Podolny and Muller - Construction and Design of Pre Stressed Concrete Segmental Bridges

Construction and Design of Prestressed Concrete Segmental Bridges

Walter Podolny, Jr., Ph.D., P.E. Bridge Division Office of Engineering Federal Highway Administration U.S. Department of Transportation

Jean M. Muller Chairman of the Board Figg and Muller Engineers, Inc.

1982 A Wiley-Interscience Publication

John Wiley &? Sons New

York

Chichester

Brisbane

Toronto

Singapore

Series Preface J

The Wiley Series of Practical Construction Guides provides the Ivorking constructor \vith up-to-date information that can help to increase the job profit margin. These guidebooks, ivhich are scaled mainly for practice, but include the necessary theory and design, should aid a construction contractor in approaching \+vork problems with more knolvledgeable confidence. The guides should be useful also to engineers, architects, planners, specification tvriters, project managers, superintendents. materials and equipment manufacturers and. the source of all these callings, instructors and their students. Construction in the United States alone will reach $250 billion a year in the early 1980s. In all nations. the business of building will continue to grow at a phenomenal rate, because the population proliferation demands new living, lvorking, and recreational facilities. This construction will have to be more substantial, thus demanding a more

professional performance from the contractor. Before science and technology had seriously affected the ideas, job plans, financing, and erection of structures. most contractors developed their knolv-holy by field trial-and-error. Wheels, small and large. jvere constantly being reinvented in all sectors, because there was no interchange of knolvledge. The current complexity of construction, even in more rural areas, has revealed a clear need for more proficient, professional methods and tools in both practice and learning. Because construction is highly competitive, some practical technology is necessarily proprietary. But most practical day-to-day problems are common to the Fvhole construction industry. These are the subjects for the Wiley Practical Construction Guides. M. D. MORRIS , P.E.

Preface J

Prestressed concrete segmental bridge construction has evolved, in the natural course of events, from the combining of the concepts of prestressing, box girder design, and the cantilever method of bridge construction. It arose from a need to overcome construction difficulties in spanning deep valleys and river crossings without the use of conventional falsework, which in some instances may be impractical, economically prohibitive, or detrimental to environment and ecology. Contemporary prestressed, box girder, segmental bridges began in Western Europe in the 1950s. Ulrich Finsterwalder in 1950, for a crossing of the Lahn River in Balduinstein, Germany, was the first to apply cast-in-place segmental construction to a bridge. In 1962 in France the first application of precast, segmental, box girder construction was made by Jean Muller to the ChoisyLe-Roi Bridge crossing the Seine River. Since then the concept of segmental bridge construction has been improved and refined and has spread from Europe throughout most of the world. The first application of segmental bridge construction in North America was a cast-in-place segmental bridge on the Laurentian Autoroute near Ste. Adele, Quebec, in 1964. This was followed in 1967 by a precast segmental bridge crossing the Lievre River near Notre Dame du Laus, Quebec. In 1973 the first U.S. precast segmental bridge was opened to traffic in Corpus Christi, Texas, followed a year later by the cast-in-place segmental Pine Valley Bridge near San Diego, California. As of this date (1981) in the United States more than eighty segmental bridges are completed, in construction, in design, or under consideration. Prestressed concrete segmental bridges may be identified as precast or cast in place and categorized by method of construction as balanced cantilever, span-by-span, progressive placement, or incremental launching. This type of bridge has

extended the practical and competitive economic span range of concrete bridges. It is adaptable to almost any conceivable site condition. The objective of this book is to summarize in one volume the current state of the art of design and construction methods for all types of segmental bridges as a ready reference source for engineering faculties, practicing engineers, contractors, and local, state, and federal bridge engineers. Chapter 1 is a quick review of the historical evolution to the current state of the art. It offers the student an appreciation of the way in which segmental construction of bridges developed, the factors that influenced its development, and the various techniques used in constructing segmental bridges. Chapters 2 and 3 present case studies of the predominant methodology of constructing segmental bridges by balanced cantilever in both cast-in-place and precast concrete. Conception and design of the superstructure and piers, respectively, are discussed in Chapters 4 and 5. The other three basic methods of constructing segmental bridgesprogressive placement, span-by-span, and incremental launching-are presented in Chapters 6 and 7. Chapters 2 through 7 deal essentially with girder type bridges. However, segmental construction may also be applied to bridges of other types. Chapter 8 discusses application of the segmental concept to arch, rigid frame, and truss bridges. Chapter 9 deals with the cable-stayed type of bridge and Chapter 10 with railroad bridges. The practical aspects of fabrication, handling, and erection of segments are discnssed in Chapter 11. In selected a bridge type for a particular site, one of the more important parameters is economics. Economics, competitive bidding, and contractual aspects of segmental construction are discussed in Chapter 12. Most of the material presented in this book is not vii

Preface

original: Although acknowledgment of all the many.source$&. not possible, full credit is given wherever the specific so;rce can be identified. Every effort has been. made to eliminate errors; the authors will appreciate notification from the reader ‘of any that remain. The authors are indebted to numerous publications, organizations, and individuals for their assistance and permission to reproduce photo-

graphs, tables, and other data. Wherever possible, credit is given in the text. WALTER PODOLNY, JEAN M. MUILEK Burke, Virginia Par%, Francr Jarmar? 1982

JK.

Contents 1

Prestressed Concrete Bridges and Segmental Construction 1.1 1.2 1.3 1.4 1.5 1.6 1.7 1.8 1.9 1.10

1.11

2

Introduction, 1 Development of Cantilever Construction, 2 Evolution of Prestressed Concrete, 4 Evolution of Prestressed Concrete Bridges, 5 Long-Span Bridges with Conventional Precast Girders, 8 Segmental Construction, 10 Various Types of Structures, 12 Cast-in-Place and Precast Segmental Construction, 17 Various Methods of Construction, 18 Applications of Segmental Construction in the United States, 26 Applicability and Advantages of Segmental Construction, 28 References, 30

Cast-In-Place Balanced Cantilever Girder Bridges 2.1 2.2 2.3 2.4 2.5 2.6 2.7

Introduction, 3 1 Bendorf Bridge, Germany, 35 Saint Adele Bridge, Canada, 37 Bouguen Bridge in Brest and Lacroix Falgarde Bridge, France, 38 Saint Jean Bridge over the Garonne River at Bordeaux, France, 4 1 Siegtal and Kochertal Bridges, Germany, 43 Pine Valley Creek Bridge, U.S.A., 46

2.8 2.9 2.10 2.11 2.12

1

2.13 2.14 2.15 2.16

3

Precast Balanced Cantilever Girder Bridges 3.1 3.2 3.3 3.4

31

Gennevilliers Bridge, France, 52 Grand’Mere Bridge, Canada, 55 Arnhem Bridge, Holland, 58 Napa River Bridge, U.S.A., 59 Koror-Babelthuap, U.S. Pacific Trust Territory, 61 Vejle Fjord Bridge, Denmark, 63 Houston Ship Channel Bridge, U.S.A., 68 Other Notable Structures, 71 Conclusion, 8 1 References, 8 1

3.5 3.6 3.7 3.8 3.9 3.10 3.11 3.12 3.13 3.14 3.15 3.16

82

Introduction, 82 Choisy Le Roi Bridge and Other Structures in Greater Paris, France, 83 Pierre Benite Bridges near Lyons, France, 89 Other Precast Segmental Bridges in Paris, 91 Oleron Viaduct, France, 96 Chillon Viaduct, Switzerland, 99 Hartel Bridge, Holland, 103 Rio-Niteroi Bridge, Brazil, 106 Bear River Bridge, Canada, 108 JFK Memorial Causeway, U.S.A., 109 Saint Andre de Cubzac Bridges, France, 113 Saint Cloud Bridge, France, 114 Sallingsund Bridge, Denmark, 122 B-3 South Viaducts, France, 124 Alpine Motorway Structures, France, 129 Bridge over the Eastern Scheldt, Holland, 134 ix

X

3.17 Captain Cook Bridge, Australia, 136 3.18 Other Notable Structures, 1 3 9 References, 147 4

Design of Segmental Bridges 4.1 4.2 4.3 4.4 4.5 4.6 4.7 4.8 4.9 4.10 4.11 4.12 4.13 4.14 4.15 4.16 4.17 4.18 4.19

5

5.4

5.6 5.7

Introduction, 225 Loads Applied to the Piers, 2 3 0 Suggestions on Aesthetics of Piers and Abutments, 232 Moment-Resisting Piers and Their Foundations, 234

5.8 5.9

148

Introduction, 148 Live Load Requirements, 149 Span Arrangement and Related Principle of Construction, 149 Deck Expansion, Hinges, and Continuity, 15 1 Type, Shape and Dimensions of the Superstructure, 159 Transverse Distribution of Loads Between Box Girders in Multibox Girders, 164 Effect of Temperature Gradients in Bridge Superstructures, 170 Design of Longitudinal Members for Flexure and Tendon Profiles, 173 Ultimate Bending Capacity of Longitudinal Members, 190 Shear and Design of Cross Section, 193 Joints Between Match-Cast Segments, 199 Design of Superstructure Cross Section, 202 Special Problems in Superstructure Design, 203 Deflections of Cantilever Bridges and Camber Design, 205 Fatigue in Segmental Bridges, 2 10 Provisions for Future Prestressing, 2 12 Design Example, 2 12 Quantities of Materials, 219 Potential Problem Areas, 220 References, 224

Foundations, Piers, and Abutments 5.1 5.2 5.3

5.5

Piers with Double Elastomeric Bearings, 24 1 Piers with Twin Flexible Legs, 253 Flexible Piers and Their Stability During Construction, 263 Abutments, 27 1 Effect of Differential Settlements on Continuous Decks, 276 References, 280

6 Progressive and Span-by-Span Construction of Segmental Bridges 6.1 6.2 6.3 6.4 6.5 6.6

Introduction, 281 Progressive Cast-in-Place Bridges, 283 Progressive Precast Bridges, 289 Span-by-Span Cast-in-Place Bridges, 293 Span-by-Span Precast Bridges, 308 Design Aspects of Segmental Progressive Construction, 3 14 References, 3 19

7 Incrementally Launched Bridges 7.1 7.2 7.3 7.4 7.5 7.6 7.7 7.8 7.9 7.10

8 225

8.3 8.4 8.5 8.6

32

Introduction, 32 1 Rio Caroni, Venezuela, 323 Val Restel Viaduct, Italy, 327 Ravensbosch Valley Bridge, Holland, 329 Olifant’s River Bridge, South Africa, 33 1 Various Bridges in France, 333 Wabash River Bridge, U.S.A., 335 Other Notable Bridges, 338 Design of Incrementally Launched Bridges, 343 Demolition of a Structure by Incremental Launching, 352 References, 352

Concrete Segmental Arches, Rigid Frames, and Truss Bridges 8.1 8.2

2,

Introduction, 354 Segmental Precast Bridges over the Marne River, France, 357 Caracas Viaducts, Venezuela, 363 Gladesville Bridge, Australia, 37 1 Arches Built in Cantilever, 374 Rigid Frame Bridges, 382

35

xi

Contents 8.7

Truss Bridges, 392 References, 399

11

11.1 11.2

9 Concrete Segmental Cable-Stayed Bridges 400 9.1 9.2 9.3 9.4 9.5 9.6 9.7 9.8 9.9 9.10

Introduction, 400 Lake Maracaibo Bridge, Venezuela, 405 Wadi Kuf Bridge, Libya, 407 Chaco/Corrientes Bridge, Argentina, 408 Mainbrticke, Germany, 410 Tie1 Bridge, Netherlands, 412 Pasco-Kennewick Bridge, U.S.A., 418 Brotonne Bridge, France, 419 Danube Canal Bridge, Austria, 427 Notable Examples of Concepts, 430 References, 439

10 Segmental Railway Bridges 10.1 10.2 10.3 10.4 10.5 10.6 10.7 10.8

10.9 10.10 10.11

Introduction to Particular Aspects of Railway Bridges and Field of Application, 44 1 La Voulte Bridge over the Rhone River, France, 442 Morand Bridge in Lyons, France, 442 Cergy Pontoise Bridge near Paris, France, 444 Marne La Vallee and Torcy Bridges for the New Express Line near Paris, France, 444 Clichy Bridge near Paris, France, 449 Olifant’s Bridge, South Africa, 452 Incrementally Launched Railway Bridges for the High-Speed Line, Paris to Lyons, France, 453 Segmental Railway Bridges in Japan, 457 Special Design Aspects of Segmental Railway Bridges, 458 Proposed Concepts for Future Segmental Railway Bridges, 464

Technology and Construction of Segmental Bridges

11.3 1.1.4

11.5 11.6 11.7 11.8

441

12

Scope and Introduction, 465 Concrete and Formwork for Segmental Construction, 466 Post-tensioning Materials and Operations, 470 Segment Fabrication for Cast-In-Place Cantilever Construction, 475 Characteristics of Precast Segments and Match-Cast Epoxy Joints, 485 Manufacture of Precast Segments, 493 Handling and Temporary Assembly of Precast Segments, 507 Placing Precast Segments, 509 References, 5 17

Economics and Contractual Aspects of Segmental Construction 12.1 12.2 12.3

13.4 13.5 13.6

Index Index Index Index

518

Bidding Procedures, 5 18 Examples of Some Interesting Biddings and Costs, 523 Increase in Efficiency in Concrete Bridges, 528 References, 535

13 Future Trends and Develofnnents 13.1 13.2 13.3

465

536

Introduction, 536 Materials, 536 Segmental Application to Bridge Decks, 542 Segmental Bridge Piers and Substructures, 543 Application to Existing or New Eridge Types, 544 Summary, 548 References, 549 of Bridges of Personal Names of Firms and Organizations of Subjects

551 555 557 559

Construction and Design of Prestressed Concrete Segmental Bridges

1 Prestressed Concrete Bridges and Segmental Construction

1.1 INTRODUCI’ION 1.2 DEVELOPMENT OF CANTILEVER CONSTRUCITON 1.3 EVOLUTION OF PRESTRESSED CONCRETE 1.4 EVOLUTION OF PRESTRESSED CONCRETE BRIDGES 1.5 LONGSPAN BRIDGES WITH CONVENTIONAL PRECAST GIRDERS 1.6 SEGMENTAL CONSTRUCTION 1.7 VARIOUS TYPES OF STRUCl-URFS 1.7.1 Girder Bridges 1.7.2 Trusses 1.7.3 Frames with Slant Legs 1.7.4 Concrete Arch Bridges 1.7.5 Concrete CabkStayed Bridges 1.8 CAST-IN-PLACE AND PRECAST SEGMENTAL CONSTRUCTION

1 . l Zntroduction, The conception, development, and worldwide acceptance of,segmental construction in the field of prestressed concrete bridges represents one of the most interesting and important achievements in civil engineering during the past thirty years. Recognized today in all countries and particularly in the United States as a safe, practical, and economic construction method, the segmental concept probably owes its rapid growth and acceptance to its founding, from the beginning, on sound construction principles such as cantilever construction. Using this method, a bridge structure is made up of concrete elements usually called segments (either precast or cast in place in their final position in the structure) assembled by post-tensioning. If the bridge is cast in place, Figure 1.1, travelers are used to allow the various segments to be constructed in successive increments and progressively

1.8.1 1.8.2 1.8.3 1.9

1.10 1.11

Characteristics of Cast-in-Place Segments Characteristics of Precast Segfnents Choice between Cast-in-Place and Precast Construction VARIOUS METHODS OF CONSTRUCTION 1.9.1 Cast-in-Place Balanced Cantilever 1.9.2 Precast Balanced Cantilever 1.9.3 Span-by-Span Construction 1.9.4 Progressive Placement Construction 1.9.5 Incremental Launching or Push-Out Construction APPLICATIONS OF SEGMENTAL CONSTRUCTION IN THE UNITED STATES APPLICABILITY AND ADVANTAGES OF SEGMENTAL CONSTRUCI’ION REFERENCES

prestressed together. If the bridge is precast, segments are manufactured in a special casting yard or factory, transported to their final position, and placed in the structure by various types of launch-

FIGURE 1.1 Cast-in place form traveler. 1

Prestressed

Concrete Stidges ad Segmental Constrt4ction

FIGURE 1.2. Oleron Viaduct, segmental construction in progress. One typical precast segment placed in the Oleron Viaduct.

ing equipment, Figure 1.2, while prestressing achieves the assembly and provides the structural strength. Most early segmental bridges were built as cantilevers, where construction proceeds in a symmetrical fashion from the bridge piers in successive increments to complete each span and finally the entire superstructure, Figure 1.3. Later, other construction methods appeared in conjunction with

______ Llzcr---/.#-------. ,% --------------/-------l-r -77 -------------.------3-r

FIGURE 1.3. Cantilever construction applied to prestressed concrete bridges.

the segmental concept to further its field of application. 1.2

Development of Cantilever Construction

The idea of cantilever construction is ancient in the Orient. Shogun’s Bridge located in the city of Nikko, Japan, is the earliest recorded cantilever bridge and dates back to the fourth century. The Wandipore Bridge, Figure 1.4, was built in the

seventeenth century in Bhutan, between India and Tibet. It is constructed from great timbers that are corbeled out toward each other from massive abutments and the narrowed interval finally capped with a light beam.’

FIGURE 1.4. Wandipore Bridge.

Develofwnent

of Cantilever Construction

3

That half an arc should stand upon the ground Without support while building, or a rest; This caus’d the theorist’s rage and sceptic’s jest. Prefabrication techniques were successfully combined with cantilever construction in many bridges near the end of the nineteenth century, as exemplified by such notable structures as the Firth of Forth Bridge, Figure 1.6, and later the Quebec Bridge, Figure 1.7, over the Saint Lawrence River. These structures bear witness to the engineering genius of an earlier’ generation. Built more recently, the Greater New Orleans Bridge over the Mississippi River, Figure 1.8, represents modern contemporary long-span steel cantilever construction. Because the properties and behavior of prestressed concrete are related more closely to those of structural steel than those of conventional reinforced concrete, the application of this material to cantilever construction was a logical step in the continuing development of bridge engineering.

FIGURE 1.7. Quebec Bridge.

FIGURE 1.8. Greater New Orleans Bridge.

Prestressed Concrete Bridges and Segmental Construction

4

This application has evolved over many years by the successive development of many concepts and innovations. In order to see how the present state of the art has been reached, let us briefly trace the development of prestressed concrete and in particular its application to bridge construction.

1.3

Evolution

of

Prestressed

Concrete

The invention of reinforced concrete stirred the imagination of engineers in many countries. They envisioned that a tremendous advantage could be achieved, if the steel could be tensioned to put the structure in a permanent state of compression greater than any tensile stresses generated by the applied loads. The present state of the art of prestressed concrete has evolved from the effort and experience of many engineers and scientists over the past ninety years. However, the concept of prestressing is centuries old. Swiss investigators have shown that as early as 2700 B.C. the ancient Egyptians prestressed their seagoing vessels longitudinally. This has been determined from pictorial representations found in Fifth Dynasty tombs. The basic principle of prestressing was used in the craft of cooperage when the cooper wound ropes or metal bands around wooden staves to form barrels.3 When the bands were tightened, they were under tensile prestress, which created compression between the staves and enabled them to resist hoop tension produced by internal liquid pressure. In other words, the bands and staves were both prestressed before they were subjected to any service loads. The wooden cartwheel with its shrunk-on iron rim is another example of prestressed construction. The first attempt to introduce internal stresses in reinforced concrete members by tensioning the steel reinforcement was made about 1886 when P. H. Jackson, an engineer in San Francisco, obtained a United States patent for tightening steel rods in concrete members serving as floor slabs. In l&S, C. E. W. DGhring of Berlin secured a patent for the manufacture of slabs, battens, and small beams for structural engineering purposes by embedding tensioned wire in concrete in order to reduce cracking. This was the first attempt to provide precast concrete units with a tensioned reinforcement. Several structures were constructed using these concepts; however, only mild steel reinforcement was available at the time. These structures at first behaved according to predictions, but because so little prestress force could be induced in the mild

steel, they lost their properties because of the creep and shrinkage of the concrete. In order to recover some of the losses, the possibility of retightening the reinforcing rods after some shrinkage and creep of the concrete had taken place was suggested in 1908 by C. R. Steiner of the United States. Steiner proposed that the bond of embedded steel bars be destroyed by lightly tensioning the bars while the concrete was still young and then tensioning them to a higher stress when the concrete had hardened. Steiner was also the first to suggest the use of curved tendons. In 1925, R. E. Dill of Nebraska took a further step toward freeing concrete beams of any tensile stresses by tensioning high-tensile steel wires after the concrete had hardened. Bonding was to be prevented by suitably coating the wires. He explicitly mentioned the advantage of using steel with a high elastic limit and high strength as compared to ordinary reinforcing bars. In 1928, E. Freyssinet of France, who is credited with the modern development of prestressed concrete, started using high-strength steel wires for prestressing. Although Freyssinet also tried the method of pretensioning, where the steel was bonded to the concrete without end anchorages, the first practical application of this method was made by E. Hoyer about 1938. Wide application of the prestressing technique was not possible until reliable and economical methods of tensioning and end anchorage were devised. From approximately 1939 on, E. Freyssinet, Magnel, and others developed different methods and procedures. Prestress began to gain some importance about 1945, while alternative prestressing methods were being devised by engineers in various countries. During the past thirty years, prestressed concrete in the United States has grown from a brand-new idea into an accepted method of concrete construction. This growth, a result of a new application of existing materials and theories, is in itself phenomenal. In Europe the shortage of materials and the enforced economies in construction gave prestressed concrete a substantial start. Development in the United States, however, was slower to get underway. Designers and contractors hesitated mainly because of their lack of experience and a reluctance to abandon more familiar methods of construction. Contractors, therefore, bid the first prestressed concrete work conservatively. Moreover, the equipment available for prestressing and related techniques was essentially new and makeshift. However, experience was gained rapidly, the quality of the work improved,

Evolution

of

Prestressed

Concrete

Bridges

5

FIGURE 1.9. Freyssinet’s Esblv Bridge on the Marne River.

and prestressed concrete became more and more competitive with other materials.

1.4

Evolution of Prestressed Concrete Bridges

Although France took the lead in the development of prestressed concrete, many European countries such as Belgium, England, Germany, Switzerland, and Holland quickly showed interest. As early as 1948, Freyssinet used prestressed concrete for the construction of five bridges over the Marne River near Paris, with 240 ft (74 m) spans of an exceptionally light appearance, Figure 1.9. A survey made in Germany showed that between 1949 and 1953, out of 500 bridges built, 350 were prestressed.

FIGURE 1.10 W a l n u t L a n e B r i d g e , Phil,~dcll~hia (courtesy of the Portland Cement Association).

Prestressing in the United States followed a different course. Instead of linear prestressing, circular prestressing as applied to storage tanks took the lead. Linear prestressing as applied to beams did not start until 1949. The first structure of this type was a bridge in Madison County, Tennessee, followed in 1950 by the well-known 160 ft (48.80 m) span Walnut Lane Bridge in Philadelphia, Figure 1.10. By the middle of 1951 it was estimated that 175 bridges and 50 buildings had been constructed in Europe and no more than 10 structures in the United States. In 1952 the Portland Cement Association conducted a survey in this country showing 100 or more structures completed or

FIGURE 1.11. AASHTO-PC1 I-girder cross sections.

