or r:p, 1 and c/J., are justified for conventional stability assessment, but they depend on Ps, e and D and are therefore not material constants. lbe volume change under cylindrical shortening with constant lateral pressure is given by the measure of dilatancy o = -d(s1 + 2s,)/ds1. The angle of sliding up v = arctan(dc/dy) is determined correspondingly by sin1ple shearing with constant vertical pressure. One can also consider cvntractancy for negative 6 or v l5]. For a drained triaxial or shear test with constant lateral or vertical pressure, 0 or vis largely negative in the beginning (densification), then positive (loosening) up to a n1axin1u1n close to the limit state. 0 = 0 or v = 0 holds for critical li1nit states and often a critical friction angle f!!c, independent of Ps, is observed (Section 2.1.2). A non-dilating grain skeleton can remain homogeneous up to the critical state, otherwise ec arises only in shear hands. 111e grain skeleton hchaviour is more completely represented by associated paths of stress and deformation, which are possible for a plane of axial symmetry (Fig. 7). TI1e paths are three-dimensional for biaxial or cuboidal deformation (s2 = 0 or s1 i- s2 i- s3 in general) and thus require a projection to two non-parallel planes. This is not shown here although test results with this representation have substantially contrihuted to the evolution of
s,
s, 2
A
s,
A
2
2
I
B 2
0
Sz
Sz
B
2 0
S2
p,,
p,,
p,,
2 °A
II : 1
! 1
'
: 0 .. ··· ·---
/r )t.· 0
..
:
0
'··
0
A~ 0
2
p,,
a
1
p,, b
P<2
c
Fig. 7. Associatt:
1.5 Constitutive laws for soils from a physical vit:-wpoint
223
constitutive laws. Four-dimensional spaces would be required for shearing because of the additional rotation of principal axes and a twofold projection for representation in planes, but the limited and disputable test results do not justify this amount of work. Proportional compressions are defined by straight deformation paths v.·ith densificaton (Fig_ ?a). Independently of the initial stress, the pressure path then tends to a straight line the slope of which is detertnined by s1 /s2. The transition to the asymptote is shorter if the initial pressures arc closer to it. As the subsequent behaviour is then determined only by stress and density one can speak of a swept out niemory (6] as internal variables are determined then by the pressures and the void ratio. For sufficiently strong densification, a void ratio independent of the initial state is achieved, as represented in the granular phase diagratn. This is not reached for a too low void ratio, hov.·ever, as the grains are destroyed due to high pressure. For deformation with constant volume ( D1 + 2IJ2 = 0), the stress path tends to a critical state with a constant stress ratio (2.30)
and remains there as long as the grains arc not damaged (Fig. 7b). The mean pressure Ps tends to the critical value related to the void ratio, as given from the beginning and is approximated by Eq. 2.16 or 2.17 (using subscript c instead of i). For e > e,.o the grain skeleton decays and can then be considered collapsible. The void ratio does not remain constant for constant Ps· If the void ratio is initially smaller than ec(Ps) the pressure ratio reaches a 1naxin1un1 which is described by the peak friction angle (2.31)
This depends strongly on the previous path and is therefore not a soil constant. Shear localisation occurs in the vicinity of such limit states. Instead of using paths, the differential stress-deformation behaviour can be represented by response polars (Fig. 8). For unit stretching rates (.~f + 2.~i = 1) with different directions .~1 /~·2, the stress rates (JJ., 1 and 1Js2) plotted from the stress point are ellipse-like curves [7}. The more they deviate from ellipses with the stress point as centre, the less elastic is the behaviour. The diameter, which can be understood as a measure of differential stiffness, increases sub linearly v.·ith Ps fron1 zero, which can be described by the power law (2.32) where h.~ and n are the sa1ne quantities as in Eqs. 2.16 and 2.21, and the factor A depends on the void ratio e. Statical limit states are characterized by having a response polar at the stress point for a certain direction of shortening. Critical states lie on the tv.·o straight lines described by Eq. 2.30, whereas other statical limit states can at best be approximated by straight line sections depending one. The representation is only sufficient if the state at a stress point, which is reached by the previous history, is the satne for all rates. For cycles this is not achieved by the void ratio in addition to the grain pressure, v.·hich is represented anyway. Response polars are good for judging and comparing differential constitutive laws. They can principally be extended to more than two different components and can then be represented by projections onto planes.
224
Gerd Gudehus
Ps1
B
Ps2
a b Fig. 8. Response polars (a) of a simple grain skeleton for axi-symmetric unit shorlening rates (b)
1,3, ...
p,,
~"-
e
s,
0,2, ...
1,3, ...
~ I
0,2, ... 2
Ps2
b
' Fig. 9. Shorlening paths (b) and changes of void ratio (c) for stress cycles (a) with sn1all (I) and large ampliludes (II)
For alternating loadings a representation with associated zig-zag paths is more adequate than with conventional diagra1ns or response polars, but even for axial symmetry it is incomplete. It is however useful to differentiate between large and small deformation atnplitudes. These become larger if grain stress cycles arc imposed with a repeated approach towards limit states (Fig. 9a). The deforn1ation paths (b) do not then become cyclic. The stepwise increase (accumulation) is bigger for larger amplitudes. The void ratio (c) tends to a nlinin1u1n ed, depending on P~· for s1nall an1plitudes and to a maximum r?c for very large atnplitudes. As long as granulometricchangcs arc negligible, the changes of void ratio and fabric arc cyclic in the asytnptote. and are s1naller with decreasing amplitude, and the accumulation becomes the same in each step. If defor1nation cycles are imposed there is a stepwise change of grain stresses (also a kind of accumulation) and asymptotical grain stress cycles can arise if granulometric changes
225
1.5 Constitutive lav.'S for soils from a physical viewpoint II
.0
s,
1,3, ...
s, II 0,2, ..