6

Prestressed

Concrete

Bridges

under construction. In 1953 it was estimated that there were 75 bridges in Pennsylvania alone. After the Walnut Lane Bridge, which was cast in place and post-tensioned, precast pretensioned bridge girders evolved, taking advantage of the inherent economies and quality control achievable with shop-fabricated members. With few exceptions, during the 1950s and early 196Os, most multispan precast prestressed bridges built in the United States were designed as a series of simple spans. T h e y w e r e d e s i g n e d w i t h s t a n d a r d AASHTO-PCI* girders of various cross sections, Figure 1.11, for spans of approximately 100 ft (30.5 m), but more commonly for spans of 40 to 80 ft (12 to 24 m). The advantages of a continuous cast-in-place structure were abandoned in favor of t h e s i m p l e r c o n s t r u c t i o n o f f e r e d b y plantproduced standardized units. At this time, precast pretensioned members found an outstanding application in the Lake Pontchartrain crossing north of New Orleans, Louisiana. The crossing consisted of more than 2200 identical 56 ft (17 m) spans, Figures 1.12 through 1.14. Each span was made of a single 200 ton monolith with pretensioned longitudinal gird*American Association of State Highway and Transportation Officials (previously known as AASHO, American Association of State Highway Officials) and Prestressed Concrete Institute.

and

Segmental

Construction

FIGURE 1.12. Lake Pontchartrain Bridge, U.S.A.

ers and a reinforced concrete deck cast integrally, resting in turn on a precast cap and two prestressed spun piles. The speed of erection was incredible, often more than eight complete spans placed in a single day. In the middle 1960s a growing concern was shown about the safety of highways. The AASHTO Traffic Safety Committee called in a 1967 report 4 for the “ . . . adoption and use of twospan bridges for overpasses crossing divided highways . . . to eliminate the bridge piers normally placed adjacent to the shoulders,” Figure 1.15. Interstate highways today require overpasses with two, three, and four spans of up to 180 ft (54.9 m) or longer. In the case of river or stream crossings,

FIGURE 1.13. Lake Pontchartrain Bridge, U.S.A.

33'4

18'4'

I

I

I (b)

FIGURE 1.14. verse section.

Lake Pontchartrain Bridge, U.S.A. (a) Longitudinal section. (b) Trans7

-

bestressed Concrete Bridges and Segmental Construction

8

STANMRD 4-SPAN INTERSTATE CROSSING I

tg

177’ 250’

FIGURE 1.15. Standard four-span interstate crossing (courtesv of the Portland Cement Association). longer spans in the range of 300 ft (91.5 m) or longer may be required, and there is a very distinct trend toward longer-span bridges. It soon became apparent that the conventional precast pretensioned AASHTO-PC1 girders were limited by their transportable length and weight. Transportation over the highways limits the precast girder to a length of 100 to 120 ft (30.5 to 36.6), depending upon local regulations.

I .5

Long-Span Bridges with Precast Girders

Conventional

As a result of longer span requirements a study was conducted by the Prestressed Concrete Institute (PCI) in cooperation with the Portland Cement Association (PCA).S This study proposed that simple spans up to 140 ft (42.7 m) and continuous spans up to 160 ft (48.8 m) be constructed of standard precast girders up to 80 ft (24 m) in length joined by splicing. To obtain longer spans the use of inclined or haunched piers was proposed. The following discussion and illustrations are based on the grade-separation studies conducted by PC1 and PCA. Actual structures will be illus-

trated, where possible, to emphasize the particular design concepts. The design study illustrated in Figure 1.16 uses cast-in-place or precast end-span sections and a two-span unit with AASHTO I girders.6 Narrow median piers are maintained in this design, but the abutments are extended into the spans by as much as 40 ft (12 m) using a precast or cast-in-place frame in lieu of a closed or gravity abutment. When site conditions warrant, an attractive type of bridge can be built with extended abutments. A similar span-reducing concept is developed in Figure 1.17, using either reinforced or prestressed concrete for cantilever abutments. An aesthetic abutment design in reinforced concrete was developed for a grade-separation structure on the Trans-Canada Highway near Drummondville in the Province of Quebec, Figure 1.18. This provided a 324 ft (9.9 m) span reduction that led to the use of type IV Standard AASHTO I girders to span 974 ft (29.7 m) to a simple, narrow median pier. A cast-in-place reinforced concrete frame with outward-sloping legs provides a stable, center supporting structure that reduces span length by 29 ft (8.8 m), Figure 1.19. This enables either standard box sections or I sections 84 ft (25.6 m) long to be used in the two main spans. This layout was used for the Hobbema Bridge in Alberta, B.C., Canada, shown in Figure 1.20. This bridge was built with precast channel girder sections, but could be built with AASHTO I girders or box sections. The median frame with inclined legs was cast in place. The schematic and photograph in Figures 1.21 and 1.22 show the Ardrossan Overpass in Alberta. It is similar to the Hobbema Bridge except that the spans are longer and, with the exception of a cast-in-place footing, the median frame is made up of precast units post-tensioned together, Figure 1.21. The finished bridge, Figure 1.23, has a

Carl-in-place Froma

SECTION

A-A

FIGURE 1.16. Extended abutments (courtesy of the Prestressed Concrete Institute, from ref. 6).

Long-Span

APPROX.

Bridges with

36’ Its’-0”

Conventional

ELEVATION r;

Precast

Girders

9

A P P R O X . *I’-

IS’-0”

T Y P E lx A A S H O OlROtR OIROER

SECTION

FIGURE 1.1’7. Cantilevered abutments (courtesy of the Prestressed Concrete Institute, from ref. 6). \\,\_ \ \\\ \\\\ \ ,, \ \\\ \\ \ \

\

FIGURE 1.18. Drummondville Bridge (courtesy of the Portland Cement Association).

pleasing appearance. The standard units were channel-shaped stringers 64 in. wide and 41 in. deep (1.6 m by 1.04 m). The use of precast units allowed erection of the entire superstructure, ineluding the median frame, in only three weeks. The bridge was opened to traffic just eleven weeks after construction began in the early summer of 1966. By use of temporary bents, Figure 1.24, standard units 60 ft (18.3 m) long can be placed over the median pier and connected to main span units with cast-in-place reinforced concrete splices located near the point of dead-load contraflexure.

ELEVATION

SECTIONS A -A

FIGURE 1.19. Median frame cast in place (courtesy of the Prestressed Concrete Institute, from ref. 6).

10

Prestressed

Concrete

Bridges

and

Segmental

Construction

from the side pier over the main pier to the hingesupport for the suspended span. The type of construction that uses long, standard, precast, prestressed units never quite achieved the recognition it deserved. As spans increased, designers turned toward post-tensioned cast-in-place box girder construction. The California Division of Highways, for example, has been quite successful with cast-in-place, multicell, posttensioned box girder construction for multispan structures with spans of 300 ft (91.5 m) and even longer. However, this type of construction has its own limitations. The extensive formwork u s e d during casting often has undesirable effects on the environment or the ecology.

FIGURE 1.20. Hobbema Bridge, completed structure (courtesy of the Portland Cement Association).

1.6 Segmental Construction This design is slightly more expensive than previ-

ous ones but it provides the most open type twospan structure. The structural arrangement of the Sebastian Inlet Bridge in Florida consists of a three-span unit over the main channel, Figure 1.25. The end span of this three-span unit is 100 ft (30.5 m) long and cantilevers 30 ft (9 m) beyond the piers to support a 120 ft (36.6 m) precast prestressed drop-in span, Figure 1.26. The end-span section was built in two segments with a cast-in-place splice with the help of a falsework bent. The Napa River Bridge at Vallejo, California (not to be confused with the Napa River Bridge described in Section 2.1 l), used a precast concrete cantilever-suspended span concept similar to the Sebastian Inlet Bridge, at about the same time. The only difference was that the cantilever girder was a single girder extending

ELE

V A

Segmental construction has been defined’ as a method of construction in which primary loadsupporting members are composed of individual members called segments post-tensioned together. The concepts developed in the PCI-PCA studies and described in the preceding section come under this definition, and we might call them “longitudinal” segmental construction because the individual elements are long with respect to their width. In Europe, meanwhile, segmental construction proceeded in a slightly different manner in conjunction with box girder design. Segments were cast in place in relatively short lengths but in fullrpadway width and depth. Today segmental construction is usually understood to be the type developed in Europe. However, as will be shown later, the segments need not be of full-roadway

T

I O N

81p-40 AASHO-PCI BOX SECTION

3’-6”

6’-6*

b’-6”

&X-IONS A - A

FIGURE

1.21.

Median frame precast (courtesy of the Prestressed Cot xrete Institute, from ref. 6).

Segmental Construction

FIGURE 1.22. Ardrossan Overpass precast median frame (courtesy of the Portland Cement Association).

width and can become rather long in the longitudinal direction of the bridge, depending on the construction system utilized. Eugene Freyssinet, in 1945 to 1948, was the first to use precast segmental construction for prestressed concrete bridges. A bridge at Luzancy over the Marne River about 30 miles east of Paris, Figure 1.27, was followed by a group of five precast bridges over that river. Shortly thereafter, Ulrich Finsterwalder applied cast-in-place segmental prestressed construction in a balanced cantilever fashion to a bridge crossing the Lahn River at Balduinstein, Germany. This system of cantilever segmental construction rapidly gained wide acceptance in Germany, after construction of a bridge crossing the Rhine at Worms in 1952, as shown in Figure 1.28,s with three spans of 330, 371, and 340 ft (100, 113, and 104 m). More than 300 such structures, with spans in excess of 250 ft (76 m), were constructed between 1950 and 1965

SECflON

FIGURE 1.23. Completed Ardrossan o\crpass (courtesy of the Portland Cement Association).

in Europe.s Since then the concept has spread throughout the world.’ Precast segmental construction also was evolving during this period. In 1952 a single-span county bridge near Sheldon, New York, was designed by the Freyssinet Company. Although this bridge was constructed of longitudinal rather than the European transverse segments, it represents the first practical application of match casting. The bridge girders were divided into three longitudinal segments that were cast end-to-end. The center segment was cast first and then the end segments were cast directly against it. Keys were cast at the joints so that the three precast elements could be joined at the site in the same position they hid in the precasting yard. Upon shipment to the job site the three elements of a girder were post-tensioned together with cold joints. l”,ll The first major application of match-cast, precast segmental construction was not consummated

A-A

FIGURE 1.24. Field spike for continuity (courtesy of the Prestressed Concrete Institute, from ref. 6).

12

Prestressed

Concrete

Bridges

and

Segmental

Construction

1.

FIGURE 1.25. Sebastian Inlet Bridge (courtesy of the Portland Cement Association). until 1962. This structure, designed by Jean Muller and built by Entreprises Campenon Bernard, was the Choisy-le-Roi Bridge over the Seine River south of Paris, Figure 1.29. This concept has been refined and has spread from France to all parts of the world. The technology of cast-in-place or precast segmental bridges has advanced rapidly in the last decade. During its initial phase the balanced cantilever method of construction was used. Currently, other techniques such as span-by-span, incremental launching, or progressive placement also are available. Any of these construction methods may call on either cast-in-place or precast segments or a combination of both. Consequently, a variety of design concepts and construction methods are now available to economically produce segmental bridges for almost any site condition. Segmental bridges may be classified broadly by four criteria:

The ultimate use of the bridge-that is, highway or railway structure or combination thereof. Although many problems are common to these two categories, the considerable increase of live loading in a railway bridge poses special problems that call for specific solutions. 2. The type of structure in terms of statical scheme and shape of the main bending members. Many segmental bridges are box girder bridges, but other types such as arches or cable-stayed bridges show a wide variety in shape of the supporting members. 3. The use of cast-in-place or precast segments or a combination thereof. 4. The method of construction. The sections that follow will deal briefly with the last three classifications.

1.7

Various Types of Structures

From the point of view of their statical scheme, there are essentially five categories of structures: (1) girders, (2) trusses, (3) rigid frames, (4) arch frames, and (5) cable-stayed bridges. 1.7.1 GIRDER BRIDGES

Box girders in the majority of cases are the most efficient and economical design for a bridge. When constructed in balanced cantilever, box girder decks were initially made integral with the piers while a special expansion joint was provided at the center of each span (or every other span) to allow

Conventional

\

Section A-A

FIGURE 1.26. Sebastian Inlet Bridge (courtesy of the Prestressed Concrete Institute, from ref. 6).

13


FIGURE 1.29. Choisy-le-Roi FIGURE 1.27.

I,uzanc~ Bridge over the Marne River.

FIGURE 1.28. & LVidmann).

b’ornx Bridge (courtesy of Dyckerhoff

CF

6lb’

E N D PIEI

176’-0’

MAIN

for volume changes and to control differential deflections between individual cantilever arms. It is now recognized that continuity of the deck is desirable, and most structures are now continuous over several spans, bearings being provided between deck and piers for expansion. Today, the longest box girder bridge structure that has been built in place in cantilever is the Koror Babelthuap crossing in the Pacific Trust territories with a center span of 790 ft (241 m), Figure 1.30.r2 A box girder bridge has been proposed for

PIER

f-

I_

1

L \

FIGURE 1.30.

Bridge.

/J 12%

._

176’-0“

I

I

12/-O”

Koror-Babelthuap Bridge, elevation and cross section (ref. 12).

14

Prestressed

Longitudmal

Concrete Bridges and Segmental Construction

section

r

1

G-r-r

Typical sections at span and over main piers

center

IF-4

FIGURE

FIGURE 1.31. The Great Belt Project.

the Great Belt Project in Denmark with a 1070 ft (326 m) clear main span, Figure 1.31. The box girder design has been applied with equal success to the construction of difficult and spectacular structures such as the Saint Cloud Bridge over the Seine River near Paris, Figure 1.32, or to the construction of elevated structures in very congested urban areas such as the B-3 Viaducts near Paris, Figure 1.33.

1.33.

R-3 Viaciuc t\. FI ‘111~ e.

The cantilever method has potential applications between the optimum span lengths of typical box girders for the low ranges and of stayed bridges for the high ranges. 1.7.3 FRAMES WITH SLANT LEGS

When the configuration of the site allows, the use of inclined legs reduces the effective span length.

1.7.2 TRUSSES

When span length increases, the typical box girder becomes heavy and difficult to build. For the purpose of reducing dead weight while simplifying casting of very deep web sections, a truss with open webs is a very satisfactory type that can be conveniently built in cantilever, Figure 1.34. The technological limitations lie in the complication of connections b e t w e e n p r e s t r e s s e d d i a g o n a l s a n d chords. An outstanding example is the Rip Bridge in Brisbane, Australia, Figure 1.35.

FIGURE

1.32. Saint Cloud Bridge, France.

FIGURE 1.34. Long-span concrete trusses.

FIGURE

1.35.

Rip Bridge, BI ishne, Xu\tl

nli,l


15

FIGURE 1.36. Long-span frame.

Provisional back stays or a temporary pier are needed to permit construction in cantilever, Figure 1.36. This requirement may sometimes present difficulty. An interesting example of such a scheme is the Bonhomme Bridge over the Blavet River in France, Figure 1.37. The scheme is a transition between the box girder with vertical piers and the true arch, where the load is carried by the arch ribs along the pressure line with minimum bending while the deck is supported by spandrel columns.

FIGURE 1.37. Bonhomme

Bridge.

1.7.4 CONCRETE ARCH BRIDGES

Concrete arches are an economical way to transfer loads to the ground where foundation conditions are adequate to resist horizontal loads. Eugene Freyssinet prepared a design for a 1000 meter (3280 ft) clear span 40 years ago. Because of construction difficulties, however, the maximum span built to date (1979) has been no more than 1000 ft (300 m). Construction on falsework is made difficult and risky by the effect of strong winds during construction. The first outstanding concrete arch was built at Plougastel by Freyssinet in 1928 with three 600 ft (183 m) spans, Figure 1.38. Real progress was achieved only when free cantilever and provisional stay methods were applied to arch construction, Figure 1.39. The world record is presently the Kirk Bridge in Yugoslavia, built in cantilever and com-

FIGURE 1.38. Plougastel Bridge, France.

16


I FIGURE 1.39. Concrete arches.

,,,)

.

.

.

\

,.

.

.

,,~

\

pleted in 1979 with a clear span of 1280 ft (390 m), Figure 1.40.

,_

i

1.7.5 CONCRETE CABLE-STAYED BRIDGES’”

FIGURE 1.40. Kirk Bridges, Yugoslavia.

When a span is beyond the reach of a conventional girder bridge, a logical step is to suspend the deck by a system of pylons and stays. Applied to steel structures for the last twenty years, this approach gained immediate acceptance in the field of concrete bridges when construction became possible

FIGURE 1.41. Long-span concrete cable-stayed bridges.

an m dr th m 1. al

Cast-in-Place and Precast Segmental

Construction

17

the structure’s deformability, particularly during construction. Deflections of a typical cast-in-place cantilever are often two or three times those of the same cantilever made of precast segments. The local effects of concentrated forces behind the anchors of prestress tendons in a young concrete (two or four days old) are always a potential source of concern and difficulties. I.82 CHARACTERISTICS OF PRECAST SEGMENTS

FIGURE 1.42. Krotonne Bridge, France.

and economical in balanced cantilever with a large number of stays uniformly distributed along the deck, Figure 1.41, The longest span of this type is the Brotonne Bridge in France with a 1050 ft (320 m) clear main span over the Seine River, Figure 1.42. Single pylons and one line of stays are located along the centerline of the bridge.

1.8

Cast-in-Place and Precast Segmental Construction

1.8.1 CHARACTERISTICS OF CAST-IN-PLACE SEGMENTS

In cast-in-place construction, segments are cast one after another in their final location in the structure. Special equipment is used for this purpose, such as travelers (for cantilever construction) or formwork units moved along a supporting gantry (for spanby-span construction). Each segment is reinforced with conventional untensioned steel and sometimes by transverse or vertical prestressing or both, while the assembly of segments is achieved by longitudinal post-tensioning. Because the segments are cast end-to-end, it is not difficult to place longitudinal reinforcing steel across the joints between segments if the design calls for continuous reinforcement. Joints may be treated as required for safe transfer of all bending and shear stresses and for water tightness in aggressive climates. Connection between individual lengths of longitudinal post-tensioning ducts may be made easily at each joint and for each tendon. The method’s essential limitation is that the strength of the concrete is always on the critical path of construction and it also influences greatly

In precast segmental construction, segments are manufactured in a plant or near the job site, then transported to their final position for assembly. Initially, joints between segments were of conventional type: either concrete poured wet joints or dry mortar packed joints. Modern segmental construction calls for the match-casting technique, as used for the Choisy-le-Roi Bridge and further developed and refined, whereby the segments are precast against each other, preferably in the same relative order they will have in the final structure. No adjustment is therefore necessary between segments before assembly. The joints are either left dry (in areas where climate permits) or made of a very thin film of epoxy resin or mineral complex, which does not alter the match-casting properties. There is no need for any waiting period for joint cure, and final assembly of segments by prestressing may proceed as fast as practicable. Because the joints are of negligible thickness, there is usually no mechanical connection between the individual lengths of tendon ducts at the joint. Usually no attempt is made to obtain continuity of the longitudinal conventional steel through the joints, although several methods are available and have been applied successfully (as in the Pasco Kennewick cable-stayed bridge, for example). Segments may be precast long enough in advance of their assembly in the structure to reach sufficient strength and maturity and to minimize both the deflections during construction and the effects of concrete shrinkage and creep in the final structure. If erection of precast segments is to proceed smoothly, a high degree of geometry control is required during match casting to ensure accuracy. 1.8.3

CHOICE BETWEEN CAST-IN-PLACE AND PRECAST CONSTRUCTION

Both cast-in-place methods and precast methods have been successfully used and produce substan-

18


tially the same final structure. The choice depends on local conditions, including size of the project, time allowed for construction, restrictions on access and environment, and the equipment available to the successful contractor. Some items of interest are listed below: 1. Speed of Construction Basically, cast-in-place cantilever construction proceeds at the rate of one pair of segments 10 to 20 ft (3 to 6 m) long ever) four to seven days. On the average, one pair of travelers permits the completion of 150 ft (46 m) of bridge deck per month, excluding the transfer from pier to pier and fabrication of the pier table. On the other hand, precast segmental construction allows a considerably faster erection schedule. a . For the Oleron Viaduct, the average speed of completion of the deck was 750 ft (228 m) per month for more than a year. b. For both the B-3 Viaducts in Paris and the Long Key Bridge in Florida, a typical 100 to 150 ft (30 to 45 m) span was erected in two working days, representing a construction of 1300 ft (400 m) offinished bridge per month, c. Saint Cloud Bridge near Paris, despite the exceptional difficulty of its geometry and design scheme, was constructed in exactly one year, its total area amounting to 250,000 sq ft (23,600 sq m). It is evident, then, that cast-in-place cantilever construction is basically a slow process, while precast segmental with matching joints is among the fastest. 2. Investment in Special Equipment Here the situation is usually reversed. Cast-in-place requires usually a lower investment, which makes it competitive on short structures with long spans [for example, a typical three-span structure with a center span in excess of approximately 350 ft (100

Ml.

In long, repetitive structures precast segmental may be more economical than cast-in-place. For the Chillon Viaducts with twin structures 7000 ft (2 134 m) long in a difficult environment, a detailed comparative estimate showed the cast-in-place method to be 10% more expensive than the precast. 3. Size and Weight of Segments Precast segmental is limited by the capacity of transportation and placing equipment. Segments exceeding 250 tons are seldom economical. Cast-in-place construction does not have the same limitation, al-

though the weight and cost of the travelers are directly proportional to the weight of the heaviest segment. 4. Environment Restrictions Both precast and cast-in-place segmental permit all work to be performed from the top. Precast, however, adjusts more easily to restrictions such as allowing work to proceed over traffic or allowing access of workmen and materials to the various piers. 1.9

Various

Methods

of

Construction

Probably the most significant classification of segmental bridges is by method of construction .41though construction methods may be as varied as the ingenuity of the designers and contractors, they fall into four basic categories: (1) balanced cantilever, (2) span-by-span construction, (3) progressive placement construction, and (4) incremental launching or push-out construction. 1.9.1 CAST-I.\‘-PL4CE

BAL,-I,VCED C.4.iTILEC’ER

The balanced or free cantilever construction concept was originally developed to eliminate falsework. Temporary shoring not only is expensive but can be a hazard in the case of sudden floods, as confirmed by many failures. Over navigable waterways or traveled highways or railways, falsework is either not allowed or severely restricted.’ Cantilever construction, whether cast in place or precast, eliminates such difficulties: construction may proceed from the permanent piers, and the structure is self-supporting at all stages. The basic principle of the method was outlined in Section 1.1 (Figure 1.3). In cast-in-place construction the formwork is supported from a movable form carrier, Figure 1.1. Details of the form travelers are shown in Figure 1.43. The form traveler moves forward on rails attached to the deck of the completed structure and is anchored to the deck at the rear. With the form traveler in place, a new segment is formed, cast, and stressed to the previously constructed segment. In some instances a covering may be provided on the form carrier so that work may proceed during inclement weather, Figure 1.44. The operation sequence in cast-in-place balanced cantilever construction is as follows: 1.

Setting up and adjusting carrier.

2.

Setting up and aligning forms.

Various Methods of Construction

CENTERJACK

FORM TRAVELLER

8,i-J?i,! -Lu ,.

! I

i-HUN I AL WORKING PLATFORM

ADDITIONAL

REAR GANG-BOARD

FIGURE

\BOTTOM FRAME WORK

~ONTAL LOWER WORKING PLATFORM

1.43. Form traveler (courtesy of Dyckerhoff & Widmann).

3. 4. 5.