p,,
b
a
Fig. 10. Stress paths (a) and grain pressure changes (c) for shortening cycles (h) 1,1,,ith small (I) and large amplitudes (II)
are negligible (Fig. 10). The grain pressure vanishes if the void ratio exceeds edo for small amplitudes and eeo for very big ones (cf. Fig. 5). In the case of water saturation and without filtration, this decay of the grain skeleton is denoted as liquefaction but a suspension capahlc to How can only arise fore > eeo· Otherwise the mean grain pressure Ps, associated withe= ed, is reached for small amplitudes and the one fore= ee with big an1plitudes. Cun1ulative changes of state hecome negligihle for extremely small amplitudes and the grain skeleton hchaviour is then - and only then - elastic. Grains becon1e rounder and smaller by abrasion and fragmentation. Cd and ec are reduced so that asymptotic cycles under alternating loading are reached only after a higger numher of cycles or never. Changes of fabric are not observable from outside and their representation is more difficult. The processes represented by Figs. 9 and 10 are idealized special cases, which arc of use for the development and selection of constitutive laws. Real alternating processes are nlore complicated. State changes of the pore fluid are described hy the pore pressure rate Pw in the case of full saturation. The rate of ftov.1 velocity and concentration of gas and ions is rarely added explicitly. Changes to the volume fraction of gas bubbles among the grains and their pressure - and thus the compressibility of the pore Jluid - can he calculated hy Eq. 2.6, where the change of the dissolved gas fraction has to he allov.:cd for if sufficient time is given for diffusion. The break-through of gas channels (capillary entry) can he estimated with Eq. 2.10, and the closure of gas channels can he recognised by means of the limit degree of saturation S,.R shown in Fig. 4. Such estimates ren1ain coarse as spatially irregular gas channels and pockets arise. 2.2.2 Thermal and mechanical activation Spontaneous changes of state can arise fron1 thermal activation v.1ithout external changes [241. Position changes of ncighhourcd molecules (Section 2.3.1) have a n1ean rate per unit of time given hy
E,,,) kT
kT exp ( -Ea) v=2- sinh ( h
kT
(2.33)
226
Gcrd Gudchus
with the absolute temperature T, the Boltzmann constant k, the Planck constant h, the activation energy Ea and the energy Em enhancing a change of position. The kinetic energy per particle fluctuates maximally and chaotically around the mean value 1.5 kT, as descrihed by the exponential Boltzmann distrihution. (If f,'u and Em are referred to one mol - with 1 kT per molecule~ 25 kJ/mol - they have to be divided hy RT in Eq. 2.33). The evaporation of water requires Ea ~ 9kT per molecule (cohesion energy, [121) and reaches noticeahle rates only at high temperature or big tensile stress. For shear flow Ea ~ kT holds, and Em is often even smaller so that Eq. 2.33 can be linearized with respect to Em/ kT. Newton's linear flow law can thus be justified and the linear permeability relation by Darcy (Section 4.2). The pore water in diffuse interfaces has higher activation energies caused hy caught counter-ions and therefore a lower mobility than free water. Typical solid soil particles have cohesion energies from about 30kT (montmorillonite) up to about 200kT (quartz). Shear dislocations in perfect crystals can reach appreciahle rates only if Em has at least the same order of magnitude. For shear melting, i.e. loss of molecular contacts by sliding, Eq. 2.33 leads to the shear rate 3
D"" Do exp [
(t-c)a ]
kT
(2.34)
with the shear stress t, the molecular cohesion c and the molecule distance a. The shear rate for"= c is Do = 2kT I h. ca 3 which corresponds to the deviatoric deformation energy during plastification associated with one molecule (Section 3.2.1). Soil grains have dislocations and eigenstresses, so that the spatial mean of Em is already substantial without external action. One can consider Eq. 2.34 as a viscoplastic constitutive relation but then has to replace c and lJo hy empirical quantities, which are difficult to be estimated from molecular dynamics. Eq. 2.34 is nearly valid for insular solid hridges at grain contacts where the rate of position change fort= c is Do""kT cd 3
/%'-p
(2.35)
with the island sized, the shear modulus G and the density p. Eqs. 2.34 and 2.35 hold even for variahle contact islands as only the islands near the activion threshold are relevant [24]. Changes of state of simple grain skeletons due to thermal activation therefore cannot be sufficiently allowed for in general, as rearrangements and grain pressure changes modify the contact properties. Monotonous processes with practically constant stretching rate are an exception. The mean grain pressure p.1 then increases with the amount of stretching rate LJ nearly via (2.36) where Psr is the amount of Ps at the reference stretching rate D = lJr. P.ir depends on the void ratio e from Eqs. 2.16 or 2.17 with the subscript for the axial stretching direction. The number of activated contact molecules per unit volume is proportional to Psr (16, 19J. The viscosity index Iv is ohtained from tests with widely varied stretching rates hetween ca 0.01 for quartz to 0.07 for montmorillonite. One can estimate the contact island sized from Eqs. 2.34 to 2.36 with the aid of the molecular cohesion c and with Iv.
l.S Constitutive l:=iws for soils from a physical viewpoint
227
Thermal and mechanical activation usually act together in grain skeletons. A kind of creep relaxation can arise due to irregular vibrations and can evolve similar to the thern1ally activated one. Thermal activation leads to rearrangements and acoustic emissions, causing further 1nove1nents in a chain reaction. Granulometric changes are mostly thennally and mechanically activated. This is not fully achieved with contact mechanics., starting from fracture mechanics., as spherical grains are then assumed. Approximations with Eq. 2.34 do not lead very far as in each case a wide spectrum of particles, activation energies and tnodes of motion is involved. The 1nechanical part of activation tends to be bigger for hard-grained and the ther1nal one for soft-grained soils., but the quantification still remains heuristic-en1pirical (Section 4.1).
2.3 Special sequences of state and stability 2.3.1
From molecules to soil elements
To visualize thern1al activation and some stability properties of molecular systems, we can consider a frictionless 1nagnet suspended ahove a row of repelling magnets adhering to the base (Fig. 11 ). The horizontal force at the pendulum is obtained from the gradient of the potential distribution. There are two potential minima near the row of magnets where the position of the swinging magnet is stable. The position is labile at the saddle point in between, elevated by the an1ount E 0 • The pendulum moves chaotically after disturhances of the base or of the suspension points. The bigger the disturbances are, the nlore frequently the position changes frotn left to right. There is a gravity-induced energy difference Em between the two energy depressions because of eccentricity of the suspension point. The position in the higher depression is tnetastable co1npared to the lower one. The kinetic energy is Boltzmann-distributed in the case of maximally chaotic 111otion and the changes of position are quasi-thermally activated (Eq. 2.33). The cquilihrium position is stabilized (i.e. the rate of position change is reduced) by increase in Ea and/or decrease in Em, and destahilizcd in the opposite case. One can speak of active stabilization if the magnet gate gets narrower or the hase slope gets s1naller. the opposite is active destabilization. E 0 and Em can he changed passively by repeated disturbances, which gradually change the position of the base magnets and the inclination. In our model system, however, the gate can only get wider spontaneously and the inclination s1naller. This system with two degrees of freedom already enables chaotic motions but is unrealistically restrained by the pendulum rod. Systems with 1nore degrees of freedoms and particles cannot be imagined mechanistically and can only be represented graphically in a restricted sense by projections. One can imagine a grid pattern of base particles representing a n1ono-1nolecular film with adsorbates above, moving in a field of attractive and repulsive forces [24]. The base pattern can he irregular due to dislocations and can change with titne. The energy distrihution of an adsorbate changes with its distance from the base and with other sufficiently close ad
228
Gerd Gudehus
•
•
c
•
o••
• •
Fig. 11. Magnet pendulum over row of magnets (a and b) with energy distribution (c and d, darker means deeper)
A stochastically homogeneous collection of particles is obtained by unlimited continuation of the pattern and its adsorbatcs. The mean particle distance correspond s to density, and the mean repulsive force to pressure. The latter is also determined by the distance from the base. Depending on temperature, i.e. mean chaotic kinetic energy, the two-dimensional gaseous, fluid or solid phase is thermodynamically stable or not. For constant pressure, stability is indicated by an increase of the free energy (for constant density free enthalpy) in the case of a phase transition, for instability the opposite holds (1]. For numerical simulation a section is considered at the rim of which particles enter and leave. Only with a careful selection of interaction potentials and the inevitably fictitious boundary conditions can realistic equations of state and statements on stability based on them be obtained.
With transition zones phase limits can also be modelled, e.g. between a liquid and a gaseous region. Thus surface energies and capillary pressures dependin g on the curvature of the interface are obtained (cf. Eq. 2.8). Transition and surface energies vanish at critical points so that phase transitions occur spontaneously, and the transition limit gets fuzzy. The critical phenomena occuring then imply the spontaneous formation of fractal capillary structures and the approximation of equations of state in the vicinity of the critical points by power laws with universal exponents [3] .
In grain skeletons fixed by filter plates, pore water and air including dissolved fractions and vapor, can have a variety o f metastable equilibrium states for given pressures and
1.5 C'.onstitutive laws for soils from a physical vie\\:point
229
mass fractions. Those with the lowest total energy are the most stable ones and correspond to the most uniform distrihution in phase space. For some neighhouring metastahle states the transition energy vanishes and critical phenomena appear. Other metastable states with non-uniform distrihution can exist for a long time because of high energy harriers_ Experiments and molecular dynamic calculations can yet only give an intuitive impression of this diversity, hut not a quantification. One therefore has to use spatial averages and can at best crudely estimate systematic inhomogeneities. Granular dynamics has not very much in common with molecular dynamics. Irregular grains are changed by each encounter so that they are not permanent, and their interactions are not energy-conserving. One is left with heuristic approaches for modelling grain changes and interactions. Concepts for the dynamics of conservative systems up to thermodynamics cannot he transferred to grain skeletons. This holds also true for notions of stahility, which at best can have a similar meaning in a heuristic sense. A simple grain skeleton which is homogeneous in a statistical sense is called a soil ele1nent and may contain statistically homogeneously distributed pore fluid. A group of grains is postulated with changes of state, which are sufficiently descrihed by evolution equations for mean values - called constitutive laws. Within the framework of this simplification some of the stability notions introduced above can he heuristically transferred. The notions temperature, entropy and total energy are of little use, but density and pressure remain useful quantities. The state of an element can be practically stable if its changes of state due to thermal and subsequent mechanical activation is negligihle within relevant periods. This notion is however suhjective, as the neglection rests ·on arbitrary tolerance criteria. A soil element undergoes a stabilization in this sense if the density increases for a given grain pressure p 5 , if the relative mean shear stress T.m/P} is reduced or if the mean grain pressure p 1. increases for a given mean shear stress T.m. Stabilization is also obtained hy equalization of inhomogeneites (fluctuations) of grain positions and forces. The internal variables which are then required are heuristic. The distribution of pore water in unsaturated grain skeletons-which cannot sufficiently he described hy mean values - is an example of such inhomogeneities. The stabilization may be called passive, if it is spontaneous due to thermal and subsequent mechanical activations. It is recognizable from a decreasing rate of changes of state. It is active, however, if it occurs due to changes of mechanical and hydraulic boundary conditons. The state of a soil element is indifferent if it does not change under a permanent stretching rate. It is called critical in soil mechanics (there are other critical states in a physical sense, Bruce and Wallace [3]). Granulometry, void ratio and grain pressure tensor remain unchanged, Eq. 2.16 or 2.17 and 2.36 (with subscript c instead of i) are approximately valid for the grain pressure. One can also consider an indifferent sequence of states if the same state of the soil element is achieved repeatedly under periodical mechanical action (Figs. 9 and 10). A soil element may be called collapsihle if its stretching rate under fictitiously permanent stresses increases with tin1e due to thermal activation only so that kinetic energy arises. Peak states in soil mechanics convention occur if a collapse is prevented by deformation control or if instead, a reduction of shear stress - called softening - occurs. Both definitions are heuristic and neither mechanically nor physically strict. For calculating the energy excess and strictly speaking the change of boundary stresses due to houndary
Gerd Gudehus
230
deformations also has to be allowed for. The softening of saturated undrained soil ele-
ments comes from the increase of pore water pressure, the one for drained soil elements from dilatancy. There is a spontaneous formation of inhomogeneities. The loosening is concentrated in narrow shear zones and the spontaneous pore pressure increase is also localized.