Placing reinforcement and tendon ducts. Concreting. Inserting prestress tendons in the segment and stressing. 6. Removing the formwork. 7. Moving the form carrier to the next position and starting a new cycle. Initially, the normal construction time for a segment was one week per formwork unit. Advances in precast segmental construction have been applied recently to the cast-in-place method in order to reduce the cycle of operations and increase the efficiency of the travelers. With today’s technology it does not seem possible to reduce the

FIGURE 1.44. Bendorf Bridge form traveler (courtesy of Dyckerhoff & Widmann).

construction time for a full cycle below two working days, and this only for a very simple structure with constant cross section and a moderate amount of reinforcing and prestress. For a structure with variable depth and longer spans, say above 250 ft (75 m), the typical cycle is more realistically three to four working days. Where a long viaduct type structure is to be constructed of cast-in-place segments, an auxiliary steel girder may be used to support the formwork, Figure 1.45, as on the Siegtal Bridge. This equip-

1.45. Siegtal Bridge, use of an auxiliary truss in cast-in-place construction.

FIGURE

20

Prestressed


ment may also be used to stabilize the free-standing pier by the anchoring of the auxiliary steel girder to the completed portion of the structure. Normally, in construction using the form traveler previously described, a portion of the end spans (near the abutments) must be cast on falsework. If the auxiliary steel girder is used, this operation may be eliminated. As soon as a double typical cantilever is completed, the auxiliary steel girder is advanced to the next pier. Obviously, the economic justification for use of an auxiliary steel girder is a function of the number of spans and the span length. I-9.2. PRECAST BALANCED CANTILEVER

For the first precast segmental bridges in Paris (Choisy-le-Roi, Courbevoie, and so on, 1961 to 1965) a floating crane was used to transfer the precast segments from the casting yard to the barges that transported them to the project site and was used again to place the segments in the structure. The concept of self-operating launching gantries was developed shortly thereafter for the construction of the Oleron Viaduct (1964 to 1966). Further refined and extended in its potential, this concept has been used in many large structures. The erection options available can be adapted to almost all construction sites. 1. Crane Placing Truck or crawler cranes are used on land where feasible; floating cranes may be used for a bridge over navigable water, Figure 1.46. Where site conditions allow, a portal crane may be used on the full length of the deck, preferably with a casting yard aligned with the deck near

one abutment to minimize the number of handling operations, Figure 1.47. 2. Beam and Winch Method If access by land or water is available under the bridge deck, or at least around all permanent piers, segments may be lifted into place by hoists secured atop the previously placed segments, Figure 1.48. At first this method did not permit the installation of precast pier segments upon the bridge piers, but it has been improved to solve this problem, as will be explained later. 3. Launching Gantries There are essentially two families of launching gantries, the details of which will be discussed in a later chapter. Here we briefly outline their use. In the first family developed for the Oleron Viaduct, Figures 1.49 and 1.50, the launching gantry is slightly more than the typical span length, and the gantry’s rear support reaction is applied near the far end of the last completed cantilever. All segments are brought onto the finished deck and placed by the launching gantry in balanced cantilever; after completion of a cantilever, after placing the precast segment over the new pier, the launching gantry launches itself to the next span to start a new cycle of operations. In the second family, developed for the Deventer Bridge in Holland and for the Rio Niteroi Bridge in Brazil, the launching gantry has a length approximately twice the typical span, and the reaction of the legs is always applied above the permanent concrete piers, Figures 1.51 and 1.52. Placing segments with a launching gantry is now in most cases the most elegant and efficient method, allowing the least disturbance to the environment. 1.9.3 SPAN-BY-SPAN CONSTRUCTION

FIGURE 1.46. Segment erection by barge-mounted crane, Capt. Cook Bridge, Australia (courtesy of G. Beloff, Main Roads Department, Brisbane, Australia).

The balanced cantilever construction method was developed primarily for long spans, so that construction activity for the superstructure could be accomplished at deck level without the use of extensive falsework. A similar need in the case of long viaduct structures with relatively shorter spans has been filled by the development of a span-by-span methodology using a form traveler. The following discussion explains this methodol13.14.15.16 %Y* In long viaduct structures a segmental span-byspan construction may be particularly advantageous. The superstructure is executed in one direc-


COUPE

21

TRANSVERSALE

FIGURE 1.47. Mirabeau Bridge at Tours, France.

tion, span by span, by means of a form traveler, Figure 1.53, with construction joints or hinges located at the point of contraflexure. The form carrier in effect provides a type of factory operation transplanted to the job site. It has many of the ad: . .

the field. The form traveler may be supported on the piers, or from the edge of the previously completed construction, at the joint location, and at the forward pier. In some instances, as in the approaches of Rheinbrticke, Dusseldorf-Flehe, the movable formwork may be supported from the ground, Figure 1.54. The form traveler consists of a steel superstructure, which is moved from the completed portion of the structure to the next span to be cast. For an above-deck carrier, large formwork elements are suspended from steel rods during concreting. After concreting and post-tensioning, the forms are released and rolled forward by means of the structural steel outriggers on both sides of the form traveler’s superstructure. For a below-deck carrier, a similar procedure is followed. Many long bridges of this type have been built in Germany, France, and other countries. Typical construction time for a 100 ft (30 m) span superstructure is five to eight working days, depending upon the complexity of the structure. Deck configuration for this type of construction is usually a monolithic slab and girder (T beam or double T), box girder, or a mushroom cross sec-

Prestressed

22

J-I

52.00 m


170ft 54.OOm _ 1 0 6 . 0 0 Ill 3 5 0 ff

180 ft

(6)

4

80.00

m

260 f t

c

FIGURE 1.51. Second family of launching gantries, Rio Niteroi Bridge.

J

1 0 6 . 0 0 Ill 3 5 0 ft

I

Cc)

FIGURE

1.49.

First family of launching gantries (Ole-

ron Viaduct). tion. This method has been used recently in the United States on the Denny Creek project in the state of Washington. In its initial form, as described above, the spanby-span method is a cast-in-place technique. The same principle has been applied in conjunction with precast segmental construction for two very large structures in the Florida Keys: Long Key Bridge and Seven Mile Bridge, with spans of 118 ft (36 m) and 135 ft (40 m), respectively. Segments are assembled on a steel truss to make a complete

FIGURE 1.50. Placing precast segments on the Oleron Viaduct.

span. Prestressing tendons then assure the assembly of the various segments in one span while achieving full continuity with the preceding span, Figures 1.55 and 1.56. The floating crane used to place the segments over the truss also moves the truss from span to span. The contractor for the Seven Mile Bridge modified the erection scheme from that used for Long Key Bridge by suspending a span of segments from an overhead falsework truss. This is the first application of a method that seems to have a great potential for trestle structures in terms of speed of construction and economy. 1.9.4

PROGRESSIVE

PLACEMENT

CONSTRUCTION

Progressive placement is similar to the span-byspan method in that construction starts at one end of the structure and proceeds continuously to the

FIGURE 1.52. Rio Niteroi launching girder.

i of Construction

23

FIGURE 1.56.

Placing segments on assembly truss for Long Key Bridge.

FIGURE 1.55. Span-by-span assembly of precast segments.

other end. It derives its origin, however, from the cantilever concept. In progressive placement the precast segments are placed from one end of the structure to the other in successive cantilevers on the same side of the various piers rather than by balanced cantilevers on each side of a pier. At present, this method appears practicable and economical in spans ranging from 100 to 300 ft (30 to 90 m). Because of the length of cantilever (one span) in relation to construction depth, a movable temporary stay arrangement must be used to limit the cantilever stresses during construction to a reasonable level. The erection procedure is illustrated in Figure 1.57. Segments are transported over the completed portion of the deck to the tip of the cantilever span under construction, where they are positioned by a swivel crane that proceeds from one segment to the next. Approximately one-third of the span from the pier may be erected by the free cantilever method, the segments being held in position by exterior temporary ties and final prestressing tendons. For the remaining two-thirds of the span, each segment is held in position by temporary external ties and by two stays passing through a tower located over the preceding piers. All stays are continuous through the tower and anchored in the previously completed deck structure. The stays are anchored to the top flange of the box girder segments so that the tension in the stays can be adjusted by light jacks. Used for the first time in France on several structures, Figure 1.58, progressive placement is being applied in the United States for the construction of the Linn Cove Viaduct in North Carolina. In this bridge the precast pier construction proceeds also from the deck to solve a difhcult problem of environmental restrictions.

24


FIGURE 1.57. Progressive placement erection procedure.

The progressive placement method may also be applied to cast-in-place construction. 1.9.5. INCREMENTAL LAUNCHING OR PUSH-OUT CONSTRUCTION

This concept was first implemented on the Rio Caroni Bridge in Venezuela, built in 1962 and 1963 by its originators, Willi Baur and Dr. Fritz Leonhardt of the consulting firm of Leonhardt and Andra (Stuttgart, Germany).”

Segments of the bridge superstructure are cast in place in lengths of 30 to 100 ft ( 10 to 30 m) in stationary forms located behind the abutment(s), Figure 1.59. Each unit is cast directly against the previous unit. After sufficient concrete strength is reached, the new unit is post-tensioned to the previous one. The assembly of units is pushed forward in a stepwise manner to permit casting of the succeeding segments, Figure 1.60. Normally a work cycle of one week is required to cast and launch a segment, regardless of its length. Operations are


FIGURE 1.60. Incremental launching (courtesy of Prof. Fritz Leonhardt).

25

sequence

superstructure under its own weight at all stages of launching and in all sections. Four methods for this purpose are used in conjunction with one another. 1.

A first-stage prestress is applied concentrically to the entire cross section and in successive increments over the entire length of the superstructure. 2. To reduce the large negative bending moments in the front (particularly just before the superstructure reaches a new pier) a fabricated structural steel launching nose is attached to the lead segment, Figure 1.62. 3. Long spans may be subdivided by means of temporary piers to keep bending moments to a reasonable magnitude. This construction technique has been applied to spans up to 200 ft (60 m) without the use of temporary falsework bents. Spans up to 330 ft (100 m) have been built using temporary supporting bents. The girders must have a constant depth, which is usually one-twelfth to one-sixteenth of the longest span. 4. Another method has been used successfully in France to control bending moments in the

26

Prestressed

FIGURE 1.61. Incremental launching (courtesy of Prof. Fritz Leonhardt).

Concrete

Bridges

and

Segmental

Construction

o n J GUI ve ‘FIGURE 1.62. Steel launching nose (courtesy of Prof. Fritz Leonhardt).

deck in the forward part of the superstructure. A system using a tower and provisional stays is attached to the front part of the superstructure. The tension of the stays and the corresponding reaction of the tower on the deck are controlled automatically and continuously during all launching operations to optimize the stress distribution in the deck, Figure 1.63.

ft (1035 m). The incremental launching technique was used successfully for the first time in the United States for the construction of the Wabash River Bridge at Covington, Indiana.

After launching is complete, and the opposite abutment has been reached, additional prestressing is added to accommodate moments in the final structure, while the original uniform prestress must resist the varying moments that occur as the superstructure is pushed over the piers to its final position. Today, the longest incrementally launched clear span is over the River Danube near Worth, Germany, with a maximum span length of 550 ft (168 m). Two temporary piers were used in the river for launching. The longest bridge of this type is the Olifant’s River railway viaduct in South Africa with 23 spans of 147 ft (45 m) and a total length of 3400

The state of the art of designing and constructing prestressed concrete segmental bridges has advanced greatly in recent years. A wide variety of structural concepts and prestressing methods are used, and at least a thousand segmental bridges have been built throughout the world. We may conclude that segmental prestressed concrete construction is a viable method for building highway bridges. There are currently no known major problems that should inhibit utilization of segmental prestressed concrete bridges in the United States. They have been successfully consummated in other countries and are increasingly being employed in the United States.

1 .I 0

Applications of Segmental Construction in the United States

27

Applications of Segmental Construction in the United States

fbJ

FIGURE 1.64. ‘Three Sisters Bridge.

Cd)

FIGURE 1.63. Incremental launching with provi-

sional tower and stays.

One of the earliest projects for which segmental construction was considered was the proposed Interstate I-266 Potomac River Crossing in Washington, D.C., Figure 1.64, otherwise known as the Three Sisters Bridge. This structure contemplated a 750 ft (229 m) center span with side spans of 440 ft (134 m) on reverse five-degree curves, built with cast-in-place segmental construction. Because of environmental objections, this project never reached fruition. The JFK Memorial Causeway (Intracoastal Waterway), Corpus Christi, Texas, Figure 1.65, represents the first precast, prestressed, segmental, balanced cantilever construction completed in the United States. It was opened to traffic in 1973. Designed by the Bridge Division of the Texas Highway Department, it has a center span of 200 ft (61 m) with end spans of 100 ft (30.5 m). The first cast-in-place, segmental, balanced cantilever, prestressed concrete bridge constructed in the United States is the Pine Valley Bridge in California, on Interstate I-8 about 40 miles (64 km) east of San Diego. Designed by the California Department of Transportation, .the dual structure, Figure 1.66, has a total length of 1716 ft (53.6 m)

FIGURE 1.65. JFK hlcnwr ial Causewav.

Corpus

FIGURE 1.66. Pine Valley Bridge (courtesy CALTRANS).

of

Christi, Texas.

Pt-estressed


of

Houston

with spans of 270, 340, 450, 380, and 276 ft (82.3, 103.6, 137.2, 115.8, and 84.1 m). As indicated previously, numerous segmental bridge projects have been constructed or are contemplated in the United States. Many of them will be discussed in detail in the following chapters. Among the most significant are the Houston Ship Channel Bridge with a clear span of 750 ft (228 m), which will be the longest concrete span in the Americas, Figure 1.67, and the Seven Mile Bridge, which will be the longest segmental bridge in North America, Figure 1.68.

1 .I 1

FIGURE

1.67.

Rendering

FIGURE 1.68. Rendering of’ Seven Mile Bridge.

Ship

Channel

Bridge.

Applicability and Advantages of Segmental Construction

Segmental construction has extended the practical range of span lengths for concrete bridges. Practical considerations of handling and shipping limit the prestressed I-girder type of bridge construction to spans of about 120 to 150 ft (37 to 46 m). Beyond this range, post-tensioned cast-in-place box girders on falsework are the only viable concrete alternative. At many sites, however, falsework is not practical or even feasible, as when crossing deep ravines or large navigable waterways. Falsework construction also has a serious impact upon environment and ecology. Prestressed concrete segmental construction has been developed to solve these problems while extending the practical span of concrete bridges to about 800 ft (250 m) or even 1000 ft (300 m). With cable-stayed structures the span range can be extended to 1300 ft (400 m) and perhaps longer with the materials available today.13 Table 1.1 summarizes the range of application of various forms of construction by span lengths. Although the design and construction of verylong-span concrete segmental structures pose an important challenge, segmental techniques may

Applicability TABLE 1.1

and

Advantages

of

Segmental

Construction

29

Range of Application of Bridge Type by Span Lengthsa

Span

Bridge Types

o- 150 ft loo- 300 ft loo- 300 ft 250- 600 ft 200- 1000 ft 800-1500 ft

I-type pretensioned girder Cast-in-place post-tensioned box girder Precast balanced cantilever segmental, constant depth Precast balanced cantilever segmental, variable depth Cast-in-place cantilever segmental Cable-stay with balanced cantilever segmental

“1 fi = 0.3048 tn.

find even more important applications in moderate span lengths and less spectacular structures. Especially in difficult urban areas or ecology-sensitive sites, segmental structures have proven to be a valuable asset. Today most sites for new bridges can be adapted for segmental concrete construction. The principal advantages of segmental construction may be summarized as follows: 1. Segmental construction is an efficient and economical method for a large range of span lengths and types of structure. Structures with sharp curves and variable superelevation may be easily accommodated. 2. Concrete segmental construction often provides for the lowest investment cost. Savings of 10 to 20% over conventional methods have been realized by competitive bidding on alternate designs or by realistic cost comparisons. 3. Segmental construction permits a reduction of construction time. This is particularly true for precast methods, where segments may be manufactured while substructure work proceeds and be assembled rapidly thereafter. Further cost savings ensue from the lessening of the influence of inflation on total construction costs. 4. Segmental construction protects the environment. Segmental viaduct-type bridges can minimize the impact of highway construction through environmentally sensitive areas. Whereas conventional cut-and-fill type highway construction can scar the environment and impede wildlife migration, an elevated viaduct-type structure requires only a relatively narrow path along the alignment to provide access for pier construction. Once the piers have been constructed, all construction activity proceeds from above. Thus, the impact on the environment is minimized. 5. Interference with existing traffic during construction is significantly reduced, and expensive detours can be eliminated. Figure 1.69 indi-

cates how precast segments may be handled while traffic is maintained with a minimum disturbance. 6. Segmental construction contributes toward aesthetically pleasing structures in many different sites. A long approach viaduct (Brotonne, Figure 1.70), a curved bridge over a river (Saint Cloud, Figure 1.7 l), or an impressive viaduct over a deep valley (Pine Valley, Figure 1.66) are some examples where nature accepts human endeavor in spite of its imperfections. 7. Materials and labor are usually available locally for segmental construction. The overall labor requirement is less than for conventional construction methods. For the precast option a major part of the work force on site is replaced by plant labor. 8. As a consequence, quality control is easier to perform and high-quality work may be expected. 9. Segmental bridges when properly designed and when constructed by competent contractors under proper supervision will prove to be practically free of maintenance for many years. Only bearings and expansion joints (usually very few for continuous decks) need to be controlled at regular intervals.

FIGURE 1.69. Saint Cloud Bridge, segments placed

over

traffic.

Prestressed

30

Concrete

Bridges

FIGURE 1.70. Brotonne Bridge approach.

10. During construction, the technique shows an exceptionally high record of safety. Precast segmental construction today is competitive in a wide range of applications with other materials and construction methods, while it adds a further refinement to the recognized advantages of prestressed concrete.

FIGURE 1.71. Saint bridge over a river.

Cloud

Bridge, France, curved

References 1 . H. G. Tyrrell, History of Bridge Engineeting, Henry G. Tyrrell, Chicago, 1911. 2. Elizabeth B. Mock, The Architecture of Bridges, The Museum of Modern Art, New York, 1949. 3. T. Y. Lin, Design of Prestressed Concrete Structures, John Wiley & Sons, Inc., New York, 1958. 4. Anon., “Highway Design and Operational Practices Related to Highway Safety,” Report of the Special AASHO Traffic Safety Committee, February 1967. 5 . Anon., Prestressed Concrete for Long Span Bridges, Prestressed Concrete Institute, Chicago, 1968.

and

Segmental

Construction

6. Anon., “Long Spans with Standard Bridge Girders,” PC1 Bridge Bulletin, March-April 1967, Prestressed Concrete Institute, Chicago. 7. “Recommended Practice for Segmental Construction in Prestressed Concrete,” Report by PC1 Committee on Segmental Construction, Journal of the Prestressed Concrete Instztute, Vol. 20, No. 2, MarchApril 1975. 8. Ulrich Finsterwalder, “Prestressed Concrete Bridge Construction,” Journal oj the Amerzcan Concrete Instztute, Vol. 62, No. 9, September 1965. 9. F. Leonhardt, “Long Span Prestressed Concrete Bridges in Europe,” Journal of the Pre.,tressed Concrete Institute, Vol. 10, No. 1, February 1965. 10. Jean Muller, “Long-Span Precast Prestressed Concrete Bridges Built in Cantilever,” Fzrst International Symposium, Concrete Bridge Design, AC1 P u b l i c a t i o n SP-23, Paper 23-40, American Concrete Institute, Detroit, 1969. 11. Jean Muller, “Ten Years of Experience in Precast Segmental Construction,” Journal of the Prestressed Concrete Instatute, Vol. 20, No. 1, January-February 1975. 12. Man-Chung Tang, “Koror-Babelthuap Bridge-A World Record Span,” Preprint Paper 3441, ASCE Convention, Chicago, October 16-20, 1978. 13. C. A. Ballinger, W. Podolny, Jr., and M. J. Abrahams, “A Report on the Design and Construction of Segmental Prestressed Concrete Bridges in Western Europe- 1977,” International Road Federation, Washington, D.C., June 1978. (Also available from Federal Highway Administration, Offices of Research and Development, Washington, D.C., Report No. FHWA-RD-78-44.) 14. Ulrich Finsterwalder, “New Developments in Prestressing Methods and Concrete Bridge Construction,” Dywzdag-Berzchte, 4-1967, September 1967, Dyckerhoff & Widmann KG, Munich, Germany. 15. Ulrich Finsterwalder, “Free-Cantilever Construction of Prestressed Concrete Bridges and MushroomShaped Bridges,” First International Symposaum, Concrete Bridge Deszgn, AC1 Publication SP-23, Paper SP 23-26, American Concrete Institute, Detroit, 1969. 16. C. A. Ballinger and W. Podolny, Jr., “Segmental Construction in Western Europe-Impressions of an IRF Study Team,” Proceedings, Conference conducted by Transportation Research Board, National Academy of Sciences, Washington, D.C., TRR 665, Vol. 2, September 1978. 17. Willi Baur, “Bridge Erection by Launching is Fast, Safe, and Efficient,” Czvzl Engineerzng-AXE, Vol. 47, No. 3, March 1977. 18. Walter Podolny, Jr., and J. B. Scalzi, “Construction and Design of Cable-Stayed Bridges,” John Wiley & Sons, Inc., New York, 1976.

2 Cast-in-Place Balanced Cantilever Girder Bridges

2.5 2.6 2.7 2.8 2.9 2.10 2.11 2.12 2.18

INTRODUCTION BENDORF BRIDGE, GERMANY SAINT ADELE BRIDGE, CANADA BOUGUEN BRIDGE IN BREST AND LACROIK FALGARDE BRIDGE, FRANCE SAINT JEAN BRIDGE OVER THE GARONNE RIVER AT BORDEAUX, FRANCE SIEGTAL AND KOCHERTAL BRIDGES, GERMANY PINE VALLEY CREEK BRIDGE, U.S.A. GENNEVILLIERS BRIDGE, FRANCE GRAND’MFRE BRIDGE, CANADA ARNHEM BRIDGE, HOLLAND NAPA RIVER BRIDGE, U.S.A. KOROR-BABELTHUAP, U.S. PACIFIC TRUST TERRITORY VEJLE FJORD BRIDGE, DENMARK

2.1

Introduction

Developed initially for steel structures, cantilever construction was used for reinforced concrete bridges as early as fifty years ago. In 1928, Freyssinet used the cantilever concept to construct the springings of the arch rib in the Plougastel Bridge, Figure 2.1. The reactions and overturning moments applied by the falsework to the lower part of the arch ribs were balanced by steel ties connecting the two short cantilevers. A provisional prestress was thus applied by the ties to the arch ribs with the aid of jacks and deviation saddles. The first application of balanced cantilever construction in a form closely resembling its present one is due to a Brazilian engineer, E. Baumgart, who designed and built the Herval Bridge over the Rio Peixe in Brazil in 1930. The 220 ft (68 m) center span was constructed by the cantilever method in reinforced concrete with steel rods extended at the various stages of construction by threaded couplers. Several other structures fol-

2.14 HOUSTON SHIP CHANNEL BRIDGE, U.S.A. 2.15 OTHER NOTABLE STRUCXURFS 2.15.1 Medway Bridge, U.K. 2.15.2 Rio Tocantins Bridge, Brazil ‘2.153 Pueute Del Azufre, Spain 2.15.4 Schubeuamdie Bridge, Canada 2.15.5 Inci- Bridge, Guatemala 2.15.6 !3etubal Bridge, Argentina 2.15.7 Kipapa Stream Bridge, U.S.A. 2.15.8 Parrots Ferry Bridge, U.S.A. 2.15.9 Magnan Via’duct, France 2.15.10 Puteaux Bridge, Frame 2.15.11 Tricastiu Bridge, France 2.15.12 Eschachtal Bridge, Germauy 2.16 CONCLUSION REFERENCES

lowed in various countries, particularly in France. Albert Caquot, a leading engineer of his time, built several reinforced concrete bridges in cantilever. Shown in Figures 2.2 through 2.4 is Bezons Bridge over the River Seine near Paris, with a clear center span of 310 ft (95 m), being constructed in successive cantilever segments with auxiliary trusses. This bridge design was prepared in 1942. The method was not widely used at that time, because the excessive amount of reinforcing steel

Jack,

/ Ties

f Overturning moment due to centering

FIGURE 2.1. Cantilever construction of arch springings for Plougastel Bridge, France. 31

FIGURE 2.2. Bezons Bridge over the Seine River, France, typical longitudinal and transverse sections.

33

Introduction

,w--. ---.-._ -_-..--._ z I .! I-

- _._.._- _______ _ :

:

h’ ..*gr- _ _ ._- -.__. --I .- ._-__ ____L_ --/ :: : .I

il

FIGURE 2.3. Bezons Bridge, construction procedure.

required to balance the cantilever moments produced the tendency toward cracking inherent in an overreinforced slab subject to permanent tensile stresses. The introduction of prestressing in concrete structures dramatically changed the situation.

Used successfully in 1950 and 195 1 by Finsterwalder with the German firm of Dyckerhoff & Widmann for the construction of the two bridges of Balduinstein and Neckarrews, balanced cantilever construction of prestressed concrete bridges experienced a continuous popularity in Germany

FIGURE 2.4. Bezons Bridges under construction.