A soil element undergoes a destabilization if it comes closer to an indifferent or collapsible state. lbis may happen passively, due to thermal activation. or actively due to imposed changes of stress or shape. 2.3.2
Earth bodies
For a granulometrically homogeneous or horizontally layered earth body with a free or uniformly pressure-loaded horizontal surface, vertical profiles suffice for describing the state and its change. A series of layers of fine sand and fine silt with a ground water table (definded by Pu, =Pa) somewhat below the surface may serve as example (Fig. 12). The pore gas may be in channels above the ground water table or in bubbles below it. The density index !rl may initially be high above the ground water table due to shrinkage and small below it and - due to flooding of a lump fabric - partially negative. The pore water pressure is smaller than Pa above the ground water table and may be partially higher than hydrostatic below. The ratio of horizontal and vertical grain pressures above the ground water table is Psh/Psv : : : : 1 due to capillary action, whereas Eq. 2.18 with a material-dependent Ko holds below. lbe total vertical pressure Pv is obtained from the specific weights related to S, and !11, and from it p 5, with Eq. 2.llb or 2.12, in the range with gas channels. Such a ground is only slowly stabilized by thermal activation. Pore water evaporates near the surface, except in very high air humidity, so that p.1 and id increase. The pore water flows off gradually from regions with higher than hydrostatic pressure so that Ps and Id increase (primary settlement, finite compression law). Under constant Psv, Id and Ko increase very slowly (secondary settlement, finite creep law). The densification-induced stabilization is strengthened by small mechanic disturbances, e.g. vibrations from traffic. A temporary destabilization can occur with strong disturbances, e.g. by blasting or strong earthquakes, as Pw increases due to prevention of contractancy (accumulation formula or differential constitutive law). Collapsible soil regions (Id < 0) can become liquid (p,, = O). and small geysers and volcanos can arise from the eruption of suspension bubbles. A subsequent stabilization with a Pw-reduction and a Id-increase is thus substantially accelerated compared to a primary consolidation.
S,
la
a
1
P...!Pa 1
PstlPsv 0
1
~
\ a
b
c
d
e
Fig.12. Partially percolated clayey layers (a), profiles of degree of saturation (b ), density in
1.5 Constitutive ls'h·s for soils from a physical viewpoint
231
The gas channels are closed if a water quantity corresponding to the gas volu111e fraction is added, so that Pw increases and therefore Ps decreases. As the saturation front propagates irregularly and can enclose gas pockets, the implied softening is not sufficiently covered by considering clements and assuming hon1ogeneity. Lump fabrics can be densified by added water due to crashing of lumps, and can get te111porarily liquid after rapid flooding. With hard quartz grains only a small elastic swelling can arise from a p 5 -reduction. Fabrics of clay particles swell much 111ore as the suction, which is lost by wetting, is much higher and as os111otic pressures play a role. With shrinkage cracks the softening is so irregular that the stability cannot be judged with spatial averages. Specific weights and hydrostatic water pressures do not always suffice for determining p 5 , and Id and p 1. arc not uniquely associated. The pressure ratio Psv! Psh is not constant: It can be close to 1 due to capillary suction, hut also exceed Ko due to tectonic shortening or temporary surcharge and then go back because of thermal activation or repeated mechanical disturbance. The stability cannot be judged with soil elen1ents under fictitious permanent lateral pressure. The state of a series of layers on an inclined hases can also be represented hy profiles if it is thin compared with its lateral extension (Fig. 13). However one has to add the velocities of the pore water and the grain skeleton driven by it, parallel to the slope. A layer series undergoes stationary creep, its state is indifferent therefore ir a critical state exists closely ahove the base. One therefore has (with the stress components by Eqs. 2.2 and 2.3) tn/ Psn = tan (Pc and (1 = 1(Section2.1.2). tn and the total pressure p 11 follow from the weight and the inclination, with Pw fro111 the water level follows Psn = p - Pw. tan(fc is - as for sliding friction - independent of the shearing rate (here [) =cos~· Ov.v/Bz). The critical void ratio ec increases due to ther111al activation however, which can be allowed for in Eqs. 2.36 and 2.16 or 2.17 (subscript c instead of i). D- and e-profilcs can then be transfonned into each other based on triaxial or shear tests with D =Dr and adapted to match the observations. Evaporation or drainage leads to a stabilization as Pw goes down, P.i· goes up and e down. Because Tn/ P.rn < tan 'Pc, D gets smaller and s111aller and p 11 • changes have to be allowed for. Conversely, an increase of Pw leads to destabilization, which can he catastrophically accelerated due to localised shear. For soils with bigger and harder grains the thermally activated creep is often negligible within technically relevant periods but mechanical disturbances have more effect. 1,, changes only slightly with Ps' but substantially with shearing during the fonnation or sliding of a slope. If the soil is in a critical state throughout from previous How, it starts to move by even the minutest disturbances, so that the How profile cannot be regular and the s, b
P.IP~
1,
1
0
0
0
1
't,./p,n
0
tsni:p 0
v,)D,h O O,Ot
-.:, ;lb-
f"
~
'"
,,
,,.
'
h
lf b
'
d
e
Fig.13. Inclined soil layer (s), (b) to (c) state profiles according to Fig. 12, (f) profile of slope-parallel velocity
232
Gerd Gudchus
surface cannot remain plane. Starting from an indifferent state, a stabilization is obtained only with substantial changes of position. For very loose packing ( 111 < 0), which can result fronl the flooding of lump fabrics with a low slope angle, even the minutest disturbances can lead to catastrophic increase of pore pressure and velocity, as the expulsion of water due to contractancy is prevented by the low permeability. Densities higher than critical, (!tl > 0) and Tn/ Psn < tan 'f'.'c, lead to shearing and loosening in narrow zones due to small disturbances. Because of shear localization in a nearly surface-parallel zone, the velocity shown above increases rapidly if Ps does not increase due to Pw-reduction because of prevention of dilatancy in the case of full saturation and low permeability. Shear bands can also have other positions resulting in earth blocks that come to rest after a substantial change of position. Even the stability of geometrically simple slopes cannot, in general, be sufficiently judged by conventional earth statics. The critical state concept yields a critical slope angle and the void ratio of the shear zone, in the case of known pore water pressure, but not the creep velocity and its change with pore pressure. One can calculate a grain pressure field, assuming the critical state everywhere, but the origin and stability of it are debatable. With shear strength and pore pressure parameters based on triaxial tests having other state paths, states are constructed which are far from reality and therefore do not enable statements to be made on stability. Three methods are worth consideration for a better judgment of stability. Most expensive is a detailed follow-up of states with initial and boundary conditions and with conservation laws and differential constitutive laws. Generally inertial forces and branchings have to be allowed for and constancy along slope-parallel planes cannot be as.sumed. A simpler method is the calculation of the excess kinetic energy by as...,uming a velocity profile. As the velocity distribution then assumed need not occur simultaneously everywhere as in a kinematic chain the analysis of propagation of plane shear waves with the aid of differential constitutive laws ( Ossinov [23]) gives a better approximation to reality. With very loose packing and full saturation a shear wave causes a transition to flow if the energy criterion indicates a loss of equilibrium. With loose packing and near saturation with gas bubbles., all stability criteria nearly coincide and a shear wave triggers the generation of further waves in the earth body. If the wave meets a dense region it can be held there by the formation of a shear band. Repeated shear waves with amplitudes typical of strong earthquakes can lead to such an increase of pore pressure in saturated soil that shear waves cannot be propagated for lack of shearing resistance and the soil Hows away. Without pore pressure increase, repeated shear waves in slopes with lower than critical inclination cause a gradual change of density and position in the sense of stabilization, otherwise destabilization occurs. A cut which is very long perpendicularly to the plane of representation (Fig. 14) with an initial state profile as shown in Fig. 12 does not behave like a slope. Mass and stiffness of structural parts near the surface may be negligible compared to those of the soil. Neglecting inertia and viscosity effects., one can replace time by the depth of the excavation monotonously increasing with it. Changes of position can be crudely estimated at best, with finite constitutive laws for stress paths estimated for it. They are more reliably obtained with differential constitutive relations. Changes of pore water pressure, due to changed hydraulic conditions and the density index, due to pressure changes and dilatancy, have to be allowed for. Shearing and loosening concentrate increasingly in narrow zones. This localisation is obtained by extended constitutive laws independently of the mesh.
1.5 Constitutive laws for soils from a physical viewpoint
233
A loss of stahility is indicated by a divergence of changes of position (shown hy an arrow in Fig. 14c). In conventional earth statics, such limit states, characterized by neighbouring equilihrium positions, are investigated with assumed slip surfaces and shear resistances along them. It remains an open question whether other limit equilibria, with other slip surfaces and state quantities assumed along them, arise and whether the limit state is indifferent or labile. More revealing are numerical limit load and bifurcation analyses of states changed by excavation. An iterative approach to failure modes (i.e. velocity fields related with the verge of bearing capacity or hifurcation) are required, as this is not an eigenvalue prohlem as for a system with conservative forces. Indifferent or collapsihle states can more easily be recognized by calculating the energy excess with semi-empirically assumed failure modes, which require differential constitutive laws. Filtration flow has to be excluded for caution in stability analyses, the volume changes are thus restricted according to the gas fraction. The stahility is actually time~dependent due to thermal activation. A first approximation is the use of instationary pore pressures, calculated with a fictitious unchanged grain skeleton but with changed hydraulic boundary conditions and degrees of saturation. The coupling of grain skeleton and pore pressure motions is helter allowed for by changes of void ratio from changes of grain pressure and dilatancy or contractancy. This coupling is
a
-1-----7 ,
-d- -
-
-
1
-
----
,
,
~"''
'B
c:V
b
d
I~-~--~
c
d B
'
Fig. 14. Cut (a) with increasing depth (b), changes of position (c), pore water pressure (d). mean shear stress (e) and density index (t) of some points \\'ith time
234
Gcrd Gudehus
simplified in the stability analysis if a filtration flow is excluded. Finally, one can allow for creep relaxation of the grain skeleton due to thermal activation of the grain contacts. This is crudely achieved with finite constitutive laws, but differential relations with viscosity
are better. Weather and pipe bursts can change the soil state considerably, but this happens fairly irregularly. One should prevent development of cracks and softening from drying or wetting by providing a cover rather than follow up with voluminous investigations. Pore pressure measurements are more useful than calculations of the pore pressure evolution in the unsaturated zone if considerable instationary pore pressures arc expected. Currently, changes of state due to variable loads and vibrations can at best be estimated with constitutive laws. Cumulative changes of state due to quasistatic alternating loads can be esti111ated by cumulation laws (Section 3.1.3), which requires a realistic determination of initial and intermediate states. Changes of state from alternating loads are more reliably obtained with differential constitutive laws. Depending on alternation and waiting times, the pore pressure change due to hydraulic-mechanical coupling and the creep-relaxation from thern1al activation can then be considered. For impacts and vibrations, changes of state can likewise be followed up with an allowance for inertial effects. Nearly cyclic changes of pore and grain pressure due to small rapid actions- e.g. passage of vehicles or earthquake waves - can be adequately estin1ated with the theory of elasticity. Cumulative changes of state due to repeated rapid small actions can be estimated with cun1ulation laws if they are small compared to the initial state. For longer waiting times, the thermal activation has to be allowed for but not in a collapse analysis. For strong in1pacts and explosions the theory of elasticity and cumulation laws fail in the range of large deformations, so that changes of state can only be followed up with differential constitutive laws. 2.3.3
Strudural bodies
The explanations shown in Fig. 14 can be transferred analogously to some other geometrical conditions and initial states. This holds in particular for earth bodies produced in layers (e.g. dan1s) and their underground. Other aspects are also induced for structural parts in contact with soil, as shown for the vertical displacement of a rigid cylinder (Fig. 15). Depending on the installation method, the soil state in the vicinity of the cylinder is different from the one without it. In a narrow zone near the surface of the cylinder the soil generally ren1ains in another state, which can be explained with extended constitutive laws. The evolution of the vertical force For the vertical displacement u with tin1e t may be given (actually the evolutions of F and u arc coupled). The development of soil state quantities can be represented as shown in Fig. 14, the interface zone between the cylinder and the soil has to be added. If pore water flow and creep-relaxation can be neglected the representation of the result with a line F(u) is of use. This is roughly obtained with finite constitutive laws based on assun1ed stress or deformation paths and more reliably with differential constitutive laws. The point of bearing capacity defined by d F /du = 0 is almost obtained by conventional calculation of base failure and more precisely by differential analysis of neighbouring equilibrium positions. If the speed of penetration exceeds the permeability by several orders of n1agnitude. the filtration flow has to be neglected in the limit state analysis so that the volume change of the soil is constrained depending on the gas fraction. With low density and high permeability, the limit load point is not reached because of geometrical stabilization from substantial penetration. With very low
1.5
(~onstitutivc
235
laws for soils from a physical viev,rpoint F
F
F
'
'' b
'I d
u F
- - - .- - - Ju a
~
~
" '\ '\
'
'
'
u
Fig.15. (a) Axi-symmetric ground and structural body, (h) force and (c) displacen1ent evolution, forccJisplact:mcnt lines for monotonous (d) and alternating (c) loading. Dashed for higher penetration velocity.