34

Cast-in-Place Balanced Cantilever Girder Bridges

FIGURE 2.5. La Voulte Bridge, France.

and surrounding countries. Nicolas Esquillan designed and built a large bridge by the cantilever method over the Rhine River in France, La Voulte Bridge (J952), where an overhead truss was used during construction, Figure 2.5. Between 1950 and 1965 more than 300 such bridges were constructed in Europe alone. Initially were prestressed by high-strength all &uctures bars, and hinges were provided at the center of the

various spans. Later other prestressing methods with parallel wire or strand tendons were also used. More important, a significant improvement in structural behavior and long-term performance was made possible by the achievement of deck continuity between the various cantilever arms. The first cantilever bridges with continuous decks were designed and built in France in 1962: the Lacroix Falgarde Bridge and Bouguen Bridge, Figures 2.6 and 2.22. Subsequently, the advantages of continuity were recognized and accepted in many countries. From 1968 to 1970 cantilever construction was considered for the Three Sisters Bridge in Washington, D.C., Figure 1.64. This project never reached the construction stage. The first cast-inplace balanced cantilever segmental bridge built in the United States is the Pine Valley Creek Bridge in California (1972 to 1974), Figure 2.7. To date, all segmental bridges constructed in the United States have been either precast or cast-in-place cantilever construction, with the following exceptions: Wabash River Bridge, incrementally launched (Chapter 7) Denny Creek and Florida Keys Bridges, span-byspan construction (Chapter 6)

FIGURE 2.6. Bouguen Bridge in Brest, France. First continuous rigid-frame structure built in balanced cantilever.

35

Bendorf Bridge, Germany

FIGURE 2.8. Bendorf Bridge (courtesy of Dvckerhoff & Widmann).

FIGURE 2.7. Pine Valley Creek Bridge. Linn Cove Viaduct, progressive placement construction (Chapter 6) The balanced cantilever method of construction has already been briefly described. In this chapter we shall see how this method has been implemented on various structures before we go on to consider specific design and technological aspects.

(west) are the river spans consisting of a symmetrical seven-span continuous girder with an overall length of 1721 ft (524.7 mj. In part two (east) are the nine-span continuous approach girders with the spans ranging from 134.5 ft (41 m) to 308 ft (94 mj and having an overall length of 1657 ft (505 mj, Figures 2.9 and 2.10. The continuous, seven-span, main river structure consists of twin, independent, single-cell box girders. Total width of the bridge cross section is 101 ft (30.86 mj. Each single-cell box has a top flange width of 43.3 ft (13.2 mj, a bottom flange width of 23.6 ft (7.2 mj, and webs with a constant thickness of 1.2 ft (0.37 m). Girder depth is 34.28 ft (10.45 m) at the pier and 14.44 ft (4.4 mj at midspan representing, with respect to the main span, a depth-to-span ratio of l/20 and l/47, respectively. Girder depth of the end of this sevenspan unit reduced to 10.8 ft (3.3 mj. The main navigation span has a hinge at midspan that is deHinge

2.2 Bendorf Bridge, Germany Longitudinal

An early and outstanding example of the cast-inplace balanced cantilever bridge is the Bendorf autobahn bridge over the Rhine River about 5 miles (8 km) north of Koblenz, West Germany. Built in 1964, this structure, Figure 2.8, has a total length of 3378 ft (1029.7 mj with a navigation span of 682 ft (208 mj. The design competition allowed the competing firms to choose the material, configuration, and design of the structure. Navigation requirements on the Rhine River dictated a 328 ft (100 m) wide channel during construction and a final channel width of 672 ft (205 mj. The winning design was a dual structure of cast-inplace concrete segmental box girder construction, consummated in two distinct portions. In part one

Cross sectton river pier

at

section

Cross section at pier G

FIGURE: 2.9. Bendorf Bridge, Part one (West), longitudinal section, plan, and cross secnons at the river pier and pier G, from ref. 1 (courtesy of Beton- und Stahlbetonbauj.

36


-~~ ss,o -L- SP.0 --L-- so0 -$A---5O$Om-

Longitudinal

-

-

~

section

Plan

FIGURE

2.10. Bendorf Bridge, Part Two (East), longitudinal section and plan, from ref. 1 (courtesy of Beton- und Stahlbetonbau).

signed to transmit shear and torsion forces only, thus allowing the superstructure to be cast monolithically with the main piers.1,2 After construction of the piers, the superstructure over the navigable portion of the Rhine was completed within one year. The repetition of the procedure in 240 segments executed one after the other offered numerous occasions to mechanize and improve the erection method.3,4 The deck slab has a longitudinally varying thickness from 11 in. (279.4 mm) at midspan to 16.5 in. (419 mm) at the piers. The bottom flange varies in thickness from 6 in. (152 mm) at midspan to 7.87 ft (2.4 m) at the piers. To reduce dead-weight bending-moment stresses in the bottom flange concrete, compression reinforcement was used extensively in regions away from the piers. Thicknesses of the various elements of the cross section are controlled partly by stress requirements and partly by clearance requirements of the tendons and anchorages. The structure is three-dimensionally prestressed: longitudinal prestressing uniformly distributed across the cross section; transverse prestressing in the top flange; and inclined prestressing in the webs. A total of 560 Dywidag bars la-in. (32 mm) in diameter resists the negative bending moment produced by a half-span, Figure 2.11.

The maximum concrete compressive stress in the bottom flange at the pier is 1800 psi (12.4 MPa). As a result of the three-dimensional prestress the tensile stresses in the concrete were negligible. The longitudinal prestressing is incrementally decreased from the pier to the hinge at midspan and to the adjacent piers; thus, shear stresses in the webs on both sides of the main piers are almost constant. Therefore, the web thickness remains constant and the diagonal prestressing remains very nearly constant. Construction began on March 1, 1962. After completion of the foundations and piers, balanced cantilever operations began from the west river pier in July 1963 and were completed at the end of that year. Segments were 12 ft (3.65 m) in length in the river span and 11.4 ft (3.48 m) in the remaining spans. Segments were cast on a weekly cycle. As the segments became shallower, the construction cvcle was advanced to two segments per week. During winter months, to protect operations from inclement weather, the form traveler was provided with an enclosure, Figure 2.12.

FIGURE

FIGURE 2.12. Bendorf‘ Bridge, protective covering

2.11. Bendorf Bridge, cross section showing tendons in the deck, ref. 2, (courtesy of the American Concrete Institute).

for form traveler (courtesy of Ulrich

Finsterwalder).

Saint Adele Bridge, Canada

FIGURE 2.13.

Ste. Adele Bridge, elevation, from ref. 5 (courtesy of

In the construction of the approach spans, the five spans from the east abutment were built in a routine manner with the assistance of falsework bents. The four spans over water were constructed by a progressive placement cantilever method (see Chapter 6), which employed a temporary cablestay arrangement to reduce the cantilever stresses.

Eng$mritzg ~V~7o.~-R~cord).

This structure, built in 1964 (the same year as the Bendorf Bridge), represents the first segmental bridge, in the contemporary sense, constructed in North America. It crosses the River of the Mules near Ste. Adele, Quebec, and is part of the Laurentian Autoroute. It is a single-cell box girder continuous three-span dual structure with a center span of 265 ft (80.8 m) and end spans of 132 ft 6 in. (40.4 m), Figure 2.13. At one end is a prestressed concrete 55 f-t (16.8 m) simple span. The bridge has a 100 ft (30.5 m) vertical clearance over the river in the canyon below.

The variable-depth girder is 16 ft 3 4 in. (4.96 m) deep at the piers and 6 ft (1.83 m) deep at midspan and its extremities, Figure 2.14. Each dual structure consists of a single-cell rectangular box 23 ft (7 m) wide with the top flange cantilevering on each side 9 ft (2.75 m) for a total width of 41 ft (12.5 m), Figure 2.15, providing three traffic lanes in each direction. Thickness of bottom flange, webs, and top flange are respectively 1 ft l# in. (0.35 m), 1 ft 6 in. (0.46 m), and 1 ft (0.3 m).5 A total of 70 prestressing tendons were required in each girder. Each tendon of the SEEE system consists of seven strands of seven 0.142 in. (3.6 mm) wires. The seven strands are splayed out through a steel ring in the anchorage and held in a circular pattern by steel wedges between each of the strands. The number of tendons anchored off at each segment end varies with the distance from the pier, increasing from an initial six tendons to eight tendons at the eighth segment, then decreasing to two tendons at the eleventh segment at midspan. There are an additional 44 positivemoment tendons in the center span located in the bottom flange.5

FIGURE 2.14. Stc. Adele HI idge, v i e w 01 variabledepth box girder (courtesy of the Portland Cement ASsociation).

FIGURE 2.15. Ste. Adele Bridge, view of end of box girder segment (courtesy of the Portland Cement Association) .

2.3

Saint Adele Bridge, Canada

38

Cast-in-Place Balanced Cantilever Girder Bridges counterweighted with 70 tons (63.5 mt) of concrete block, which was gradually diminished as construction proceeded and the depth of the segments decreased. The first pair of segments (at the pier), each with a length of 21 ft 23 in. (6.47 m), were cast on a temporary scaffolding braced to the pier, Figure 2.18, which remained fixed in position throughout the erection process.5 Construction of four segments per week, one at each end of a cantilever from two adjacent piers, was attained by the following five-day construction cycles:

FIGURE 2.16. Ste. Adele Bridge, dual structure under construction by the balanced cantilever method, from ref. 5 (courtesy of Engineering News-Record).

Forty-seven segments are required for each structure, eleven cantilevered each side of each pier, a closure segment at midspan of the center span, and a segment cast in place on each abutment. Segments cast by the form traveler were 10 ft 78 in. (3.24 m) in length.5 Four traveling forms were used on the project: one pair on each side of the pier for each of the dual structures, Figures 2.16 and 2.17. The forms were supported by a pair of 42 ft (12.8 m) long, 36 in. (914.4 mm) deep structural steel beams spaced 15 ft (5.57 m) on centers, that cantilevered beyond the completed portion of the structure. Initially the cantilevered beams were

Traveling forms moved, bottom flange First day: formed, reinforced, and cast. In the parallel span there was a one-day lag such that crews could shift back and forth between adjacent structures. Second day:

Third day: Concrete placed for webs and top

flange, cure begun. Fourth day: T e n d o n s p l a c e d a n d p r e s t r e s s i n g jacks positioned while concrete was curing. Fifth day: Prestressing accomplished. Forms stripped; preparations made to repeat cycle.

The cycle began on Monday. Since there was a lag of one day on the parallel structure, a six-day work week was required. Upon completion of the eleventh segment in each cantilever the contractor installed temporary falsework to support the abutment end and then cast the closure segment at midspan. Counterweights were installed at the abutment end to balance the weight of the closure forms and segment weight. After installation and stressing of the continuity tendons, abutment segments were cast and expansion joints installed.5

2.4

FIGURE 2.17. Ste. Adele Bridge, view of form travelers cantilevering from completed portion of the structure, from ref. 5 (courtesy of Engineering News-Record).

Reinforcement placed for webs and

top flange.

Bouguen Bridge in Brest and Lacroix Falgarde Bridge, France

The Bouguen Bridge in Brittany, West Province in France, is the first rigid-frame continuous structure built in balanced cantilever (1962 to 1963). The finished bridge is shown in Figure 2.6, while dimensions are given in Figure 2.19. It carries a three-lane highway over a valley 145 ft (44 m) deep-Le Vallon du Moulin H Poudre-and provides a link between the heart of Brest city and Le Bouguen, a new urban development. The total length of bridge is 684 ft (208 m). The main structure is a three-span rigid frame with

Bouguen Bridge in Brest

and Lacroix

Falgarde

39

Bridge, France

FIGURE 2.18. Ste. Adele Bridge, schematic of construction sequence, from ref. 5 (courtesy of Engineering News-Record).

box girder is 10 ft (3 m); web thickness also is constant throughout the deck and is equal to 9$ in. (0.24 m). Piers consist of two square box columns 10 ft by 10 ft (3 x 3 m) with wall thickness of 9$ in. (0.24 m) located under each deck girder. Two walls 84 in. (0:22 m) thick with a slight recess used for architectural purposes connect the two columns. Both piers are of conventional reinforced concrete construction, slip-formed at a speed reaching 14 ft (4.25 m) per day in one continuous operation.

piers elastically built-in on rock foundations with span lengths of 147,268, and 147 ft (45,82, and 45 m). At one end the deck rests on an existing masonry wall properly strengthened; at the other end a shorter rigid frame with a clear deck span of 87 ft (26.5 m) provides the approach to the main bridge. The deck consists of two box girders with vertical webs of variable height, varying from 15 ft 1 in. (4.6 m) at the support to 6.5 ft (2 m) at midspan and the far ends of the side spans. Width of each

Midspan

section

Pier

section

Plan section at pier

(b)

FIGURE 2.19. Bouguen Bridge, France, general dimensions. (a) Longitudinal section. (6) Cross sections.

40


FIGURE 2.20. Bouguen Bridge, construction of east

cantilever. The superstructure box girders are connected to the pier shaft by transverse diaphragms made integral with both elements to insure a rigid connection between deck and main piers. Construction of the deck proceeded in balanced cantilever with 10 ft (3 m) long segments cast in place in form travelers with a one-week cycle, Figures 2.20 and 2.21. High-early-strength concrete was used and no steam curing was required. Concrete was allowed to harden for 60 hours before application of prestress. The following cube strengths were obtained throughout the project: 60 hours (time of prestress) 7 days 28 days 90 days

3700 psi (25.5 MPa) 5500 psi (37.9.MPa) 7000 psi (48.3 MPa) 8200 psi (56.5 MPa)

Only one pair of form travelers was used for the entire project, but each traveler could accommodate the construction of both girders at the same time.

.\’

FIGURE

2.21.

Bouguen Bridge, view of’ the traveler.

During construction of the deck, much attention was given to the control of vertical deflections. Adequate camber was given to the travelers to fully compensate for short- and long-term concrete deflections. The cumulative deflection at midspan of the first cantilever arm was 14 in. (40 mm) at time of completion. Concrete creep caused this deflection to reach 3 in. (75 mm) at the time the second cantilever arm reached the midspan section. Proper adjustment of the travelers allowed both cantilever arms to meet within t in. (3 mm) at the time continuity was achieved. Flat jacks were provided over the outer supports to allow for any further desired adjustment. The structure is prestressed longitudinallv by tendons of eight 12 mm strands: 76 tendons over the top of the pier segment, 32 tendons at the bottom of the crown section, 20 tendons in the side spans, and transversely by tendons of seven 12 mm strands. The Lacroix Falgarde Bridge over Ariege River in France, built in 1961 and 1962, is similar to the Bouguen Bridge and represents the first continuous deck built in balanced cantilever (see the photograph of the finished bridge, Figure 2.22). It consists of three continuous spans 100, 200, and 100 ft (30.5, 61, and 30.5 m). The single box girder has a depth varying between 4 ft 5 in. and 10 ft 6 in. (1.35 to 3.2 m). Dimensions are given in Figure 2.23. The superstructure rests on both piers and abutments through laminated bearing pads. The deck was cantilevered and the construction started simultaneously from the two piers with four travelers working symmetrically. During con-

FIGURE 2.22. Lacroix-Falgardc Bridge, view of’ the

structure during construction.

Saint Jean Bridge Over the Gardonne River at Bordeaux, France

FIGURE 2.23.

Lacroix-Falgarde Bridge, elevation and cross section.

struction, the deck was temporarily fixed to the piers by vertical prestress. The structure is prestressed longitudinally by tendons of twelve 8 mm strands and transversely by tendons of twelve 7 mm strands.

2.5

Saint Jean Bridge over the Garonne River at Bordeaux, France

Completed in April 1965, the Saint Jean Bridge in Bordeaux is a remarkable application of the new concepts developed at that time in cast-in-place cantilever construction. The main structure has an overall length of 1560 ft (475 m) and is continuous with expansion joints only over the abutments. The deck is f’ree to expand on neoprene bearings located on all river piers, Figure 2.24. A very efficient method of pier and foundation construction was also developed, which will be described in more detail in Chapter 5. The bridge was built in the heart of the city of Bordeaux over the Garonne River between a 175year-old multiple-arch stone structure and a 120year-old railway bridge designed by Eiffel, the engineer who designed the Eiffel Tower. The main structure includes six continuous spans. The central spans are 253 ft (77 m) long and allow a navigation clearance of 38 ft (11.60 m) above the lowest water level, while the end spans are only 222 ft (67.80 m) long. Short spans at both ends, 50 ft (15.40 m) long, provide end restraint of the side spans over the abutments. The overall width of the bridge is 88 ft (26.80 m), consisting of six traffic lanes, two walkways, and two cycle lanes. Superstructure dimensions are shown in Figure 2.25.

41

The deck consists of three box girders. The constant depth of 10.8 ft (3.30 m) has been increased to 13 ft (3.90 m) over a length of 50 ft (15 m) on each side of the piers to improve the bending capacity of the pier section and reduce the amount of cantilever prestress. No diaphragms were used except over the supports. The results of a detailed analysis performed to determine the transverse behavior of the deck confirmed this choice (see detailed description in Chapter 4). Longitudinal prestressing consists of tendons with twelve 8 mm and twelve t in. strands. Transverse prestressing consists of tendons with twelve 8 mm strands at 2.5 ft (0.75 m) intervals. Vertical prestressing is also provided in the webs near the supports. As indicated in Figure 2.26, three separate pier columns support the three deck girders. They are capped with large prestressed transverse diaphragms. The piers are founded in a gravel bed located at a depth of 45 ft (14 m) below the river level by means of a reinforced concrete circular caisson

FIGURE 2.24. Saint Jean at Bordeaux, view of the completed structure.

COUPE LONGITUDINALE CULEE R D

CVLEE NE - _ 5

FIGURE 2.25. Saint Jean ar Bordeaux. (a) Longitudinal and (6) cross sections.

FIGURE 2.26. 42

Saint Jean Bridge at Bordeaux, typical section at river piers.

Siegtal and Kochertal Bridges, Germany

FIGURE 2.27. Saint Jean Bridge at Bordeaux, work progress on piers and deck. 18.5 ft (5.60 m) in diameter and 10 ft (3 m) high, floated and sunk to the river bed and then opendredged to the gravel bed. Precast circular matchcast segments prestressed vertically make up the permanent walls of caissons, while additional segments are used temporarily as cofferdams and support for the deck during cantilever construction. A lower tremie seal allows dewatering and placing of plain concrete fill inside the caisson. The reinforced concrete footing and pier shaft are finally cast in one day. The superstructure box girders were cast in place in 10 ft (3.05 m) long segments using twelve form travelers, allowing simultaneous work on the three parallel cantilevers at two different piers. The 20 ft (6.1 m) long pier segment was cast on the temporary supports provided by the pier caissons, allowing the form travelers to be installed and cantilever construction to proceed. Six working days were necessary for a complete cycle of operations on each traveler. Work progress is shown in Figures 2.27 and 2.28. Total construction time for the entire 130,000 sq ft (12,000 m*) was approximately

FIGURE 2.28. Saint Jean Bridge at Bordeaux, cantilever construction on typical pier.

43

one year, as shown on the actual program of work summarized in graphic form in Figure 2.29. To meet the very strict construction deadline of the contract, it was necessary to bring to the project site another set of three travelers to cast the last cantilever on the left bank and achieve continuity with the southern river pier cantilever. Altogether, meeting the two-year construction schedule was recognized as an engineering achievement. Exactly one hundred years earlier, Gustave Eiffel had built the neighboring railway bridge in exactly two years-food for thought and a somewhat humbling reflection for the present generation.

2.6

Siegtal and Kochertal Bridges, Germany

The Siegtal Bridge near the town of Sieger, north of Frankfort, Germany, represents the first industrial application of cast-in-place cantilever construction with an auxiliary overhead truss. This method was initially developed by Hans Wittfoht and the firm of Polensky-und-Zollner and subsequently used for several large structures in Germany and other countries. One of the most recent and remarkable examples of this technique is the Kochertal Bridge between Ntiremberg and Heilbron, Germany. Both structures will be briefly described in this section, while a similar application in Denmark is covered in another section of this chapter. Siegtal Bridge is a twelve-span structure 3450 ft (1050 m) long resting on piers up to 330 ft (100 m) high, with maximum span lengths of 344 ft (105 m), Figure 2.30. Two separate box girders carry the three traffic lanes in each direction for a total width of 100 ft (30.5 m), Figure 2.31. Structural height of the constant-depth box girder is 19 ft (5.8 m), corresponding to a span-to-depth ratio of 18. The deck is continuous throughout its entire length, with fixed bearings provided at the three highest center piers and roller bearings of highgrade steel for all other piers and end abutments. Piers have slip-formed reinforced concrete hollow box shafts with a constant transverse width of 68 ft (20.7 m) and a variable width in elevation with a slope of 40 to 1 on both faces. The superstructure was cast in place in balanced cantilever from all piers in 33 ft (10 m) long segments with an auxiliary overhead truss supporting the two symmetrical travelers, and a cycle of one week was obtained without difficulty for the construction of two symmetrical 33 ft (10 m) long seg-

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FIGURE 2.29.

Saint Jean Bridge at Bordeaux, actual program of work.

Elevation t Cross section 1

‘Cross section 2 Horizontal section

FIGURE 2.30. Siegtal Bridge, general dimensions.

45

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FIGURE 2.31. Siegtal Bridge, typical cross section.

merits. The auxiliary truss was also first used to cast the pier segment above each pier, Figure 2.32, before cantilever construction could proceed, Figure 2.33. Because the pier shafts are flexible and have limited bending capacity, it was inadvisable to subject them to unsymmetrical loading conditions during deck construction. Thus, the overhead truss also served the purpose of stabilizing the cantilever arms before continuity was achieved with the previous cantilever. The auxiliary steel truss is made of high-strength steel (50 ksi yield strength). Prestressing is applied to the upper chord, which is subjected to high tensile stresses in order to reduce the weight of the equipment. The overall length of the truss is 440 ft (135 m) long to accommodate the maximum span length of 344 ft (105 m). The total weight of the truss and of the two suspended travelers, allowing casting of two 33 ft (10 m) long segments, was 660 t (600 mt). Deck concrete was pumped to the various segments through pipes carried from the finished deck bv the auxiliary truss, Figures 2.34 and 2.35. Work commenced on the superstructure in March 1966. The first box girder was completed in

FIGURE 2.32. Sicgtal Bridge, pier segment ca5ting.

April 1968. The truss and travelers were immediately transferred to the second box girder, which was completed in September 1969. Thus, the average rate of casting was as follows: 3450 ft (1050 m) in 25 months, or 140 ft (42 m) per month

First bridge:

3450 ft (1050 m) in 17 months, or 200 ft (62 m) per month Both bridges: 6900 ft (2100 m) in 42 months, or 160 ft (50 m) per month Secohd bridge:

An outstanding contemporary example of the same technique is the Kochertal Bridge in Germany, shown in final progress in Figure 2.36. General dimensions of the project are given in Figure 2.37. Total length is 3700 ft (1128 m) with typical spans of 453 ft (138 m) supported on piers up to 600 ft (183 m) in height. The single box girder superstructure with precast outriggers carries six traffic lanes for a total width of 101 ft (30.76 m). Box piers were cast in climbing forms with 14.2 ft (4.33 m) high lifts. The top section is constant for all piers with outside dimensions of 16.4 by 28.2 ft

FIGURE 2.33. Siegtal Bridge, canClever construction.

II


46

FIGURE 2.34.

: 4

105m

105m

Siegtal Bridge, elevation of overhead truss and travelers.

(5 by 8.6 m). The four faces are sloped to increase the dimensions at foundation level to a maximum of 31.2 by 49.2 ft (9.5 by 15 m) for the highest piers. Wall thickness varies progressively from top to bottom, to follow the load stresses, from 20 in. (0.5 m) to 36 in. (0.9 m). The constant-depth superstructure is cast in two stages, Figure 2.38: (1) the single center box with a width of 43 ft (13.1 m) and a depth of 23 ft (7 m), and (2) the two outside cantilevers resting on a series of precast struts. To meet the very tight construction schedule of 22 months it was necessary to use two sets of casting equipment, working simultaneously from both abutments toward the center. Each apparatus was made of an overhead truss

equipped with a launching nose to move from pier to pier and two suspended travelers working in balanced cantilevers, casting segments on a oneweek cycle, Figure 2.39.

FIGURE 2.35. Siegtal Bridge, typical section of truss

F I G U R E 2 . 3 6 . Kochertal Bridgr,

and

travelers.