density and low permeability a collapse-like sinking after the limit load point is possihlc, as can be seen fron1 the energy excess with the aid of a state field, an assumed failure mode and a differential constitutive law. In the case of high density shear bands grow from the edge of the cylinder so that the bearing load is reduced, a collapse becomes possible if it is exceeded. Similar diagrams are of use with thern1al activation if the velocity of penetration is nearly constant (dashed in Fig.15 for nearly tenfold velocity). Due to the i ncrcasc in the resistance to rearrangement from Eq. 2.36, similarly inc1cascd penetration resistances arc ohtaincd. For force control, an increase in displacement under constant force is nearly obtained with finite creep laws and more precisely with differential constitutive laws with viscous extension. If the penetration velocity is much higher than the pcrmcahility the penetration is delayed hy restricted compression and the temporary excess pore pressures related to it, which can be approximately calculated with finite constitutive laws (conventional
consolidation theory) and more precisely with differential constitutive laws. A delayed collapse (creep rupture) can occur in strongly compactablei rather impcrmcahlc soils. fOr alternating loading, the penetration with equal maximal force is bigger than for monotonic loading. For constant pore water pressure this is due to additional dcnsification, with constant density due to saturation and without filter How because of pore pressure increase, which in both cases is justified hy a stronger tendency for contractancy than dilatancy. Cycles with small amplitude can be followed up with the theory of elasticity and small changes of state from cycles with accumulation formulae. fOr a large intended penetration - e.g. vibratory driving - this is no longer sufficient and the soil hchaviour near the cylinder has to he covered by interface elements (Section 4.3). Diagrams such as Fig. 15d and e do not suffice for structural bodies with n1ore than one degree of freedom in representing the evolution of states and for the analysis of stability. One gets farther with force and displacement paths as shown in the example of a rotating and penetrating block (Fig. 16 ). For monotonous increase of forces and displacements (I), limit states can he reached (dashed in Fig. 16b) which arc characterized by the existence
236
Gerd Gudehus
r-------
3 _,.
__ 1
f _ .......... 2
/
/
/ 2 I
I
II
I
a
1
3
3
''
2 I
0
0
b
c
M
----~r'
t
-----..i' ____
2
II
J_
J
u T
Fig.16. Force (a) and displacement paths (b) of a rotating and penetrating block (a)
of differentially neighbouring equilibrium positions_ These can be labile (e.g. in Fig. 16a) because of the increase of M with 9J, or indifferent (not in the figure). Because of the equalizing effect of the structural body which deforms little cotnpared with the soil, the determination of force and displace1nent paths is achieved with few soil elements and simplifying assumptions for distribution of state quantities in the soil. Finite constitutive relations arc of use for coarse estimations hut for the analysis of stability, differential constitutive laws are indispensable. Mutually associated paths (as in Fig. 16b and c) are also suitable for 111ore than two degrees of freedom (e.g. sin1ultaneous settle111ent, tilting and sliding). An extension to cover foundation groups held together by a superstructure is also possible. As in the case of Fig. 15, thermal activation in the grain skeleton can be allowed for with the aid of Eq. 2.36, if the soil is homogenous with respect to the viscosity index. Thus li111it state curves depending on the rate of displace111ent are ohtained, or displace111ents depending on the load duration. The time effect of filtration has to be allowed for, depending on permeability and hydraulic boundary conditions. Special effects can occur for paths with several strong changes of direction (alternating paths, e.g. II in Fig. 16). It suggests, as with soil clements far from lin1it states, that one should assume linearly elastic behaviour and to allow for small changes of state by finite cu111ulation laws. This is insufficient however, if parts of the ground, in spite of sufficient overall stability, repeatedly reach limit states (e.g. for Fig.16 near the edge), leading to softening and rocking of the structural body. Only a refined follow-up of states can show how far simplified approaches suffice for the paths and whether local softening can be avoided hy structural modifications. We now turn to the mutual influence of neighbouring structural hodics through the ground (Fig. I7). Changes of state in the near-field of a structural body act upon its neighbours largely independently of their spatial distribution, so that a simplification of houndary conditions is justified. Therefore a linearly distrihutcd botto111 pressure under a structural body suffices to calculate its quasistatic action upon the neighbour with respect to displacements or forces (a). Changes of the pressure distribution arc lin1ited by the plasticity of the soil. Small changes of grain pressure under the neighbouring structural body can be estimated with linear elasticity, larger changes can be followed up with differential constitutive laws.