2.7 Pine Valley Creek Bridge, U.S.A. The first prestressed concrete cast-in-place segmental bridge built in the United States was the Pine Valley Creek Bridge on Interstate I-8 between San Diego and El Centro, California, Figures 1.66 and 2.7, opened to traffic late in 1974. This structure is located approximately 40 miles (64 km) east of San Diego and 3 miles (4.8 km) west of the

project.

gt~nenl vim o f

FIGURE 2.37. Kochertal Bridge, elevation, plan and

cross

section.

community of Pine Valley and within the Cleveland National Forest. Interstate I-8 crosses over a semiarid region that is highly erodible when the ground cover is disturbed; consequently stringent controls were imposed on access roads and ground-cover disturbances. Structure type was influenced by the following factors: site restrictions, economics, ecological considerations, and Forest Service limitations. After comparing various possible schemes such as steel arch, deck truss, or steel box girder, the California Department of Transportation selected a concrete box girder bridge medicated on the use of cantilever seg-

(b) FIGURE 2.38. Kochertal Bridge, typical cross sections. (a) First stage casting. (6) Final stage.


48

FIGURE 2.39. Kochertal Bridge, cantilever construc-

tion. mental construction, particularly well suited to the site because the depth and steep slopes of the valley made the use of falsework impractical. Also, the

DATUM

LLLV.

cantilever method minimized scarring of the natural environment, which was a major consideration for a project located in a National Forest. The bridge has an average length of 17 16 ft (523 m) and consists of twin two-lane single-cell, trapezoidal box girders each 42 ft (12.8 m) out-toout. The deck is 450 ft (137 m) above the creek bed. The superstructure consists of five spans of prestressed box girders 19 ft (5.8 m) deep. The center span is 450 ft (137 m) in length, flanked by side spans of 340 ft (103.6 m) and 380 ft (115.8 m), with end spans averaging 270 ft (82.3 m) and 276 ft (84.1 m). The bridge was constructed with four cantilevers. Pier 2 has cantilevers 115 ft (35.1 m) in length, piers 3 and 4 have 225 ft (69.6 m) cantilevers, and pier 5 has 155 ft (47.2 m) cantilevers,6*7*8 Figure 2.40. Provisions were made in the design to permit the portions of spans 1 and 5 adjacent to the abutments to be constructed segmentally or on falsework at the contractors’ option. The later option was exercised by the contractor.g*10

woo;

ELEVATION

FIGURE

2.40.

Pine Valley Creek Bridge, elevation and typical section, from ref. 8.

49

Pine Valley Creek Bridge, USA.

Hinges were provided in spans 2 and 4 at the end of the main cantilevers. In the preliminary design, consideration was given to the concept of a continuous structure for abutment to abutment without any intermediate joints. Continuity has manv advantages insofar as this particular structure’is concerned. However, it has the significant disadvantage of large displacements under seismic loading conditions. Because of the extreme difference in height and stiffness between piers, it was determined that all the horizontal load was being transmitted to the shorter piers, which were not capable of accepting it.s The pier foundations posed some interesting construction problems. The top 20 ft (6 m) of the rock material at the structure site was badly fissured, with some fissuring as deep as 40 ft (12 m). Narrow footings only 1 ft (0.3 m) wider than the pier shafts, tied down with rock anchors, were preferred to the conventional spread footings to minimize the amount of excavation. Although the piers are spectacular because of their size, they are not unique in concept. The two main piers, 3 and 4, are approximately 370 ft (113 m) in height and are made up of two vertical cellular sections interconnected with horizontal ties. In a transverse direction the piers have a constant. width to facilitate slip-form construction, while in the longitudinal direction the section varies parabolically, with a minimum width of 16 ft (4.9 m) approximately one-third down from the top. At this point they flare out to 23 ft by 24 ft (7 by 7.3 m) at the soffit. The pier wall thickness is a constant 2 ft (0.6 rn).‘jps Earthquake considerations produce the critical design load for the piers. The 1940 El Centro earthquake was used as the forcing function in the design analysis. Design criteria required that the c,ompleted structural frame withstand this force level without exceeding stress levels of 75% of yield. The pier struts are an important element in the seismic design of the piers. They provide ductility to the piers by providing energy-absorbing joints and an increased stability against buckling for the principal shaft elements. Because of the size of the struts in relation to the pier legs, the majority of the rotation in the strut-leg joint occurs in the strut. Thus, a very high percentage of transverse confining reinforcement was required in the strut to insure the ductility at this location.“j9 Although preliminary design anticipated the slip-forming technique for construction of the piers, the contractor finally elected to use a selfclimbing form system. Steel forms permitted 22 ft

(6.7 m) high lifts, and they were given a teflon coating to facilitate stripping while producing a high-quality finished concrete surface. Construction of the pier caps was especially challenging. The pier caps, Figure 2.41, consist of two arms 60 ft (18.3 m) in height, which project outward at an approximate angle of 60” from each stem of the pier shafts. These arms are constructed in four lifts in such a manner that the forms for each lift are tied into the previous lift. Upon completion of the pier cap arms they are tied together and the top strut is formed, reinforcement placed, and cast. The pier cap is prestressed transversely in order to overcome side thrust from-the superstructure. The superstructure consists of two parallel trapezoidal box girders 42 ft ( 12.8 m) wide and 19 ft (5.8 m) deep with a 38 ft (11.6 m) space between the boxes, such that an additional box girder may be constructed for future widening, Figures 2.40 and 2.42. The boxes, in addition to being posttensioned longitudinally, have transverse prestressing in the deck slab, together with sufficient mild steel reinforcement to resist nominal construction loads, allowing the transverse prestressing operations to be removed from the critical path. The

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FIGURE 2.41. Pine Valley Creek Bridge, elevation, side view, and cross section of pier, from ref. 7 (courtesy of the Portland Cement Association).

50


FIGURE 2.42. Pine Valley Creek Bridge, typical box girder cross section, from ref. 7 (courtesy of the Portland Cement Association). sloping webs and large deck overhangs were used

to minimize the slab spans and the number of girder webs and to accentuate a longitudinal shadow line, thus reducing the apparent depth. The web thickness of 16 in. (406 mm) was selected to permit side-by-side placement of the largest tendon then being used in bridge construction and to keep the shear reinforcement to a reasonable size and spacing, Figure 2.42. The bottom slab at midspan is 10 in. (254 mm) thick and flares out to 6.5 ft (1.98 m) at the pier. 6*7,g Construction of the superstructure proceeded in a balanced cantilever fashion, Figures 2.7 and 2.43. As shown in Figure 2.44, the erection scheme proposed by the contractor allowed all superstructure work to be performed in a continuous sequence, essentially from the top. Four form travelers were used for the cantilever construction of this project, one at each end of each cantilever arm. Basically, one traveler consisted of an overhead steel truss used to support the formwork for the typical 16.5 ft (5 m) long segments. The truss is anchored, at the rear, to the previously cast segment, while the front end is equipped with hydraulic jacks used for grade adjustment. Highdensity plywood was used for all formed surfaces. A total of 172 cast-in-place segments were required for the entire structure. Falsework was required close to abutments 1 and 6 to complete the side spans beyond the balanced cantilever arms. Formwork used in that portion of the structure could be reused above each intermediate pier cap to construct the 35 ft (10.7 m) long pier segment before the actual cantilever construction proceeded. The cross section of the superstructure allowed

PIE

FIGURE 2.43. Pine Valley Creek Bridge, auxiliary bridge, from ref. 7 (courtesy of the Portland Cement Association).

an auxiliary truss to be located between the two concrete box girders, Figure 2.43. This auxiliary bridge consisted of a structural steel truss 10 ft (3.05 m) square in cross section and 320 ft (97.5 m) in length. In a stationary position it was supported at the leading end on the pier cap strut and at the rear end of a steel saddle between the two concrete boxes. It was designed such that the front end could be cantilevered out 225 ft (68.6 m), which is one-half the main span. Electric winches allowed longitudinal launching between the concrete box girders. When pier 5 was completed, the auxiliary bridge was erected in span 5-6, utilizing temporary support towers near abutment 6. Subsequent 30 ft (9.1 m) lengths of auxiliary truss were attached at the abutment and incrementally launched toward pier 5, until its front end was supported on the pier cap. The pier table was then constructed and

cantilever construction commenced until the structural hinge in span 4-5 was reached. Upon completion of the closure joint in span 5-6 the auxiliary truss was launched forward until the front end reached pier 4. The form travelers were dismantled from the tip of the cantilever and reerected on the pier table at pier 4, and cantilever FIGURE 2.44. (Opposite) Pine Valley Creek Bridge, erection scheme proposed by the contractor, from ref. 10.

0 \__

Stage 1 Cantilever Construction from pier 5

construction on conventional

($l +L

\ stage 2 from pier 4

\ Stage 3 from pier 3

Stage 5 cOmpletion

52


construction was started again. This cycle was repeated until closure was achieved in span l-2. The use of the auxiliary truss had the following advantageslO: 1. Men and materials for the superstructure could reach the location of construction from abutment 6 over the auxiliary bridge and the already completed portion of the superstructure without interfering with the valley below. 2. The construction equipment (tower cranes and hoists) at the piers was required only for the actual construction of the piers and could be relocated from pier to pier without waiting for completion of the superstructure. 3. Except for construction of abutment 1 and pier 2, site installation for the entire project was located at one location, near abutment 6. Concrete was supplied from a batching plant located approximately 2 miles (3.2 km) from the site. Ready-mix trucks delivered the concrete at abutment 6. The concrete was then pumped through 6 in. (152 mm) pipes down the slope to the foot of piers 5 and 4. The concrete for the superstructure was pumped through a pipeline installed in the auxiliary truss right into the forms. A second pump with a similar installation was located at abutment 1 to supply concrete for abutment 1 and pier 2.1° A 5000 psi (35 MPa) concrete was specified for the superstructure, presenting no unusual problems. However, to maintain a short cycle for the construction of the individual segments it was necessary to have sufficient strength for prestressing 30 hours after concrete placement. This was difficult to achieve, since the specifications did not allow type III cement and certain additives. A solution Gas to prestress the individual tendons necessary to support the following segment to 50 percent of their final force. The form carrier could then be advanced and the remainder of the prestressing force applied after the concrete reached sufficient strength and before casting the next segment.r” Prestressing was achieved using lf in. (32 mm) diameter Dywidag bars. Longitudinal tendons were provided in 33 ft (10 m) lengths and coupled as the work progressed. Temporary corrosion protection of the bars was obtained by blowing “VPI” powder into the ducts and coating each bar with vinyl wash or “Rust-Van 310.“*

2.8 Genneuilliers Bridge, France The Gennevilliers Bridge, Figures 2.45 and 2.46, is a five-span structure with a total length of 2090 ft (636 m). At its southern end it is supported on a common pier with the approach viaduct from the port of Gennevilliers. It crosses successively an entrance channel to the port, a peninsula situated between the channel, and the Seine River itself, Figure 2.47. It is part of the Al5 Motorway, which traverses from the Paris Beltway through Gennevilliers, Argenteuil, the valley d’Oise, and on to the city of Cergy-Pontoise. The present structure provides a four-lane divided highway with provision for a future twin structure. The superstructure is a variable-depth two-cell box girder with spans of 345, 564, 243, 564 and 371 ft (105, 172, 74, 172 and 113 m). Depth varies from 29.5 ft (9 m) at intermediate piers to 11.5 ft (3.5 m) at midspan of the 564 ft (172 m) spans and its extremities, with a depth of 23 ft (7 m) at midspan of the short center span, Figure 2.46. Depth-to-span ratios of the 564 ft (172 m) spans at midspan and at the piers are respectively l/49 and l/19. The curved portion of the structure has a radius, in plan, of 2130 ft (650 m). The longitudinal grade is a constant 1.5 percent within the zone of curvature. Because the short center span is subjected to negative bending moment over its entire length, the structure behaves much as a continuous three-span beam. In cross section, Figure 2.48, the two-cell box girder has a bottom flange varying in width from 42.2 ft (12.86 m) at midspan to 30.5 ft (9.3 m) at the pier, for the 564 ft (172 m) span. Thickness of the bottom flange varies from 47 in. (1.2 m) at the pier to 8 in. (20 cm) at midspan. The top flange has

FIGURE 2.45. Gennevilliers Bridge, view of curved five-span structure.

Gennevilliers Bridge, France

FIGURE 2.46.

Gennevilliers Bridge, plan and elevation, from ref. 11.

an overall width of 60.6 ft (18.48 m) with a 6 ft (1.83 m) overhang on one side and 6.2 ft (1.88 m) on the other. Thickness of the top flange is a constant 8 in. (20 cm). The center web has a constant thickness of 16 in. (400 mm). Exterior webs, which are inclined 18” to the vertical, vary in thickness

FIGURE 2.47.

from 16 in. (400 mm) at the pier to 12 in. (300 mm) at midspan. Diaphragms, Figure 2.49, are located at the supports. The superstructure is prestressed in three directions, with strand tendons being utilized longitudinally and transversely and bar tendons utilized for the webs. Interior anchorage

Gennevilliers Bridge, aerial view of the completed bridge.


54

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FIGURE 2.48. Gennevilliers Bridge, cross section, from ref. 11.

FIGURE 2.49. Gennevilliers Bridge, interior view showing diaphragm.

blocks for the longitudinal prestressing are located at top slab level. The superstructure is fully continuous over its total length of 2090 ft (636 m) between the northern abutment and the southern transition pier with the approach viaduct. The deck rests upon the four main piers supported by large elastomeric pads. The superstructure was cast in place using the balanced cantilever method according to the step-by-step scheme shown in Figure 2.50. Segments over the piers (pier segments) were constructed first on formwork, in a traditional manner, except for their unusual length [26 ft (7.9m)I and weight [850 t (770 mt)]. Four travelers were used for casting the typical 11 ft (3.35 m) long segments varying in weight from 242 t (220 mt) near the piers to 110 t (100 mt) at midspan. l1 The travelers were specially designed to achieve maximum rigidity and prevent the usual tendency to crack a newly cast segment under the deflections of the supporting trusses of conventional travelers. The framework used for this purpose was made of self-supporting forming panels assembled into a monolith weighing 120 t (110 mt) and prestressed to the preceding part of the superstructure to make the unit substantially deflection free, Figure 2.5 1. Stability, especially under wind loads or in the event of an accidental failure of the travelers during the construction period, was maintained by a pair of cables on each

FIGURE

2.50.

Gennevilliers Bridge, erection sequence, from ref. 11.

This three-lane cast-in-place segmental bridge is located on Quebec Autoroute 55 and crosses the St. Maurice River approximately 3 miles (4.8 km) north of Grand’Mere, Quebec, Figure 2.52. Water depth at the bridge site is over 110 ft (35.5 m), with an additional 150 ft (45.75 m) depth of sand, silt,

and debris above bedrock. The river flow at the bridge site is 3.6 ft/sec (1.1 m/set). During the preliminary design stage in 1973 and 1974, several structural solutions were considered. The use of short spans of precast concrete AASHTO sections or structural steel girders requiring a number of piers was immediately abandoned because of river depth and current velocity at the site. Site conditions required the development of an economical long clear span with as few piers as possible in the river. Options available

FIGURE 2.51. Gennevilliers Bridge, superstructure

FIGURE 2.52. GrandMere

side of the pier connecting the superstructure to pier base. 2.9 Grand’Mere Bridge, Canada

under

construction.

Bridge, general view showing parabolic soffit of center span, (courtesy of the Portland Cement Association).

56


were structural steel, post-tensioned precast segmental, and several options of cast-in-place prestressed concrete, varying in span, cross section, and pier requirements. The design finally selected for preparing the bid documents was a concrete cantilever single-cell box with a center span of 540 ft (165 m), a 245 ft (75 m) western land span, and a 150 ft (46 m) eastern land span for a total length of 935 ft (285 m). The design actually used for construction, Figure 2.53, for the same total length, has a main span increased to 59.5 ft (181 m) and two equal 170 ft (52 m) long side spans. The corresponding slight increase in cost of the superstructure was far more than offset by eliminating the need to build a caisson in 48 ft (15 m) of water 98 ft (30 m) above bedrock for the west pier. This redesign, developed in cooperation with the contractor, allowed an overall saving of approximately 20% of the project cost. The two identical 170 ft (5 1.9 m) long land spans cantilever from the main piers and act as counterweights for the main span. From a depth of 32 ft (9.8 m) at the main piers they taper to a depth of 28 ft (8.5 m) at a point 130 ft (39.6 m) from the main

t

\

piers, where they are supported by a secondary pair of 4 ft by 4 ft (1.2 by 1.2 m) bearing capped piers. The 40 ft (12.2 m) wedge-shaped shore ends of the land spans taper from the secondary piers to grade at the top of the abutment. The abutments, which are just 16 in. (406 mm) thick, are designed to support the approach slab only, Figure 2.54.

595’

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170’

ELEVATION

TYPICAL

DETAIL

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SECTION

ABUTMENT

FIGURE 2.54. Grand’Mere Bridge, general arrangement. (a) Elevation. (6) Typical section. (c) Detail of abutment.

Grand’mere Bridge, Canada

Modular, confined rubber expansion joints are provided in the roadway above the abutments. The wedge portions of the land spans are solid conCrete, helping counterbalance the weight of the main span under service conditions as well as during the construction stage. The land spans have a web thickness of 2 ft (0.6 m), a 3 ft (0.9 m) thick bottom slab, and a 15 in. (38 1 mm) thick top flange. A 2 ft (0.6 m) thick diaphragm is located 78 ft (23.8

57

m) outboard of the secondary piers to form a chamber between the solid wedge end and the diaphragms. This chamber was incrementally filled with gravel in three stages to counterbalance the main span as it was progressively constructed. The bottom soffit of the west land span was supported on temporary steel scaffolding. However, because of the terrain slope, the east land-span bottom soffit was plywood-formed on a bed of sand spread

Elevation fb)

FIGURE 2.55. Arnhem Bridge. (a) Plan. (b) Elevation.

58


over the rock. Upon completion of concreting and curing, the sand was hosed out from under the formwork, allowing it to be stripped.‘”

2.10 Arnhem Bridge, Holland

The Arnhem Bridge, Figure 2.55, is a cast-inplace, lightweight concrete, segmental bridge crossing the Rhine River with a center span of 448 ft (136 m), a south end span of 234 ft (71 m), and a north end span of 238 ft (72 m) connecting to approach ramps. It is a dual structure composed of two-cell box girders, Figure 2.56a. The western structure has two 30 ft (9.1 m) roadways for automobile traffic. The eastern structure has a 23 ft (7 m) roadway reserved for bus traffic, a 17 ft (5.3 m) roadway for bicycles and motorcycles, and a 7 ft (2.1 m) pedestrian walkway. Ramp structures are of prestressed flat slab construction, Figure 2.566. The main three-span river crossing with an overall width of 122.7 ft (37.4 m) consists of two

double-cell box girders that vary in depth from 6.5 ft (2.0 m) at midspan to 17 ft (5.3 m) at the piers. The western rectangular box girder has a width of 49 ft (14.8 m) with 10 ft (3 m) top flange cantilevers for an overall width of 68.4 ft (20.84 m). The eastern rectangular box girder has a width of 35.4 ft (10.8 m) with top flange cantilevers of 8.6 ft (2.62 m) for a total width of 52.6 ft (16.04 m), Figure 2.56a. Construction of the main spans is by the conventional cast-in-place segmental balanced cantilever method with form travelers. The form travelers are owned by the Dutch Government and leased to the contractors. Strand tendons were used for post-tensioning, and the lightweight concrete had a weight of about 110 lb/ft3 (1780 kg/m3), Figure 2.57. Temporary supports at the pier were used for unbalanced loading during construction, Figure 2.58. Precast exposed aggregate facia units were used for the entire length of the structure and its approaches, Figures 2.59 and 2.60.

E, 0.13

j

2.62

1 VAR

fb) FIGURE 2.56. Arnhem Bridge, typical cross sections of main bridge and flat-slab ramp. (a) Main structure. (6) Prestressed flat-slab ramp.

Napa River Bridge, U.S.A.

FIGURE 2.57. Arnhem Bridge, center-span cantilev-

ers.

FIGURE 2.58. Arnhem Bridge, temporary pier sup-

ports

for

unbalanced

moments.

FIGURE

flat-slab

59

2.59.

ramp

Arnhem

HI-idgc, Ge\\- of‘ prwtwssed

structure.

FIGURE 2.60. Arnhem Bridge, precast exposed

aggregate facia

units.

2.11 Napa River Bridge, U.S.A. The Napa River Bridge, Figure 2.61, is located on Highway 29 just south of the city of Napa, California, and provides a four-lane, 66 ft (20 m) wide roadway over the Napa River to bypass an existing two-lane lift span and several miles of city streets. The 68 ft (20.7 m) wide, 2230 ft (679.7 m) long bridge consists of 13 spans varying in length from 120 to 250 ft (36.58 to 76.2 m) and a two-cell trapezoidal box girder varying from 7 ft 9 in. (2.36 m) to 12 ft (3.66 m) in depth, Figure 2.62. Three hinged joints were provided at midspan in spans 2, 6, and 10. These joints involved elaborate connections incorporating elastomeric bearing pads and hard-rubber bumper pads to withstand severe movement and shock during an earthquake, Figure 2.63. All other joints between the cantilevers The were normal cast-in-place closure joints.13 superstructure is fixed to the piers, primarily for seismic resistance. The Structures Division of the California Department of Transportation (CALTRANS) developed plans and specifications for three alterna-

FIGURE 2.61. Napa River Bridge, aerial view.

tive types of construction, Figure 2.62, as follows: A. A conventional continuous cast-in-place prestressed box girder bridge of lightweight concrete. B. A continuous structural-steel trapezoidal box girder composite with a lightweight concrete deck. C. A cantilever prestressed segmental concrete box girder bridge allowing either cast-in-place


60

108 +20 PRVC Elev 63 18

2820’ vc k/C q -0 24863 % Sto

‘SO&

PROFILE GRADE

Pier 2

3

4

5

6

7

i

9

IO

II

I2

13

ELEVATION

CmveAmol tip P/S Llghtwelght Cone B o x Girder ALTERNATIVE A

FIGURE 2.62.

Welded Steel Box Girder ALTERNATIVE

B

Contllever Segmental P / S Lightweight Cone B o x Gtrder ALTERNATIVE

C

Napa River Bridge, profile grade, elevation, and alternate sections. or precast segments. Erection was allowed on falsework or by the free cantilever method.

Anchor Bolt

L Elastomeric Pad

FIGURE 2.63. Napa River Bridge, mid-span hinge joint with seismic bumbers.

Because of poor foundations and a readily available aggregate supply, all alternatives utilized lightweight concrete in the superstructure. Alternative C utilized transverse prestressing in the deck to reduce the number of webs to three, as compared to seven webs required in alternative A. Of seven bids received and opened on November 6, 1974, six were for alternative C and the seventh and highest was for alternative B. No bids were submitted for alternative A. Design of the superstructure required lightweight concrete with a compressive strength of 4500 psi (3 16 kg/cm2) at 28 days and 3500 psi (246 kg/cm2) prior to prestressing. The three-web winning alternative required a minimum of formed surfaces and forced the majority of longitudinal prestressing into the flanges, resulting in maximum prestress eccentricity, and therefore an economical solution. Contract plans showed the minimum prestress force required at each section and permitted the use of either 270 ksi (1862 MPa) strand or 150 ksi (1034 MPa) bar tendons. Prestressing force diagrams were provided for both materials. The contractor had the option of balancing segment length against prestress force to achieve the most economical structure. In addition, the plans provided the contractor with the option of a combination of diagonal prestressing and conventional reinforce-

Koror-Babelthuap, U.S. Pacajic

ment in the webs for shear reinforcement or the utilization of conventional stirrup reinforcement only. The design was based upon a 40,000 psi (276 MPa) prestress loss for the 270 ksi (1862 MPa) strand and 28,000 psi (193 MPa) loss for the 150 ksi (1034 MPa) bars. Because the loss of prestress is a function of the type of lightweight aggregate used, the contractor was required to submit test values for approval concerning the materials to be used and relevant calculations.t4 The contractor elected to use the cantilever cast-in-place alternative supported on falsework until each segment was stressed, Figure 2.64. Falsework bents with ten 70 ft (21.3 m) long, 36 in. (914 mm) deep, wide-flange girders support each balanced cantilever. The falsework was then moved to the next pier, leaving the cantilever free-standing, Figure 2.65. The entire formwork, steel girders, and timber forms were lowered by winches from the cantilever girder after all negative post-tensioning was completed. Positive posttensioning followed midspan closure pours.13

The 250 ft (76.2 m) long navigation span was constructed with a complicated segment sequence because of a U.S. Coast Guard requirement that a 70 ft (21.3 m) wide by 70 ft (2 1.3 m) high navigation channel be maintained. Approximately 60 ft (18.3 m) of span 4, over the navigable channel, was constructed in three segments on suspended falsework by the conventional cast-in-place segmental method.13 All transverse and longitudinal post-tensioning tendons consist of t in. (12.7 mm) diameter strands. Longitudinal tendons are twelve t in. (12.7 mm) diameter strand, with anchorages located in the top and bottom flanges such that all stressing was done from inside the box girder. Loops are used for economy and efficiency, as shown in Figure 2.66. The longest span over the navigation channel is prestressed by 50 (twelve 4 in. strand) tendons. Transverse prestress in the top flange allowed a 10 ft (3 m) cantilever on each side of the two-cell box girder. Transverse tendons consist of four -f in. diameter strands encased in flat ducts 2.25 by 0.75 in. (57 by 19 mm) with proper splay at both ends to accommodate a flat bearing at the edge of the deck slab.

2 .12

FIGURE 2.64. Napa Kiver Bridge, free-standing cantilever and supporting bents for falsework

FIGURE 2.65. Napa

River Bridge, falsework bents (courtesy of Phil Hale, CALTRANS).

61

Trust Territory

Koror-Babelthuap, U.S. Pacijk Trust Territory

This structure currently represents (1979) the longest concrete cantilever girder span in the world. It connects the islands of Koror and Babelthuap, which are part of the Palau Island chain of the Caroline Islands located in the United States Trust Territory some 1500 miles (2414 km) east of the Philippines, Figure 2.67.

FIGURE 2.66. Napa tendons.

River Bridge, longitudinal loop

62


FIGURE 2.67. Koror-Babelthuap Bridge, location map, from ref. 15. In elevation this structure has a center span of 790 ft (241 m) with side spans of 176 ft (53.6 m) that cantilever another 61 ft (18.6 m) to the abutments, Figure 1.30. Depth of this single-cell box girder superstructure varies parabolically from 46 ft (14 m) at the pier to 12 ft (3.66 m) at midspan of the main span, Figure 2.68. The side span decreases linearly from the main pier to 33 ft 8 in. (10.26 m) at the end piers and then to 9 ft (2.74 m) at the abutments. The structure has a symmetrical vertical curve of 800 ft (243.8 m) radius from abutment to abutment with the approach roadways at a 6% grade.15 Superstructure cross section, Figure 1.30, is a single-cell box 24 ft (7.3 m) in width with the top flange cantilevering 3 ft 9$ in. (1.16 m) for a total top flange width of 31 ft 7 in. (9.63 m), providing two traffic lanes and a pedestrian path. The webs have a constant thickness of 14 in. (0.36 m). Bottom flange thickness varies from 7 in. (0.18 m) at midspan of the center span to 46 in. (1.17 m) at the

FIGURE 2.68. Koror-Babelthuap Bridge, parabolic soffit of main span (courtesy of Dr. Man-Chung Tang, DRC Consultants, Inc.).

main pier and then to 21 in. (0.53 m) at an intermediate diaphragm located in the end span. This diaphragm and the one at the end pier form a ballast compartment. Another ballast compartment is located between the end-pier diaphragm and the abutment. The bottom flange under the ballast compartments is 3 ft (0.9 m) thick in order to support the additional load of ballast material. Top flange thickness varies from 11 in. (0.28 m) at midspan of the main span to 17 in. (0.43 m) at the main pier and has a constant thickness of 17 in. (0.43 m) in the end spans.15 The superstructure is monolithic with the main piers, with a permanent hinge at midspan to accommodate concrete shrinkage, creep, and thermal movements. The hinge can only transfer vertical and lateral shear forces between the two cantilevers and has no moment-transfer capacity.15 The superstructure was constructed in segments with the end spans on falsework and the main span in the conventional segmental cantilever manner, were comusing form travelers. After f-oundations pleted, a 46 ft (14 m) deep by 37 ft (11.3 m) pier segment was constructed, Figure 2.69, in three operations: first the bottom flange, then the webs and diaphragm, and finally the top flange. Upon completion of the pier segment, form travelers were installed and segmental construction begun. Two form travelers were used to simultaneously ad-

FIGURE 2.69. E;oror-Baheltlluap Bridge, pier segment (courtesy of Dyckerhoff & Widmann).

Vejle Fjord Bridge, Denmark

FIGURE 2.70. ~oror-K;tt,clthrl;lp

Bridge. main-span

cantilevers advancing (courtesy of Dyckerhoff PC Widmann). Vance the main-span cantilevers, Figure 2.70. Segments for this project were 15 ft (4.57 m) in length. l5 On this project, each segment took slightly more than one week to construct. A typical cycle was as follows : I5 1.

2.

3.

4.

5. 6. 7.

When the concrete strength in the last segment cast reached 2500 psi (17.2 MPa), a specified number of tendons, ranging from six to 12, were stressed to 50 percent of their final force, thus enabling the form traveler to advance in preparation for the following segment. Advancing the form traveler also brought forward the outside forms of the box. The forms were cleaned while rough adjustments of elevation were made. Reinforcement and prestressing tendons were placed in the bottom flange and webs. The inside forms were advanced and top flange reinforcement and tendons placed. After the previous segment concrete had reached a strength of 3500 psi (24.1 MPa), the remaining tendons were stressed. The previous segment had to be fully prestressed before concrete for the subsequent segment could be placed. Fine adjustment of the forms for camber and any required correction was made. New segment concrete was placed and cured. When the new segment reached a concrete strength of 2500 psi (17.2 MPa), the cycle was repeated.

The structure was prestressed longitudinally, transversely, and vertically. Three hundred and

two longitudinal tendons were required at the pier segment. As the cantilever progressed, 12 to 16 tendons were anchored off at each segment, with eight longitudinal tendons remaining for the last segment in a cantilever at midspan. As the structure has a hinge at midspan, there were no continuity tendons in the bottom flange. Transverse tendons in the top flange were spaced at 22 in. (0.56 m) centers. Vertical tendons were used in the webs to accommodate shear. Spacing for the vertical web tendons was 30 in. (0.76 m) in the center span and 15 in. (0.38 m) in the end spans. All tendons were la in. (32 mm) diameter barsI Side spans were constructed on falsework resting on compacted fill. The sequence of segmental construction in the side spans was coordinated with that in the main span, so that the unbalanced moment at the main pier was maintained within prescribed limits. 2.13


This structure crosses the Vejle Fjord about 0.6 mile (1 km) east of the Vejle Harbor. It is part of the East Jutland Motorway, which will provide a bypass around the city of Vejle, Denmark. A total length of 5611 ft (17 10 m) makes it the second longest bridge in Denmark. Bid documents indicated two alternative designs, one in steel and one in concrete. The steel alternative called for a superstructure composed of a central box girder with cantilevered outriggers supporting an orthotropic deck and fjord spans of 413 ft (126 m). The second alternative required a prestressed concrete superstructure with a central box girder to be constructed by the balanced cantilever method utilizing either precast or cast-inplace segments, with fjord spans of 361 ft (110 m). The successful alternative was the cast-in-place segmental prestressed concrete box girder. The bridge, in plan, is straight without any horizontal curvature. It does have a constant grade of 0.5% falling toward the north. Navigation requirements were a minimum 131 ft (40 m) vertical and 246 ft (75 m) horizontal clearance. Water depth in the fjord is generally 8 to 11.5 ft (2.5 to 3.5 m) except at the navigation channel, where the depth increases to 23 ft (7 m). Under the fjord bed are layers of very soft foundation materials, varying in depth from 26 to 39 ft (8 to 12 m). Therefore, the piers in the fjord are founded on 8 in. (0.2 m) square driven reinforced concrete piles varying in length from 100 to 130 ft (30 to 40 m), Figure 2.71. Piers on the south bank are founded on

64


m CAST I FlXCD rcmuwan USTw4nEw.Y _---_------_..,---

R E D PDII I L ErsS 0 150 CM

as1 MICACEOUS

,‘,“‘,‘.‘,‘,‘,‘,

FIGURE 2.71. Vejle Fjord Bridge, fjord piers

FIGURE 2.72.

founded on driven reinforced concrete piles.

on bored piles.

bored reinforced concrete piles, 59 in. (1500 mm) in diameter, Figure 2.72. On the north bank one pier is founded on driven reinforced concrete piles and one is supported directly on a spread footing. The cross section of the bridge, Figure 2.73, which carries four traffic lanes with a median barrier, is a variable-depth single box with a vertical web and prestressed transverse ribs. Total width between edge guard rails is 87 ft (26.6 m). Box girder width is 39.4 ft (12 m), with a depth varying from 19.7 ft (6 m) at the pier to 9.8 ft (3 m) at midspan. Each segment is cast with a length of 11.3 ft (3.44 m). Transverse top flange ribs are spaced at 22.6 ft (6.88 m) centers-that is, every other segment joint. The total bridge length is divided into four separate sections by three expansion joints located at the center of spans 4-5, 8-9, and 12-13. Longitudinal prestress is achieved by Dywidag (twelve

Vejle Bridge, land .piers founded I Fjord _

0.6 in. diameter strand) tendons, as are the transverse prestress in the top slab and the continuity prestress in the bottom slab. A 492 ft (150 m) long steel launching girder and two special form travelers were used for casting in place the full width of the 11.3 ft (3.4 m) long segments in balanced cantilever. Insulating forms followed the form travelers in order to prevent the formation of fissures due to adverse temperature gradients. In addition, the steel girder stabilized the concrete structure during construction and was used for the transportation of materials, equipment, and working crew. The total weight of the girder including the two travelers was approximately 660 t (600 mt). A typical longitudinal section of a cantilever is shown in Figure 2.74, along with the structure erection procedure. Work on the bridge started in the summer of 1975 and was scheduled for completion in 1980.

C R O S S S E C T I O N 1:200 2660

l

50

300

50

750

5010030

AT MID SPAN

FIGURE

2.73.

- - I - -

OVER

PIER

Vejle Fjord Bridge, elevation, plan, and cross section.

BOX-TYPE GIRDER z

LONGITUDINAL

POSITION

SECTION

OF

PRESTRESSING

TENDONS

I I SUPERSTRUCTURE, PRINCIPLE OF EXECUTION

FIGURE

2.74.

AUXILIARY EQUIPMENT ETC.

CONSTRUCTION

Vejle Fjord Bridge, longitudinal section and erection sequence.

PRINCIPLES


67 :: \\\ \\ \ \\\\

\\

FIGURE 2.75. Vejle Fjord Bridge, launching girder.

,:< \ \ \\

>~‘

.:,

FIGURE 2.77. Vejle Fjord Bridge, pier segment with diaphragm.

FIGURE 2.78. Vejle Fjord Bridge, construction \iew, spring 1978 (courtesy of H. A. Lindberg).

\

;->, : \

I_\\

1

FIGURE 2.76. Vejle Fjord Bridge, transverse ribs. Construction progress in the spring of 1978 is illustrated in Figures 2.75 through 2.78. Figure 2.79 is an aerial view showing the structure nearing completion. To keep within the construction schedule, it was finally necessary to use two complete sets of launching girders and twin travelers working simultaneously from both ends of the bridge.

2.79. Vqjle Fjord Bridge. aerial view from the northwest. FIGURE


68 2.14

Houston Ship Channel Bridge, U.S.A.

This bridge, a rendering of which is shown in Figure 1.67, includes a main structure over the Ship Channel in Houston, Texas, and two approach viaducts. The main structure is a three-span continuous box girder, cast in place in balanced cantilever. Span lengths are 375, 750, and 375 ft (114, 229, and 114 m). The navigation channel is 700 ft (213 m) wide at elevation 95 ft (29 m) and 500 f-t (752 m) wide at elevation 175 ft (53.4 m), Figure 2.80. The three-web box girder carries four traffic lanes separated by a 2 ft 3 in. (0.7 m) central barrier and has two 3 ft 9 in. (1.14 m) parapets. The box girder is fixed to the top of the main piers to make the structure a three-span rigid frame. Support for the box girder is provided by elastomeric bearings on top of the transition piers, where it is separated from the approach viaducts by expansion joints.

shrinkage, superimposed dead loads, and live loads). They are, therefore, heavilv reinforced; their dimensions are: Total height (from top of footing to bottom of pier segments): 160 ft 10 in. (49 m) Length (parallel to centerline of highway): 20 ft constant (6.1 m) Width: variable from 38 ft at the bottom to 27 ft 7 in. at the top (11.6 to 8.4 m) Pier cross section: rectangular box, with 2 ft (0.6 m) constant wall thickness

The transition piers support the last segment of the main structure side span and the last span of the approaches. The pier shaft is a rectangular box with 1 ft 4 in. (0.4 m) thick walls. Their heights are 152 ft (46 m) at one end and 164 ft (50 m) at the other end of- the bridge. The length, parallel to the centerline of the highway, varies from 18 to 8 ft (5.5 to 2.4 m); the width is 38 ft (11.6 m) constant. Atop the pier, a 6 ft 8 in. (2 m) cap carries the perFoundations The two center piers and two tranmanent elastomeric bearings and all the temporary sition piers rest on 24 in. (610 mm) diameter jacks and concrete blocks that will be used at the driven steel pipe piles. The center piers each rest time of the side-span closure pour. All four piers upon 255 piles with a unit pile capacity of 140 t are slip-formed. (127 mt). Footings are 81 ft (24.7 m) wide, 85 ft (26 Box Gzrder Superstructure Dimensions of the m) long, and 15 ft (4.6 m) deep. These footings are variable-depth box girder were dictated by verv surrounded by a sheet pile cofferdam and are stringent geometry requirements. Vertical alignpoured on a 4 ft (1.2 m) thick subfooting seal conment of the roadway was determined by the crete. The transition pier footings are 50 ft (15.2 maximum allowable grade of the approach viam) wide, 35 ft (10.7 m) long, and 5.5 ft (1.7 m) ducts and the connection thereof with the roadway thick and rest on 70 piles each of 100 t (90 mt) system on both banks. The clearance required fat bearing capacity. the ship channel left, therefore, only a structural depth of 2 1.8 ft (6.6 m) at the two points located Piers The main piers provide for the stability of 250 ft (76 m) on either side of the midspan section. the cantilevers during construction (unbalanced The soffit is given a third-degree parabolic shape construction loads and wind loads) and participate to increase the structural depth near the piers in in the capacity and behavior of the structure under order to compensate for the very lirnited height of service loads (long-term loads due to creep and

FIGURE 2.80.

Houston

Ship

Channel

Bridge,

longitudinal

section.

Houston Ship Channel Bridge, USA

the center portion of the main span. Maximum depth at the pier is 47.8 ft (14.6 m), with a spanto-depth ratio of 15.3. Minimum depth at midspan is 15 ft (4.6 m), with a span-to-depth ratio of 49. Over the 500 ft (152 m) center portion of the main span the span-to-depth ratio is 23, compared to a usual value between 17 and 20. Typical dimensions of the box section are shown in Figure 2.8 1. Posttensioning is applied to the box section in three dimensions:

69

Longitudinal prestress is provided by straightstrand tendons (twelve 0.6 in. diameter or nineteen 0.6 in. diameter strands), as shown schematically in Figure 2.82. Transversely, the top slab is post-tensioned by tendons (four 0.6 in. diameter strands) in flat ducts placed at 2 ft (0.6 m) centers. Vertically, the three webs are also post-tensioned as prescribed in the specifications to a minimum

t

k Bridge

FIGURE 2.81.

Houston Ship Channel Bridge, box section.

T r a n s v e r s e t e n d o n s 4xO.G;

Cantilever

FIGURE

prestress

over

main

piers

/

/

Tendons

( 1 2 x 06%a..ond (19xO.6’dia. I

Continuity

prestress

at

2.82.

Houston Ship Channel Bridge, longitudinal prestress.

mid

-span

70


FIGURE

2.83.

Houston Ship Channel Bridge, details of travelers

compressive stress equal to 3Ji; that is, 230 psi (1.6 MPa) for a concrete strength J‘i = 6000 psi (41.4 MPa). Details of the form traveler are shown in Figure 2.83. Pier segments over the main piers are of unusual size and posed a very interesting design problem, arising from the transfer of the superstructure un-

balanced moments into the pier shafts. Additional vertical post-tensioning tendons are provided in the two 2 ft (0.6 m) thick pier diaphragms for this purpose. End segments over the transition piers were designed to allow either the approaches or the main structure to be completed first, as these are two separate contracts. It is possible to make an adjustment at the end piers to compensate either for differential settle-

71

Other Notable Structures

(a)

2.15.1 MEDWAY

BRIDGE, U.K.

One of the first very long-span cantilever bridges was the Medway Bridge. This structure used a series of temporary falsework bents to provide stability during construction, Figure 2.84. 2.15.2 RIO TOCANTINS

BRIDGE, BRAZIL

This structure has a center span of 460 ft (140 m) and two side spans of only 174 ft (53 m), Figures 2.85 and 2.86. 2.15.3 PUENTE FIGURE 2.84. Xlrti~av Bridge, U.K. ((I) I‘)pical struction sequence. (h) View of’ finished bridge.

COII-

ments or for any deviation of the deflections from the assumed camber diagram used for construction. Provisions have been made for unexpected additional concrete shrinkage and creep problems; empty ducts have been placed in the pier segment diaphragms and at midspan to allow for future possible installation of additional tendons located inside the box girder but outside the concrete section, should the need for such tendons arise.

DEL AZUFRE, SPAIN

This bridge is located very high over a deep canyon of the Rio Sil. Cantilever cast-in-place was the ideal answer to allow construction with a minimal contact with the environment, Figures 2.87 and 2.88. 2.15.4 SCHUBENACrlDIE

BRIDGE, CANADA

This three-span bridge with a center span of 700 ft (213 m) crosses the Schubenacadie River, near Truro, Nova Scotia. High tidal range, swift currents, ice, and adverse climatic conditions made the construction of this structure very challenging, Figures 2.89 and 2.90. 2.15.5

INCIENSO

BRIDGE,

GUATEMALA

2.15 Other Notable Structures

There are so many outstanding and interesting cast-in-place cantilever bridges in the world today that it is impossible to discuss the subject adequately in the space available here. Mention should be made, however, of several notable structures not yet covered by a detailed description.

The main three-span rigid frame structure with a center span of 400 ft (122 m) is of cast-in-place balanced cantilever construction, and the approach spans are of precast girders, Figures 2.91 and 2.92. The very severe 1977 earthquake left the center structure completely undamaged, while the usual damage took place in the approach spans.


72

1

FIGURE 2.85.

1.72S

j6.55 1

1.725

1

Rio Tocantins Bridge, Brazil, typical elevation and cross section.

2.15.6 SETUBAL BRIDGE, ARGENTINA

This three-span structure with a main span of460 ft (140 m) rests on two main river piers with twin vertical walls and piles, with a transition footing at water elevation, Figures 2.93 and 2.94. 2.15.7 KIPAPA STREAM BRIDGE, U.S.A.

This bridge is located in the Island of Oahu in the State of Hawaii. The dual structure has an overall

width of 118 ft (36 m) to accommodate six traffic lanes, three in each direction, and consists of two double-cell box girders of constant depth with interior spans of 2.50 ft (76.2 m), Figures 2.95 and 2.96. Construction was by cast-in-place cantilever with segments 15 ft 3 in. (4.65 m) long. The bridge has pleasant lines, which blend aestheticallv with the rugged deep-valley site. 2.15.8

PARROTS FERRY BRIDGE, U.S.A.

This structure, built in California for the Corps of Engineers, represents a major application of lightweight concrete for cast-in-place cantilever construction, Figure 2.97. 2.15.9

FIGURE 2.86. Rio Tocantins Bridge, Brazil, view of the finished bridge.

MAGNAN

VIADUCT,

FRANCE

Located just off the French Riviera in Southern France, this four-span continuous structure rests on 300 ft (92 m) high twin piers of an I-shaped section. Superstructure was cast in place in two stages (first the bottom slab and webs and then the top slab) to reduce the weight and cost of travelers. Figures 2.98 and 2.99 show the principal dimensions and views of one cantilever and the finished structure, Figure 2.100.

Other Notable Structures 6S.00

130 00

73 I

*+t

cm

I I I

Ad

*

I

FIGURE

2.87.

d

Puente de1 Azufre, Spain, typical elevation and sections.

2.15.10 PUTE4UX BRIDGE, FRANCE

These are twin bridges crossing the Seine River near Paris. Because of very stringent clearance and geometry requirements, the available structural depth was only 5.9 ft (1.8 m) for the clear 275 ft (83.8 m) span and 4.8 ft (1.47 m) for the clear 214

ft (65.3 m) span, making both structures very slender, Figures 2.101 and 2.102. Stiff “V” piers in both structures help reduce the flexibility of the deck. 2.15.11 TRICASTIN BRIDGE, FRANCE

This structure spans the Rhone River with no piers in the river, which necessitates a long center span and two very short side spans anchored at both ends against uplift. The center portion of the main span is of lightweight concrete, while the two zones over the piers where stresses are high are of conventional concrete, Figures 2.103 and 2.104. 2.15.12 ESCHACHTAL BRIDGE, GERMANY

FIGURE 2.88.

Puente &%I Azuir e, Spun.

This bridge is located near Stuttgart, Germany. The superstructure consists of a large single-cell box girder with large top flange cantilevers supported by precast struts. Because of the weight involved, the central box was cast in one operation; struts were installed and flanges cast subsequently, Figures 2.105 and 2.106.

Elevation

16'~0"

1

16’4” Q I

,6,-o”

i

16’4”

Section at Midspan

II

20’~0” Section pver

4 Piers

FIGURE 2.89. Shubenacadie Bridge, elevation and sections, from ref. 16.

FIGURE 2.90. Shubenacadie Bridge, supper t avstem for unbalanced cantilever moment at pier (courtesy of the Portland Cement Association). 74

FIGURE 2.91. Incknso Bridge, Guatemala, view of the structure.

ELEVATION

@@Gp 7 50 MAIN ‘/2

SECTION ON SUPPORT

BRIDGE

‘/2

SECTION ON SPAN

FIGURE 2.92. Incienso Bridge, Guatemala, dimensions.

FIGURE 2.93. Setubal Bridge, Argentina, dimensions. 75

FIGURE bridge.

2.94.

Setubal Bridge, Argentina, view of the

Abut 2 3

2

1

29’&”

c

4 Elevation

5

-...v.&-.-.-__.

7

6

29’4”

._~ . __. i)

~~,2,, -- - FIGURE

2.95.

Kipapa Stream Bridge, elevation and cross section.

FIGURE 2.96. Kipapa Stream Bridge, construction view (courtesy of Dyckerhoff & Widmann).

FIGURE 2.97. Ferry Bridge, ref. 17.

=2?

Parrots dimensions,

COUPE LONGITUDINALE ?? 99

0 T

00

@ UC

FIGURE 2.98. Magnan Viaduct, longitudinal section.

FIGURE 2.99.

Ilagnan Viaduct, view of a cantilever.

FIGURE 2.100. Magnan Viaduct, aerial view of the completed bridge.

FIGURE

2.101.

Puteaux Bridge, aerial view of the completed bridge.

Ill1 rlnrn - Ml

h “t/ \

10.00 5 00

5.00

c

1 2 . 4 0

++-j 2 . 4 0

FIGURE 2.103. Tricastin Bridge, dimensions.

79

FIGURE 2.104. Tricastin Bridge, view of finished bridge.

FIGURE 2.105. Eschachtal Bridge, casting flange

FIGURE 2.106. Eschachtal Bridge, view of outrigger

cantilevers.

struts.

80

81

References

2.16

Conclusion

‘I‘he I~;III\~ structures described above show the versatilitv of’ cast-in-place balanced cantilever construction, particularl\~ in the field of vet-v-long-span bridges with tew repetitive spans. The design aspect 01‘ these structures will be discussed in Chapter 4 attd construction problems in Chapter 11.

References 1. H. I‘llUl, “RlYlcLentMll,”

JAIJI-g;~t~g.

Hctt 5.

Bdot/- uuct S~nhlh/e~r//~crtc, 6 1

\I;ti 1966.

2 .

L‘lt-ich Fitistrrwaldet-. “Prestressed Concrete BI-idge <:onst~~tction.” Jounrcll of tha .4ttrwicntr COPIUP~P It~ctitlrtr. Vol. 62. So. 9, Seprember 1965.

3 .

L’lricti Finster\val(ler. “ S e w D e v e l o p m e n t s i n Prestreshing .\Iethotls a n d C o n c r e t e B r i d g e Construcrioti.” I)~711/,/1~-Br,-rchlr, 4-1967, S e p t e m b e r 1 9 6 7 , I)\ckerhof’t K- Widmann Kc;, hlunich, Germanv.

4 . L‘lt-ic-h Fiiister~valder, “Fi-ee Cantilever Construction 01 Prestressed CoIlcrete B r i d g e s a n d MushroonShaped Bridges,” I;/,.\[ I~tprrccct~or~rtl Svmpo.tiu~, Cow or/r Hr/f/gfj Dr.\rgt/, r\Cl Publication- SP-23, Papet SP23-26. American Concrete Institute. Detroit, 1969.

8 . R i c h a r d A . D o k k e n , “CAL.I‘RANS Experience in Segmental Bridge Design.” Bririp Sotu, Division of Structures, Departmenr ot .[‘I-ansportation, State of California, Vol. XVII, No. 1, March 1975. 9 . A . P . Berzone, “ P i n e V a l l e y C r e e k BridgeD e s i g n i n g f o r S e g m e n t a l C o n s t r u c t i o n , ” \leeting Prepr-int 1 9 4 4 , AXE N a t i o n a l S t r u c t u r a l E n gineering hleeting, April 9-13, 1973, San Francisco. 10.

Richard Heinen, “Pine Valley Creek Bridge: Use ot Cantilever Construction,” Meeting Preprint 198 I, ASCE Sational Strucrural Engineering Meeting, .-\pt-il 9 - 1 3 , 1 9 7 3 , Sail Francisco.

1 1. “ A . 1 5 e t A . 8 6 r a c c o r d e m e n t autoroutiel- dans le nerd du departement ties hauls-de-seine,” Ministere d e L’Equipemenr D i r e c t i o n Departemental de L’Equipement des Hauts-de-Seine, Paris, September 1976. 12. “Bridge Has 595 CoKam-tio~l

‘VfWS,

ft Post-tensioned S p a n , ” 2, 1976.

Hmy~

August

IS. “Napa R i v e r B r i d g e , Sapa, Calif~~rnia,” Portland Cement Association, Bridge Report, SR 194.01 E, 1977. 1 4 . “ A l t e r n a t e B i d d i n g f o r C a l i f o r n i a ’ s Napa R i v e t Bridge Won by Cast-in-Place Prestressed Concrete Segmental Construction,” Prestressed Concrete Institute, Post-‘I‘ensioning Division, Special Bridge Report.

.5.

“Bridge Built :Itop the I‘m\ elcrc,” Etrgitrwrttcg

With Cantilevered .\‘Pu~.\-RPCO~, June 18, 1964.

1 5 . hian-Chung ‘Tang, “Koror-Babelthuap B r i d g e - A World Record Span,” Preprint Paper 3441, ASCE Convention, Chicago, October 16-20, 1978.

6.

Dale F. I)o\vning, “Cantilever Segmental Prestressed Cast-in-Place Construction of’ the Pine Valley Creek Bridge.” presented to the X.-\SHO Annual &leering, I.0 .-\ngeles. Caliti~rnia. So\,ember 1 I- 15, 19i3.

1 6 . D . W . Macintosh and R. A. Whitman, “The Shubenacadie Bridge, .Maitland, Nova Scotia,” Annual Conference Preprints, Roads and ‘I‘ransportation Xssociation of Canada, Ottawa, 1978.

,

“Pine \‘alle\ Creek Bridge, Calit’ornia,” Bridge Report SK 16 1 .O 1 E. Portland Cement Association, Skokie. 111.. 19i4.

15. “Concrete A l t e r n a t e W i n s C o m p e t i t i v e B i d d i n g Contest f’or Long Span California Bridge,” Bridge Report, PostGensioning Institute, April 1977.

Scenery

3 Precast Balanced Cantilever Girder Bridges 3.1 3.2 3.3 3.4

INTRODUCITON CHOISY-LE-ROI BRIDGE AND OTHER STRUCTURES IN GREATER PARIS, FRANCE PIERRE BENITE BRIDGES NEAR LYON, FRANCE OTHER PRECAST SEGMENTAL BRIDGES IN PARIS 3.4.1 3.4.2 3.4.3 3.4.4

3.14 B-3 SOUTH VIADUCTS, FRANCE 3.15 ALPINE MOTORWAY STRUCTURES, FRANCE 3.16 BRIDGE OVER THE EASTERN SCHELDT, HOLLAND 3.17 CAPTAIN COOK BRIDGE, AUSTRALIA 3.18 OTHER NOTABLE STRUCTURES

Paris Belt (Downstream) Paris Belt (Upstream) Juvisy Bridge Twin Bridges at GmfIans

3.18.1 Calix Bridge, France 3.18.2 Vail Pass Bridges, U.S.A. Tent Viaduct, U.K. 3.18.3 3.18.4 L32 Tauernautobahn Bridge, Austria 3.18.5 Kishwaukee River Bridge, U.S.A. 3.18.6 Kentucky River Bridge, U.S.A. 3.18.7 I-205 Columbia River Bridge, U.S.A. 3.18.8 Zilwaukee Bridge, U.S.A. Ottmarsheim Bridge, France 3.18.9 3.18.10 Overstreet Bridge, Florida, U.S.A. 3.18.11 F-9 Freeway, Melbourne, Australia

3.5 OLERON VIADUCT, FRANCE 3.6 CHILLON VIADUCT, SWITZERLAND 3.7 HARTEL BRIDGE, HOLLAND 3.8 RIQNITEROI BRIDGE, BRAZIL 3.9 BEAR RIVER BRIDGE, CANADA 3.10 JFK MEMORIAL CAUSEWAY, U.S.A. 3.11 SAINT ANDRk DE CUBZAC BRIDGES, FRANCE 3.12 SAINT CLOUD BRIDGE, FRANCE 3.13 SALLINGSUND BRIDGE, DENMARK

3.1

Zntroduction

As indicated in Chapter 1, precast segmental construction had its origins (in the contemporark sense) in France in 1962 as a logical alternative to the cast-in-place’ method of construction. To the advantage of segmental cantilever construction, primarily the elimination of conventional falsework, the technique adds the refinements implicit in the use of precasting. The characteristics of precast segmental construction are: 1. Fabrication of the segments can be accomplished while the substructure is under construction, thus enhancing erection speed of the superstructure. 2. BY virtue of precasting and therefore maturity. of the concrete at the time of erection, the time required for strength gain of the concrete is removed from the construction critical path. 82

REFERENCES

3.

As a result of the maturity of the concrete at the time of erection, the effects of concrete shrinkage and creep are minimized.

4. Superior quality control can be achieved for factory-produced precast concrete. However, geometric control during fabrication of segments is essential, and corrections during erection are more difficult than for cast-in-place segmental construction. In addition, the connection of longitudinal ducts for post-tensioning tendons and the continuity of reinforcing steel, if they are required in the design, are less easily achieved in precast than in cast-in-place methods. Although precast segmental had been used as early as 1944 for the Luzancy Bridge over the Xlarne River, Figure 1.27, wide acceptance began lvhen match-casting techniques were developed. Basically, the principle of fabrication of precast segments is to cast them in a series one against the other in the order in which they are to be assem-

Choisy-le-Roi Bridge and Other Structures in Greater Paris, France bled in the structure. The front face of a segment, thus, serves as a bulkhead for casting the rear face of the subsequent segment. Methods of fabrication of precast segments will be discussed in Chapter 11. Segments are erected in balanced cantilever starting from a segment over the pier, which is the first to be placed. Modifications to the initial principle hau e further inc-rea>& the %,ex;lbcl(clty of eye
83

The first precast segmental bridge to be built on the North American Continent was the Lievre River Bridge on Highway 35,s miles (13 km) north of Notre Dame du Laus, Quebec, with a center span of 260 ft (79 m) and end spans of 130 ft (40 m), built in 1967. It was followed in 1972 by the Bear River Bridge, Digby, Nova Scotia (Section 3.9), with six interior spans of 265 ft (81 m) and end spa-m of ‘Lo4 ft (65i -i-ix\. The 3FU KcnQxia( Causeway, Corpus Christi, Texas (Section 3.10), opened to traffic in 1973, was the first precast segmental bridge to be constructed in the United States. In the United States, as of this writing, the authors are aware of more than 30 precast segmental bridge projects that are either completed, under construction, or in the design stage. Some are listed in Table 3.1 .3

3.2

Choisy-le-Roi Bridge and Other Structures in Greater Paris, France

The first bridge to use the precast segmental cantilever technique with epoxied match-cast joints was built at Choisy-le-Roi near Paris between 1962 and 1964. It carries National Highway 186, a part of the Paris Great Belt system, over the Seine River just east of Orly Airport, Figure 3.1. This structure is a three-span continuous bridge of constant depth with end spans of 123 ft (37.5 m) and a center span of 180 ft (55 m), Figures 3.2 and 3.3. This bridge replaced one constructed in 1870, which had a superstructure of six steel girders with five spans of approximately 75 ft (23 m). This structure, determined to be no longer adequate as early as 1939, was severely damaged during World War II. It in turn had replaced an ancient bridge of five 66 ft (20 m) oak arch spans designed by the famous mathematician Claude-Louis-Marie Navier.4 In 1961, a study by the Administration of Bridges and Roads allowed two options, one in prestressed concrete and the other in steel, each having three continuous spans of 123 ft (37.5 m), 180.4 ft (55 m), and 123 ft (37.5 m). Four prestressed concrete solutions were considered. The successful solution is illustrated in Figure 3.2. The overall width of the superstructure for this dual bridge is 93.2 ft (28.4 m), Figure 3.3. Each bridge consists of two single-cell rectangular box girders. The superstructure accommodates dual two-lane roadways of 23 ft (7 m), two 13 ft (4 m) sidewalks, and a 10 ft (3 m) median.4*5 Individual box girders have a constant depth of 8.2 ft (2.5 m),

Precast Balanced Cantilever Girder Bridges

84 TABLE

3.1.

Name and Location Lievre River, Notre Dame du Laus, Quebec Bear River, Digby, Nova Scotia JFK Memorial Causeway, Corpus Christi, Texas Muscatuck River, U.S. 50, North Vernon, Indiana Sugar Creek, State Route 1620, Parke County, Indiana Vail Pass, I-70 West of Denver, Colorado (4 bridges)

Penn DOT Test Track Bridge, Penn Sate University, State College, Pa. Turkey Run State Park Parke County, Indiana Pasco-Kennewick, Columbia River between Pasco and Kennewick, Washington (cable-stay spans) Wabash River, U.S. 136, Covington, Ind. Kishwaukee River, Winnebago CO . near Rockford, Ill. (dual structure) Islington Ave. Ext., Toronto, Ontario Kentucky River, Frankfort, Ky. (dual structure) Long Key, Florida (contract let late 1978) Linn Cove, Blue Ridge Parkway, N.C. (contract let late 1978) Zilwaukee, Michigan (dual structure) (bids opened late 1978)

Precast Segmental Concrete Bridges in North America

Date of Construction

Method of Construction”

Span Lengths, tt (m)

1967

B.C.

1972

B.C.

1973

B.C.

1975

B.C.

1976

B.C.

1977

B.C.

1977

O.F.

130-260- 130 (39.6 - 79.2- 39.6) 203.75 - 6 (12 265 - 203.75 (62.1 - 6 ((I 80.77 - 62.1) loo-200- 100 (30.5 - 6 t - 30.5) 95 190-95 (29 - 58 - 29) 90.5 - 180.5 - 90.5 (27.6 - 53 - 27.6) 134 - 200 - 200 - 134 (40.8 - 61 - 61 - 40.8) 134-200-200145 (40.X-61 -61 -44) 151-155-210-210-154 (46-47.2-64-64-47) 153-210-210154 (46.6 - 64 - 64 - 47) 124 (37.8)

1977

B.C.

1978

B.C.

1978

1.L.

1979

B.C.

1979

B.C.

1979

B.C. S.S. P.P.

B.C.

93.3 - 4 (a 187 - 93.5 (28.5 - 3 (@ .57 - 28.5) 170-3 @I 250- 170 (51.8 - 3 G 76.2 - 51.8) 2 @ 161 -200-5 @ 272 (2 @ 49 - 61 - 5 (if X3) 228.5 - 320 - 228.5 (69.6 - 97.5 - 69.6) II3 - 101 fin 118 - 113 (34.4 - 101 @i 36 - 34.4) 9X.5- 163-4@ 1X0- 163-98.5 (30 - 49.7 - 4 Q 54.9 - 49.7 - 30) 26 N.B. spans total length 8.087.5 (2,465) 25 S.B. spans total length 8.0575 (2,456) maximum span 392 (119.5)

“Method-of-construction notation: B.C.-balanced cantilever, l.L.-incremental placement, S.S.-span-by-span.

top flange width of 21.65 ft (6.6 m), and a bottom flange width of 12 ft (3.66 m). Webs have a constant thickness of la in. (0.26 m), and the top flange is of constant section throughout its length with a minimum thickness of 7 in. (0.18 m) at its

180 - 1x0 (54.9 - 54.9) 406.5 - 98 1 - 406.5 (124 - 299 - 124)

launching, O.F.-on talsework. P.P.-Progressive

crown, Figure 3.3. The bottom flange thickness is 6 in. (0.15 m), except near the river piers where the thickness increases to 15.75 in. (0.4 m) to accommodate cantilever b e n d i n g s t r e s s e s . T h e downstream half of the bridge (consisting of two

85


Precast Segmental Bridges Choisy-le-Roi 1962-64 Courbevoie 65-66 Ring .Motorlva\ 66-68 Ring .Motor\vav 6i-68 St Cloud 72-74 Juvis) 66-68 Conflans 50-72 78 St Maurice Interchange B-3 South L’iaduct 71-72 Marne la Vallee 7.s77 Torcv RR 78 Clichv RR 78

Cast-in-Place Segmental

Bt-idges

1974-76 13 Gennevilliers 14 North \Vest A-86 Intel-change 78 15 Clichy High\va\ i3-i
Location map of’ segmental bridges in greater Paris, France.

box girders) ~\‘as constructed first, alongside the esisting bridge. After removal of the existing bridge. the second or upstream half was constructed. Each dual structure was constructed b\ the balanced cantilever method utilizing Frevssinet tendons f’or the longitudinal prestressing. Box girder segments \vere 8.2 ft (2.5 m) in length and lveighed 22 tons (20 nit), except the pier segments

FIGURE 3.2. Choisv-lc-Roi

Bridge.

which were 16.4 ft (5 m) in length and weighed 60.6 tons (55 mt). The pier segments also contained two diaphragms which provided continuitv with the inclined wall piers, Figure 3.3. The segments were fabricated in a precasting vard on the left bank of the Seine approximately a mile (1.6 km) upstream of the project site, Figure 3.4. Although this bridge might be considered of moderate importance with respect to span lengths, its importance lies in the method of fabrication. It was the first to use segments precast by the matchcasting technique. Segments were cast in the precasting yard as a series of 8.2 ft (2.5 m) long units, one against the other, on a continuous soffit form which had been carefully adjusted to the intrados profile of the bridge with allowance for camber. This came to be known as the “long-line” method (see Chapter 11). Two sets of steel forms riding the soffit form and overnight steam curing allowed the production of two segments per working day. To prevent bonding of the segments to each other in the casting form, a special peel-off bond breaker was sprayed over the end of the segment before the adjacent segment was cast. The segments were

86


Elevation

Elevation and cross section of river piers

,

-..I mr-&aL&. ,66

3M

I--1-e-. ,x130

20‘ohp--- A Cross section of superstructure

FIGURE 3.3. Choisy-le-Roi Bridge, dimensions: elevation, elevation and cross section of River piers, cross section of superstructure.

subsequently stripped from the soffit form at their match-castjoints and reassembled at the bridge site in balanced cantilever on each side of the river Diers.4 A floating crane handled the segments at the casting yard. After the units were loaded on barges and transported to the project site, the same crane placed the segments over a retractable jig rolling inside the box girder in the completed portion of the bridge and was thus freed for another segment placing operation. A platform mounted on jacks on the jig, Figure 3.5, allowed for adjustment of the segment at the desired position.4 A 1 ft (0.3 m) wide gap was temporarily maintained between the faces of the segments to allow workmen to apply


FIGURE

P------

3.4.

a7

Choisy-Iv-Koi Kritlge. view of’ the precasting yard.

-J FIGURE 3.5.

Choisy-le-Roi

the epoxy joint material. The jig was then retracted and prestressing tendons were placed and stressed to connect the two symmetrical segments on each side of the previously completed portion of the cantilevers on either side of the pier.5 Placing of the precast segments in a cantilever fashion on each side of the pier progressed step by step, as indicated in Figure 3.6. Tendon layout is illustrated in Figure 3.7. Upon completion of the two twin cantilevers from the river piers, a cast-inplace closure pour was consummated at midspan and a second series of prestressing tendons were placed in the bottom flange to achieve continuity between the two center-span cantilevers. These tendons were given a draped profile to allow the location of tendon anchorages in the top flange of the box girder. Both series of tendons, cantilever and continuity, overlap each other and contribute

Bridge,

retractable

erection

jig.

FIGURE 3.6. Choisy-le-Rot with floating crane.

Bridge, segment placing

Precast Balanced Cantiher Girder Bridges

l3cdes1208 ) / 8cablesl2# 7

FIGURE

3.7. Choisv-le-Roi Bridge, tendon lavout

to a substantial reduction in the shear forces in the webs as a result of the vertical component of the prestress. The side spans were constructed in a similar manner. The three precast segments adjacent to the abutments were assembled on falsework. After a closure pour between these segments and the cantilever from the river pier, positive-moment tendons were placed and stressed in the end span to achieve continuity. Because the midspan area of the center span had little capacity to withstand moment reversal under ultimate load,

FIGURE 3.8.

additional short tendons were located in the top flange to achieve full reinforcement continuity with the longest cantilever tend0ns.j The same construction technique used for the Choisy-le-Roi Bridge was used for the Courbevoie Bridge, built between 1965 and 1967, which also crosses the Seine in the northwest suburb of Paris, Figure 3.1. The bridge has three symmetrical spans of 130,200, and 130 ft (40,60, and 40 m) for a total length of 460 ft (140 m), Figure 3.8. Four box girders of constant depth carry the 115 ft (35

Courbevoie Bridge, elevation.

Pierre Benite Bridges near Lyon, France

m) wide deck, Figure 3.9. The available depth of only 7.5 ft (2.28 m) made necessary a very slender structure; depth-to-span ratio for the main span is 1 /26.5,6 Each river pier is an assembly of two half-piers, Figures 3.9 and 3.10, which are fixed at the level of the foundation. Each half-pier consists of a rectangular shaft 9 by 26 ft (2.8 by 8 m), which supports two pairs of prestressed concrete walls, above the normal water level, in the form of a parallelogram of 18 in. (0.45 m) thickness and 10.5 ft (3.2 m) width. The walls are arranged in a “V” in the transverse direction of the bridge and have a dimension of 6.7 ft (2.05 m) out-to-out of walls in the longitudinal direction.6 The girders are fixed at the piers and supported on elastomeric bearings at the abutments. A total of 148 precast segments of 12.5 ft (3.8 m) length were required for the superstructure. They were fabricated in four months at the rate of two segments per day, in two sets of steel forms, electrically heated and insulated with polyurethane 1ining.j Erection at the site was accomplished by a floating crane. After careful adjustment of the pier segments, they were erected at the rate of four per day. The temporary jig used at Choisy-le-Roi for adjustment of the segments was replaced in this project by two temporary steel beams bolted to the top of each segment and connected to the completed section of the cantilever by prestressing bars.j The girder was prestressed longitudinally and transversely, through three longitudinal cast-inplace strips between the top flange cantilevers of the box girders. The completed structure is shown in Figure 3.10.

3.3

89

FIGURE 3.9. Courbevoie Bridge, cross section at rive1 pier and abutment.

FIGURE

3.10.

Courbevoie Bridge, view of completed

3.11.

Pierre Benite Bridge, view of the

bridge.

Piewe Benite Bridges Near Lyon, France

These two large bridges carry the motorway from Paris to the Riviera south of Lyon near the Pierre Benite hydroelectric plant, Figure 3.11. There are two separate bridges, one over the draft channel of the power plant and the other over the Rhone River. Both structures are twin bridges, each bridge consisting of two single-cell box girders. Typical dimensions in longitudinal and cross sections are shown in Figures 3.12 and 3.13. The same constant depth of 11.8 ft (3.6 m) is used for all spans of the two bridges. However, a haunch under the intrados of the box girders increases the

FIGURE

finished bridge.


90 1

5600

I

I9400

I

I

84@J

!

* 7wotslr Ibxo 56m -4

I

I m 0 4

0

ISJOO

Bridge over draft channel (a) 259,OO

/

7500

FIGURE 3.12. Pierre Benite Bridge, longitudinal sections. (a) Bridge over draft channel. (b) Bridge over Rhone River.

16.92 - ..3.26-- ( i t

13.00

FIGURE 3.13.

..?.O

16

30

166

Pierre Benite Bridge, typical cross section.

structural depth over the piers to a maximum of 14 ft (4.28 m) for the 276 ft (84 m) span. All piers rest on compressed-air caissons and are made of solid cylindrical columns 6.5 ft (2 m) in diameter which support the cast-in-place pier segment, including skew diaphragms between the two individual box girders of each bridge. This pier segment served as the starting base for precast segment placing in balanced cantilever for the superstructure. The 528 segments were precast near the southern bank of the draft channel. This application of precast segmental construction was the occasion to conceive and develop for the first time the short-

line precasting method, whereby the segments are cast in a formwork located in a stationary position. Each segment is cast between a fixed bulkhead and the preceding segment, in order to obtain a perfect match. After a learning curve of a few weeks, each of the two short-line-method casting machines was used to cast one segment every day. Details and specific problems of the short-line method will be described in Chapter 11. Figure 3.14 shows the precast segments as they were fabricated, temporarily stored, loaded on barges by a very simple portal structure equipped with winches, and finally transported to the construction site.

Other Precast Segmental Bridges in Paris

91

construction site with segment placing in progress is shown in Figure 3.16. Both precasting and placing operations were carried out successfully. All the segments were placed in the structures in 13 months. The only regret was that this erection system did not provide for precast pier segments. The geometry of the cast-in-place pier segments was further complicated by the skew of the bridges, such that the contractor expended as much labor on this aspect of construction as in precasting and positioning all the precast segments.. 3.4

Other Precast Segmental Bridges in Paris

The first two match-cast bridges, Choisy-le-Roi and Courbevoie, were followed by a series of other crossings over the Seine River. All contracts for design and construction were obtained on a competitive basis with other types of materials or construction methods: The next two structures were for the construction of the Paris Belt Motorway which crosses the Seine at two locations, one downstream of the city and one upstream; see the location map, Figure 3.1. They were followed by several others, which are briefly described in this section. 3.4.1 PARIS BELT (DOWNSTREAM)

These twin bridges, Figure 3.17, carry four traffic FIGURE 3.14. Pierre Henitc Bridge, precasting yard and loading portal. (a) Precasting yard. (6) Loading portal. lanes. Dimensions are shown in Figures 3.18 and

Placing of all segments in the two twin structures was achieved in balanced cantilever, using the cast-in-place pier segments as a starting base. This project used the newly developed “beam-andwinch” erection system, illustrated in Figure 3.15 together with a close-up view of a typical segment-placing operation. Electric winches are supported in a cantilever position from the completed part of the deck to allow each segment to be lifted off the barge and placed in its final position. Because of high-velocity river currents on one structure, it was considered advisable to transfer the segments from the barge to the winch system close to the piers to allow temporary anchorage of the barge. Therefore, segments had to be moved longitudinally from the barge position to thtir final location. A special trolley carried the winches and the suspended segment while riding along rails fixed to the finished deck. A general view of the

3.19. Maximum span length is 302 ft (92 m) and the structural depth of the four box girders is 11 ft (3.4 m), increased toward the piers to a maximum of 21.3 ft (5.5 m) by straight haunches. Because of the skew between the axis of the bridge and the flow of the Seine, the pier shafts were given a special lozenge shape, which proved very efficient for the hydraulic flow and is of pleasant appearance. The limited bending capacity of the shafts called for temporary supports during cantilever construction operations. Precast segments were manufactured on the bank of the Seine with two casting machines (short-line method). For the part of the bridge superstructure located over the river, segments were placed with a floating crane, Figure 3.20. In fact, almost half the bridge length was placed over land out of reach of the floating crane. The beamand-winch equipment used at Pierre Benite Bridge was substituted for the crane to place these segments. There was also need of additional falsework on one bank to compensate for the unusually long

FIGURE 3.16.