1.5 Constitutive laws for soils fron1 a physical viev.rpoint
a
237
b
c
d
Fig. 17. Simplified near-fields of neighbouring structursl bodies: foundation base (a), penetration (b ), excavstion (c), building pit (d)
This holds true also for dynamically loaded neighbouring structural bodies (e.g. driving, Fig. 17b). Resulting forces and their temporal evolution, which are obtained from detailed analysis, may be distributed at the shaft and the foot of the penetrating structural body in a sin1plified inanner. Nearly cyclic state changes under the neighbouring structural body can be estimated with linear elasticity and gradually increasing state changes with the number of cycles by cun1ulation laws. This is not as easy for substantial pore pressure developn1ent due to the tendency for contractancy and bad drainage, but this case should be avoided geotechnically. In the case of a neighbouring excavation the density change caused by the tool is considerable only in a narro\\' near-field (Fig. 17c). The action upon a structure is adequately allo\\ ed for by a stepped reduction of the original soil pressures in the region of excavation. Analogously, the botto1n pressure under a neighbouring fill is increased in steps and simply distributed as redistributions. Spreading shear stress scarcely influences the neighbouring structure. Apart from the bigger mass and stiffness of the neighbouring structure the explanations given in Fig. 14 still hold. The interaction is complicated more by combination and smaller distances (Fig. 17d). To avoid a large amount of numerical calculation it is still covered in the simplest possible n1anner. Si1nplified distributions also suffice for neighbouring changes of pore \\ ater motions and pressures (e.g. due to injection or sinking). Ho\\ ever, hydraulic localization because of predeter1nined or widening channels has to be exempted, 1
1
1
3 Stress-strain relations 3.1
Finite constitutive laws
3.1.1 Ela
L'.T, =2GE+2G--ltrE 1 - 2v
(3.1)
Where the symbol tJ. indicates that- other than with solids- a non-vanishing initial stress Ts is necessary. According to Eq. 2.32 the shear modulus increases with the mean grain
238
Gerd Gudehus
pressure p,1 as follows: G = &dhs(Ps)1-11 h.,
(3.2)
and vanishes for Ps = 0. Granulate hardness hs and the exponent n depend on the grain properties, the factor gJ changes with the density index IJ. The Poisson's ratio v lies between about 0.3 und 0.45. Molenkamp [17] gives a similar formula. Elastic hehaviour is observed for deformations helow typical threshold values, ranging fron1 about 10-5 for clay to io- 8 for ballast. The parameters in Eq. 3.2 can he determined from resonant column tests [26, 28] or estimated according to Section 3.3.3. The propagation speeds of elastic con1pression and shear waves are ohtained fron1 Eq. 3.1 as c,,=/:ITl-v)G/(1-2v)p,
(3.3)
so that v can he determined from cp/Cs· Application lin1its have to be kept in mind when calculating wave propagations and vihrations with Eq. 3.1. Eq. 3.2 shows that even a granulometrically homogeneous soil with horizontal surface is inhomogeneous with respect to G. A homogenization is ohtained with pore water underpressure in the vicinity of the soil surface, which is bigger with smaller grains. When the deforn1ation threshold is exceeded the application of linear elastokinetics is en1pirically justified at best. As the speed of wave propagation exceeds the permeability by several powers of ten, a filtration of pore water is in1possible during vihrations. Volun1e changes are therefore increasingly constrained with increasing degree of saturation S,.. If gas huh hies between the grains are so hig that the pressure difference Pr;-Pw compared to the pore water pressure Pu; is negligible, its change with the volun1e change tr Eis obtained by f!i.pw = -Pw
I+ e
.
e(I - ,,)
trE
(3.4)
Eqs. 3.1 and 3.4 have to be comhined using Eq. 2.lla, resulting into reversihle pore pressure changes for the propagation of longitudinal waves. Eq. 3.4 is not valid for very small gas bubbles due to capillary action and pressure-dependent solubility of gas. For full saturation (S,. = 1), the constraint tr E = 0 replaces Eq. 3.4, so that compression waves hecome theoretically impossible. The linearly elastic hehaviour of simple grain skeletons is generally anisotropic. An orthotropic generalization of Eq. 3.1 can be considered for compact clay soils.. hut the directional parameters can as yet only be obtained from experiments (inherent anisotropy). The anisotropy related with grain stress (called induced) can be allowed for with differential constitutive laws (Sections 3.2.3 and 3.3.3). 3.1.2
Pseudo-elasticity
For alternating loading of the soil with small amplitudes, the changes of state are mostly reversible so that only their cumulation can he registered after many directional changes (Section 3.1.3). Changes of state within one cycle can he sufficiently explained using Eq. 3.1, and in the case of pore water with gas bubbles Eq. 3.4 can also be used. (The energy dissipation is not allowed for in a physically adequate way hy linear viscous damping or with complex shear moduli.)
1.5 Constitutive laws for soils from a physic
239
Pseudo-moduli can be given as ratios /j,Ts//j,E of particular components for particular tnonotonous paths for anelastic changes of state. For uniaxial compression in the oedometcr test one can use the so-called stiffness tnodulus (J.5) with the axial components ~T5 1 and /j,£1. ~ Tsi! ~E1 is called the defortnation modulus Ev for cylindrical shortening with constant lateral pressure. For shearing with constant pressure a pseudo-modulus (3.6)
can he used. These quantities are called tangent moduli for the transition from differences to differentials and secant tnoduli when using the differences between path reversal points. For a given initial grain stress, void ratio and path direction they lie hctwccn minimal values (Esp, Evp or Gsp) and maximal ones (Ese, Ev~ or G 5 ~) for tnaxilnum plastic or maximum elastic behaviour. (The notions static and dynatnic moduli arc misleading, as the differences have nothing to do with different velocities for static or dynatnic cases. It is also not advisable to use the notions first loading, unloading and reloading, as these could be made objective only hy means of special constitutive laws, which are physically not necessary.) For unccmcntcd grain skeletons the difference comes from higgcr or smaller irregularitiy of the grain contact forces (Sections 2.1.1, 3.2.3 and 3.3.3). The ratio of highest and lowest values is nearly constant, (3.7) with me between about 3 and 10. Assuming isotropy, one can transform the maximal values by using Eq. 3.1. This is no longer allowed if anelastic portions dotninate because of higgcr deformations. In particular, dilatancy and contractancy have nothing to do with Poissons's ratio and decreasing differential stiffness, when approaching limit states, has nothing to do with elasticity. The stnallest values can be estitnated using hypoplasticity in their dependence on grain pressure and void ratio (Section 3.3.2). For monotonous processes, upper hounds arc ohtaincd for displacetnents with given forces, and lower hounds for forces with given displacements. Such an application of elasticity theory can only serve as crude approximation, as the physical suppositions (reverseability and linearity) arc not given. Pressure distributions in the soil can thus also he approximated (Poulos [25]). There are non-linear finite constitutive laws for special monotonous paths. For proportional compression, the pseudo-equations of state given in Section 2.1.2 arc of use. If the assu1ned initial states are not given however, this leads to a scarcely tractahle underestitnation of stiffness. Working lines for special loading conditions can be approxitnated by different formulae, which can be used to fit differential constitutive laws, but for lack of ohjectivity not for solving initial boundary value problems. An exception is made hy power laws like (3.8) using a reference pressure p 5 o, a state-dependent factor gp and an exponent~· As long as f~ is constant, which can empirically he the case at least for constant pressure, forcedisplaccmcnt-laws of the same form are concluded fro1n Eq. 3.8.