FIGURE 3.15. Pierre Benite Bridge, segment placing scheme (left and top right). 92

Pierre Benite Bridge, under-

FIGURE 3.17. Paris Belt (Downatrearn), finished bridge.

construction.

\ itw o f

9950

FIGURE 3.18.

-------rng.3

._._ ~i6_~-~~j~~--rlps--~ ___--.-

Paris Belt (Downstream), typical longitudinal section.

FIGURE 3.19. Paris Belt (Downstream), typical cross section.

: i-


94

FIGURE 3.20. Paris Belt (Downstream), segment placing.

FIGURE 3.21. Pal i\ Belt (C‘pstlum), \iew of the finished bridge.

end span, which could not be changed because of stringent pier location requirements.

direction, Figure 3.21. The twin bridges have dimensions similar to those of the downstream bridge, and each structure has two parallel box girders connected by transverse prestress. Dimensions are shown in Figures 3.22 and 3.23. A circular intrados profile was used in lieu of the straight haunches. All segments were precast on the river bank in the immediate vicinity of the bridge, using

3.4.2 PARIS BELT (UPSTREAM)

On the other sihe of Paris another segmental structure, also carrying the Belt Motorway over the Seine, was designed for five traffic lanes in either asa

56,62

I

Id GAUCHE

-. w

FIGURE 3.22. Paris Belt (Upstream), longitudinal section. 4.50

4. 0

E

g

II 18 3.50 m

1

3.50 m

3.50 n-l

1

II

3.50 m

71

1

3.50

m

i Ill I

3.50 m

3.50 m I

3.50 m

IlII

FIGURE

3.23.

3.50 m

3.50 m

-

Paris Belt (Upstream), typical cross section.

A

/\

Other Precast Segmental Bridges in Paris S?QUl?4CCS~~ISCKceMRucnar

RUSES D’EMCUTIDN D” T-

FIGURE 3.24.

Paris Belt (Upstream), typical segment placing scheme.

the same two casting machines used previously for the downstream bridge. Placing segments in the structure posed some interesting problems, as shown in the sequence diagrams of Figure 3.24. Pier segments were too heavy to be handled as one unit and were subdivided into two segments, assembled upon the pier shaft before cantilever placing could start. A

crane, either on crawlers or on a barge, together with the beam-and-winch equipment handled all segment placing. 3.4.3 JUVISU

This bridge, Figure 3.25, is also on the Seine just south of Choisy-le-Roi; see the location map, Figure 3.1. Dimensions are shown in Figure 3.26. Segments were cast by the short-line method near the site and placed with a floating crane. An auxiliary falsework on both banks allowed segment placing and assembly beyond the reach of the floating crane. 3.4.4

FIGURE 3.25. JUVISV Bridge, completed stl ucture.

BRIDGE

TWIN BRIDGES AT CONFLANS

These twin bridges, Figure 3.27, placed about 320 ft (100 m) apart to allow for interchange ramps on both banks, are upstream of Paris where the Seine and Marne Rivers merge; see the location map, Figure 3.1. Dimensions and construction methods were similar to those of the Courbevoie Bridge already described.


96 I

24:3@

1

2413’

IO’

Cl

1

I33

Ed .-

FIGURE 3.26. Juvisy Bridge, cross section.

FIGURE 3.27. Twin Bridges at Conflans, finished

Balanced cantilever construction was accomplished utilizing a launching gantry for erection. In the approach spans the superstructure has a constant depth of 8.2 ft (2.5 m). Depth of the center spans varies from 14.9 ft (4.5 m) at the piers to 8.2 ft (2.5 m) at midspan, Figure 3.29. The rectangular box segment has a bottom flange width of 18 ft (5.5 m) and a top flange width of 34.8 ft (10.6 m). Webs have a constant thickness of 12 in. (0.3 m), while the top and bottom flanges are 8 in. (0.2 m) and 7 in. (0.18 m) thick, respectively, Figure 3.30. Typical segment length is 10.8 ft (3.3 m). Expansion of the deck is provided in every fourth span by a special stepped (ship-lap) joint with horizontal elastomeric bearing pads, Figure

bridge.

3.5 Oleron Viaduct, France The Oleron Viaduct provides a link between the mainland of France and the resort island of Oleron off the Atlantic West Coast 80 miles (128 km) north of Bordeaux. This structure has a total length between abutments of 9390 ft (2862 m). In the navigable central part of the structure are 26 spans of 260 ft (79 m), Figure 3.28. Approach spans consist of two at 194 ft (59 m), sixteen at 130 ft (39.5 m), and two at 94 ft (29 m). The superstructure is supported by 45 piers and was assembled by prestressing match-cast segments, using epoxy joints.

FIGURE 3.28. Olevon Viaduct, complcred

strllcrllre.

97

Oh-on Viaduct, France 3sllo’ 3!2’

I

j I

i t FIGURE 3.29.

, I

2916’

18'

34!9”

c

/

3!2’

I

I

I\

Oleron Viaduct, typical cross section, from ref. 5 (courtesy of

the American C:oncrete

Institute).

3.30. Throughout the total length of structure there are ten expansion joints: one at each abutment and eight intermediate ones. The latter are located at points of contraflexure in a typical interior span subjected to a continuous uniform load.” The segments with the expansion joint have the same length as typical segments and are in fact two half-segments that are temporarily preassembled with bolts, with a special layout of temporary and permanent prestressing tendons. It is then possible to maintain the balanced cantilever erection procedure beyond the expansion joint to midspan. Later on, when continuity has been achieved in the adjacent spans, the expansion.joint segment is ‘!unlocked” to perform in the intended manner.

The precasting plant was located in the vicinity of the mainland abutment. Production in this plant was scheduled so that the 24 segments required for a typical 260 ft (79 m) central span could be fabricated in nine working days. Segments were produced by the long-line method, described in Chapter 11. Four sets of steel forms rode a bench that was carefully aligned to the longitudinal profile of the roadway and the variable-depth soffit with due provision for camber. Segments were match-cast in the same relative order in which they were subsequently assembled at the site.5 An aerial view of the casting yard is shown in Figure 3.31. Handling of segments in the casting and storage yard was accomplished by a special railwaymounted gantry capable of handling loads varying

FIGURE 3.30. Oleron Viaduct, typical center span elevation, from ref. 5 (courtesy of the American Concrete Institute).


FIGURE FIGURE

3.31.

Oleron Viaduct, aerial view of casting

yard.

from 45 tons (42 mt) for the center-span segment to 80 tons (73 mt) for the pier segment. A lowboy dolly riding on rails of the finished bridge and pushed by a farm tractor transported the segments from storage to their location for assembly. Cantilever erection at the site was accomplished by a launching gantry, Figure 3.32. This gantry was the key to the successful operation of this project. Although the structure is erected over water, the use of floating equipment would have been difficult, expensive, and subject to uncertainty because of the great tidal range and the shallowness of water in most of the area traversed by the structure. Floating equipment would have been able to reach the approach piers only at high tide. During low tide the marsh area, which is the location of France’s famed Marennes oyster beds, could not accept any tire-mounted or crawlermounted equipment. Consequently, it was decided to work entirely from above with a launching gantry. This new technique was developed for the first time for this structure and was later refined for other structures. For the typical central spans the erection cycle required between eight and ten working days.5 Construction began in May 1964, three months after design work had started. The first segment was cast in July and placed in August 1964. Side spans laid on a curve were completed in December and the launching gantry was then modified for construction of the center spans. The last of the 870 precast segments was in place in March 1966, and the bridge opened to traffic in May, after an overall construction time of two years5; see the summary of the work program in Figure 3.33. A

3.32. Oleron Viaduct, construction view showing cantilever span, from ref. 5 (courtesy of the American Concrete Institute).

view of the final structure is shown in Figures 3.28 and 3.34. The Oleron Viaduct was the first application of the launching-gantry concept for placing segments in cantilever. Several structures were later designed and built with the same construction method. Mention should be made here of three special bridges: 1. Blois Bridge over the Loire River The principal dimensions are given in Figure 3.35. The superstructure box girders rest on the pier shafts through twin elastomeric bearings, which allow thermal expansion while providing partial restraint for bending-moment transfer between deck and piers. Consequently, savings are obtained both in the deck and in the foundations. All segments were placed in the bridge with an improved version of the launching gantry first designed for the Oleron Viaduct. High-strength steel and stays were used to provide minimum weight with a satisfactory stiffness during operations, Figure 3.36. High-strength bolt connections were used throughout to make the gantry completely capable of dismantling and easily transportable to other construction sites. 2. Aramon Bridge over the Rhone River This was the next structure where the same gantry could be used, Figure 3.37. 3. Seudre Viaduct Located just a few miles south of Oleron over the Seudre River, this 3300 ft (1000 m) long viaduct was also of precast segmental construction and used the same launching gan-

Chillon Viaduct, Switxerland

99

CONTINENT

OLERON

i

PIERS

ON FOOTINGS

1 1 1 1

FIGURE

PIERS ON PILES -------$

3.33.

PIERS ON FOOTINGS i

Oleron Viaduct, program of work.

try. The finished structure is shown in Figure 3.38. Foundations for the center spans were built inside sheet pile cofferdams in spite of very swift tidal currents.

3.6 Chillon

)-

Twin rectangular slip-formed shafts were used for the piers, varying in height from 10 to 150 ft (3 to 45 m). Stability during construction was excellent and required little temporary bracing except between the slender walls to prevent elastic instability.’ With the exception of three piers in each

Viaduct, Switzerland

The 7251 ft (2210 m) long dual structures of the Chillon Viaduct are part of European Highway E-2 and are located at the eastern end of Lake Geneva passing through an environmentally sensitive area and very close to the famed Castle of Chillon, Figure 3.39. In addition, the structures have very difficult geometrical constraints consisting of 3% grades, 6% superelevation, and tight-radius curves as low as 2500 ft (760 m). Each structure has 23 spans of 302 ft (92 m), 322 ft (98 m), or 341 ft (104 m). The variable spans allowed the viaduct to be fitted to the geology and topography, providing minimum impact on the scenic forest. The viaducts are divided by expansion joints into five sections of an approximate length of 1500 ft (457 m).

FIGURE bridge.

3.34.

Oleron Viaduct, aerial view of’ finished

100


@

MVATlOli Ech

61.50 t

Culie

l/ZCd

9.1 m

I

P\OO

1

R.G

-ELEVATION

I

9l.00

1

Pl

P2

P3

1

L

61,SO

PL

t

Cult*

R

D

0 COUPE TRAtlSVfRSALr Erh :

CROSS

IIlOO=

SECTION

zqoo

,l D ",

I too

I

7,oo

L

3.35.

Blois Bridge, elevation and typical cross section.

viaduct, all piers are hinged at the top. The piers that are less than 72 ft (22 m) high are hinged at the base; taller piers are fixed at their base, being sufficiently flexible to absorb longitudinal move-

FIGURE 3.36. Blois Bridge, operating on the superstructure.

$00

I

to 4.79 m at midspan

FIGURE

I

zoo

launching

ment of the superstructure. The superstructure consists of a single-cell rectangular box with a cellular cantilever top flange, Figure 3.40, and with a depth varying from 18.5 ft

gantry

FIGURE 3.37. Aramon Bridge, launching gantry.

Chillon

FIGURE

3.38.

Seudre Bridge, fItli\hcd \I

I

Viaduct, Switzerland

FIGURE 3.39. Chlllon Viaducl.

I C 1111 e.

(5.64 m) at the longer-span piers to 7.2 ft (2.2 m) at midspan. Widths of top and bottom flange are respectively 42.7 ft (13 m) and 16.4 ft (5 m). Dimensions of the tw& typical cantilevers are noted in Figure 3.4 1. Maximum segment weight was 88 tons (80 nit). A cellular cantilever top flange was used because the overall width of the top flange ex-

101

aerial LICI\.

ceeded 40 ft (approx. 12 m) and the cantilever length was 13.15 ft (4 m). An alternative would have been to provide stiffening ribs as used in the Saint Andre de Cubzac Viaducts (Section 3.11) and the Sallingsund Bridge (Section 3.13). Segments were precast in a yard at one end of the structure with five casting machines, allowing

Over supports (4

b 4

500 At mid-span (b)

FIGURE

3.40.

Chillon Viaduct, cross sections. (a) Over supports. (b) At midspan.

i

PORTIQUE - TYPE 48.OOm

PORTIOUE-TYPE

4 2 OOm

I

.,a0 **lo

73x320 EOXSOIE 42

00

i

boo

O,Q20

1 L---

CDNSOlC

42 00

82 DO

FIGURE

3.41.

Chillon Viaduct, longitudinal sections of typical cantilevers.

Hartel Bridge, Holland

103

Sections I, II, and V, conventional cast-in-place prestressed concrete box girders Sections III and IV, precast prestressed concrete segmental box girders Two steel bascule bridges.

FIGURE 3.42. Ch~llon

\.~ndu~t, precasting yard.

an average production of 22 to 24 segments per week (see aerial view, Figure 3.42). Erection was by the conventional balanced cantilever method with a launching gantry designed to accommodate the bridge-deck geometry in terms of curve and variable superelevation. The overall length of the gantry was 400 ft (122 m) and the total weight 250 tons (230 mt). Special features of this gantry will be discussed in Chapter Il. Cantilever placing of precast segments is shown in Figure 3.43. This structure is truly an achievement of modern technology with emphasis upon the aesthetic and ecological aspects of design. 3.7 Hartel Bridge, Holland The 1917 ft (584.5 m) long Hartel Bridge crosses a canal in Rotterdam, Figure 3.44, and consists of the following elements:

FIGURE 3.43. Chillon

Viatiuct, tion with launching gantry.

c a n t i l e v e r constrUC-

The original design contemplated that the total structure would be constructed as conventional cast-in-place box girders on falsework. Substitution at the contractor’s request of cast-in-place segmental construction by precast skgmental construction for sections III and IV saved the extensive temporary pile foundation system necessary to avoid uneven settlement of falsework because of initial soil conditions. The redesign proposed two single-cell rectangular box girders as opposed to one three-cell box girder, Figure 3.44, omitting the center portion of the bottom flange and providing thinner webs and a thicker bottom flange. In the segmental box girder design the dimensions of the deck slab are constant over the entire length, girder depth varies from 4.92 ft (1.5 m) to 17 ft (5.18 m), the webs have a constant thickness of 13.8 in. (0.35 m), and the bottom flange thickness varies from 10 in. (0.26 m) to 33 in. (0.85 m). Up to a depth of 9.35 ft (2.85 m) the segments have a length of 15.8 ft (4.8 m); over 9.3 ft (2.85 m) the length decreases to 12.3 ft (3.75 m). The vertical curvature of the bridge was made constant for the full length of sections III and IV by increasing the radius from 9842.5 ft (3000 m) to 19,029 ft (5800 m), which resulted in a repetition of eight times half the center span. This repetition justified precast segments. A long-line casting bed (see Chapter 11) was constructed on the centerline of the bridge box girders at ground level, Figure 3.45. Thus, a portal crane was able to transport the cast segments to the storage area and also erect them in the superstructure, Figure 3.46. The end spans have three more segments than half the center span; these were supported on temporary falsework until all the prestressing tendons were placed and stressed, Figure 3.46. The first segment cast was the pier segment; each of the remaining segments was then matchcast against the preceding segment. The pier segment was positioned on bearings on top of the pier, Figure 3.47, and the two adjoining segments were positioned (one after the other) and the joints glued with epoxy resin. Temporary high-tensile bars located on the top of the deck slab and in the bottom flange were stressed to prestress the three

Precast Balanced Cantilever Girder Bridges III

IV

Flevation

Cross sections of the redesign

Cross section of the original design

FIGURE 3.44. Hartel Bridge, typical dimensions: elevation, cross sections of the original design, cross sections of the redesign (courtesy of Brice Bender, BVNLSTS).

segments together. After the epoxy had hardened,

the permanent tendons were placed and stressed. The two segments adjoining the pier segment were supported during erection on flat jacks on the top

FIGU?E 3.43.

of the outside struts of a steel scaffolding bearing on the pier foundation. Thus, the flat jacks were used for adjustment of the segments to achieve proper geometry control. The remaining segments were

Hare1 BridgG nxrbod of castjng segments /courtesy of Brjce

Bender, BVNISTS.

Hartel Bridge, Holland

FIGURE

FIGURE 3.46. H,~~tcl H~~tigc, p t,tl (I
105

3.48.

H,II tel RI idgc. complctcd \tl II< ture.

for h‘in-

erected in the conventional balanced cantilever method. The completed structure is shown in Figure 3.48. Other structures using precast segmental construction were subsequently designed and built in the Netherlands. Shown in Figure 3.49 is the bridge over the I.jssel at Deventer, where segments in the 247 f’t (74 m) spans were placed with a launching gantry. The overall length of’ the gantry was 520 f‘t (156 m), allowing the legs to bear on the permanent concrete piers and impose no loading on the deck during construction, Figure 3.50.

FIGURE 3.49. Ikventex Bridge, placing segments with the launching gantry.

FIGURE 3.47. Hartel Bridge, erection sequence and detail of temporary pier bracing (courtesy of Brice Bender, BVN/STS).


106 I-

156 m (520 ft) 74 m (247 ft)

7 8 m (260 ft)

Real

rl

Front

Max bridge span 74 m (247 ft)

FIGURE 3.50. Deventer Bridge, elevation of gantry.

3.8 Rio-Niteroi Bridge, Brazil The Rio-Niteroi Bridge crosses the Guanabara Bay connecting the cities of Rio de Janeiro and Niteroi, thereby avoiding a detour of 37 miles (60 km). This structure also closes the gap in the new 2485 mile (4000 km) highway that interconnects north and south Brazil and links the towns and cities on the eastern seaboard, Figure 3.51. Although the route taken by the bridge across the Bay seems somewhat indirect, it was selected because it avoids very deep water and is clear of the flight path from Santos Dumont Airport.

Total project length is approximately 10.5 miles (17 km), of which about 5.65 miles (9.1 km) is over water. The alignment begins at the Rio side with a 3940 ft (1200 m) radius curve, then a straight section, within which are located steel box girder navigation spans totaling 2872 ft (848 m) in length. This is followed by an island, where the viaduct is interrupted by a road section of 604 ft (184 m), and finally another 3940 ft (1200 m) radius curve arriving at Niteroi. The precast segmental concrete viaduct sections have a total length of 27,034 ft (8240 m) representing a total deck area of 2,260,OOO sq ft (210,000

The

- Rio-Niteroi Brii

/n

FIGURE

3.51.

_.RmdeJaneir

Rio Niteroi Bridge, site location map

Rio-Niteroi Bridge, Brazil

m*), making this bridge the largest structure of its type. An aerial view of the crossing under traffic is shown in Figure 3.52. The superstructure has 262 ft (80 m) continuous spans with an expansion joint at every sixth span, Figure 3.53. It consists of two rectangular box girders for a total width of 86.6 ft (26.4 m) and a constant depth of 15.4 ft (4.7 m). A 2 ft (0.6 m) cast-in-place longitudinal closure joint

between the top flange cantilevers provides continuity between the two box girder segments. Typical segments have a length of 15.75 ft (4.8 m) and weigh up to 120 tons (110 mt). The pier segments are 9.2 ft (2.8 m) in length. Special segments are used for expansion joints. Longitudinal prestressing tendons consist of twelve f in. (13 mm) diameter strands in the top and bottom flanges with a straight profile, while the resistance to shear stresses is obtained by vertical web prestress, Figure 3.54. All segments were manufactured in a large precasting yard on a nearby island. Ten casting machines (eight for the typical segments and two for the pier and hinge segments) were laid in two independent parallel lines, each equipped with a portal crane for carrying the segments to the storage area and the loading dock. More than 3000 segments were subsequently barged to their location in the structure and erected by four launching gantries working simultaneously on each of the two parallel box girders and on either side of the bay, Figures 3.55 and 3.56. The rate of segment placing was remarkable. A typical span was assembled and completed in five working days. Between the months of February and July 1973, an average of

FIGURE 3.52. Rio-Sire] oi 131 dge, view of the completed structure.

Cross

section

Elevation fb)

FIGURE 3.53. Rio-Niteroi Bridge, cross section and elevation. (a) Cross section. (b)

Elevation.

107


ELEVATION

PLAN

CABLAGE

SUPERIEUR *

PLAN

FIGURE

3.54.

CABLAGE

INFERIEUR

Rio-Niteroi Bridge, typical span dimensions and tendon layout.

278 precast segments per month were installed in the structure by the four launching gantries, representing an area of 180,000 sq ft (17,000 m’) of finished bridge per month. At the same speed, Oleron Viaduct could have been built in two months. Such is the measure of the determination and enthusiasm of engineers and constructors of the New World.

3.9 Bear River Bridge, Canada The Bear River Bridge is about 6 miles (9.7 km) east of Digby, Nova Scotia, on trunk route 101 between Halifax and Yarmouth, near the Annapolis Basin; it replaces an 85-year-old structure. Preliminary studies showed, and construction bid prices verified, that precast segmental was more economical than steel construction by nearly 7%.7*8

JFK Memorial Causeway, U.S.A.

F I G U R E 3 . 5 5 . RIO-NIICI strut tion.

oi 131 dgc, wnrile\el

COII-

Total structure length is 1998 ft (609 m) with six interior spans of 265 ft (SO.8 m) and end spans of 204 ft (62.1 m), Figure 3.57. The layout has very severe geometry constraints. In plan, the east end of the bridge has two sharp horizontal curves connected to each other and to the west end tangent by two spiral curves; minimum radius is 1150 ft (350 m). In elevation, the bridge has a 2044 ft (623 m) vertical curve with tangents of 5.5 and 6.0 percent. Two sets of short-line forms employed by the contractor to cast the segments met the variable geometry requirements admirably. The accuracy of casting was such that only nominal elevation adjustments were required at the abutments and the center-span closure pours.s The single-cell box girder superstructure is continuous for the total length of the bridge. Typical cross-section dimensions are indicated in Figure 3.58. Prestressing tendon layout is illustrated in Figure 3.59 for a typical interior span. Fifty-five tendons were required for negative moments and 22 for positive-moments. The majority of nega-

FIGURE 3.56. Rio-Niteroi

tries.

RI

dgr.

launching gan-

109

tive-moment tendons were inclined in the web and anchored at the face of the segments. Anchorage of six tendons at the face of the first segment adjacent to the pier segment (three in each web) produced a large upward shear force at the face of the pier segment, which was not overcome until the erection of several additional segments. The midspan positive-moment tendons are continuous through the cast-in-place closure joint at midspan. These tendons, indicated by capital letters in Figure 3.59, were placed in preformed ducts upon completion of erection of the segments in a span and the closure pour consummated. All positivemoment tendons were anchored in the top flange. The precast segments are typically 14 ft 2 in. (4.3 m) in length and the closure pour at midspan is 4 ft 4 in. (1.3 m) long.7,R The precast segments are reinforced with prefabricated mild steel reinforcement cages, in addition to the primary longitudinal prestressing tendons, Figure 3.60, and transverse prestressing in the top flange. Web shear reinforcement varies depending on the location of the segment. The 145 precast segments were cast in a plant located near the bridge. This plant was equipped with two casting molds, each producing one segment per day. A 12-hour steam curing period was used and a concrete strength at 28 days of 5000 psi (34.5 MPa) was achieved.’ Because of the curved layout of the bridge and its relative shortness, the use of a launching gantry would have been uneconomical. Segments were placed by a 200 ton (180 mt) mobile crane on land, or on a barge over water, Figure 3.61. Construction of this bridge started in May of 1971, and it was opened to traffic on December 18, 1972. 3.10 JFK Memorial Causeway, U.S.A. A portion of the JFK Memorial Causeway represents the first precast, prestressed, segmental box girder completed in the United States. Opened to traffic in 1973, this 3280 ft (1000 m) long structure spans the Gulf Intercoastal Waterway in Texas to connect Corpus Christi and