240
Gerd Gudehus
Pseudo-elastic finite constitutive laY.:s require certain stress and deformation paths that can be estimated on this basis. but more precisely determined only Y.:ith differential constitutive Jai,vs. They arc only suitable for estimating displacements or forces. After judging all the other requirements. one should decide \\.:hcther a differential constitutive la\.\ is necessary. 1
3.1.3
Accumulation and creep-relaxation
Nearly cyclic changes of state arise from alternating loading of simple grain skeletons Y.:ith small deformation amplitudes. If thermal activation is negligible, there is a rateindependent accumulation of state quantities. Finite constitutive la\\.:s arc then generally not suitable. Calculations i,vith differential constitutive laws are so cumbersome for large numbers of cycles however. that one has to work Y.:ith simplified approximation formulae. One can refer to a series of cycles if stress and deformation amplitudes remain nearly constant inside it. An elastic constitutive law holds approximately true i,vithin each cycle Y.:here the shear modulus depends on void ratio and pressure (Section 3.1.1 ). For almost constant pressure the void ratio tends to a pressure-dependent loY.:er bound e11 (Fig. 10). The reduction of void ratio Y.:ith the number of cycles N can then be approximated by (3.9)
i,vhere Ee is the deformation amplitude and that
K
a material constant. It therefore folloi,vs
(3.10)
i,vith the initial void ratio ea· For grain pressure cycles the deformations increase Y.:ith each cycle. It is often assumed that a deformation component increases by dE = K,dN/N
(3.11)
so that the increase is reduced by increasing N. and E = £0
+ K, lnN
(N::: 1)
(3.12)
is obtained.Ke depends on grain pressure components. Similar approaches can be derived for the mean pressure reduction Y.:ith deformation cycles and constant volume (Fig. 10). These approaches and the parameters therein, are often fitted to the results of special cyclic tests, but this is debatable because of the arbitrariness of the initial states i,vithout an allo\\i·ance for asymptotic properties of the grain skeleton. Better accumulation formulae and more relevant parameters can be obtained by integration \\.:ith differential constitutive laws (Sections 3.2.3 and 3.3,3). They serve for numerical simplification in order to avoid a non-linear integration over numerous cycles. At the end of a series of cycles - arbitrarily delimited by tolerances - cumulative changes of state quantities have to be corrected by means of a constitutive law for a monotonous correction path, so that equilibrium and compatibility relations are satisfied. Continuity of the differential stiffness matrix is a necessary assumption for such an approximation method.
1.5 Constitutive laws for soils from a physical vicv.•point
241
Tlle stationary viscous ftoH..' due to thermal activation has already be dealt with in Section 2.2.2. In Eq. 2.17 the mean grain pressure (instead of Eq. 2.36) is given by (3.13)
Wherein Psr is the Ps for stationary flow with D = Dr and e = er. In Eqs. 2.34 and 2.35 the viscosity index is G Iv =kT 21 c d
(3.14)
where only the proportionality with T and G is of use, whereas the contact quantities c and d can only be crudely estin1ated. According to Eqs. 3.13 and 3.14, Ps increases for constant void ratio due to an increase of D or T, whereas for constant p,1 the void ratio increases with an increase of D or T. Eq. 3.13 holds for proportional compression with a lower e1, and e goes down gradually. For uniaxial shortening Dis related to the rate of void ratio e by D = e/(l + e1). With constant grain pressure (Ps = 0), Eq. 3.13 then gives
e
e
(3.15)
- " ' I!! e·2 C ,.
With the initial condition e = e0 , e = e = eo - C,1,,(1
+ e,) ln(f /to)
e0 and fort
=to one can conclude
(t :Oto)
(3.16)
When using this empirically known relation for secondary consolidation [27], one takes the duration of primary consolidation as to. However Eq. 3.16 is not fully correct: The factor 1 +I, ln(D/ D,) has been replaced by 1 in Eq. 3.15, ea= C,l,,/to docsnot generally hold fort = lo and Eq. 2.17 cannot hold for very small and very big values of p,. Eq. 2.36 has to be replaced by a linear viscosity relation for extremely small D, so e tends to a limit value for Ps = 0, but only after extremely long periods. The reduction of void ratio can also be described by Eq. 3.16 for other creep cases with constant skeleton pressure. For estimating the increase in deformation con1ponents, one can consider their ratio as constant for a proportional compression. Such simple approaches are no longer justified for sn1all deformations after a change of path direction, as the grain skeleton's state is no longer described then only by void ratio, grain pressure tensor and stretching rate. T11ere are also no physically justified finite constitutive laws for the relaxation of a fixed grain skeleton.
3.2 3.2.l
Elastoplasticity Simple How condition and associated flow rule
For plastic deformations of a perfect solid one has to consider only the dcviators l
•
1
(3.17) T' =T- -ltrT, E=E--ltrT 3 3 as the volume changes are elastic if extremely high pressures or tension ~tresses are excluded [20]. 1be deviatoric deformation energy per unit volun1e in the case of isotropy,
U'=trT* 2 /2G
(3.18)
242
Gerd Gudehus
is the potential of the elastic deviatoric deformation.
E; =au* /aT*
(3.19)
in accordance with Eq. 3.1. If U* satisfies the flow condition 1
F:=U'--c2 /G=0 2
(3.20)
with cohesion c, a plastic deformation rate hy the as.vociated flow rule
Ef!
will arise, the direction of which is given (3.21)
with an as yet undetermined factor A > 0. As a result of Eqs. 3.18 and 3.19, tional to T* and E;.
Ep is propor-
Dy dislocations, which arise already during crystallisation and continue during plastic deformations, spatially fluctuating eigenstresses arise and the isotropy is lost. One can follow up the composite behaviour with an assumed initial array and stress-Held for perfect crystallites. However, frequently one works instead with a fictitious cigcnstrcss a and uses Eqs. 3.18 to 3.21 with the stress difference [18]
(3.22) instead of T*. The change of a with plastic deformation is described by a hardening law, (3.23) where A denotes a constant and T* = T* / I T$ I is the deviatoric stress direction. lJsing the consistency condition required for continued plastic deforn1ation, . ( aF . F = tr UT* T*
+
aF ·)
nu. a
= 0
(3.24)
one obtains ,< = G tr(T
*• T')/ A tr(T'T') tr(T;i")
(3.25)
With the decomposition
t
=
t, + tp
(3.26)
of the deformation rate into an elastic and a plastic rx>rtion, one then obtains the differential constitutive law
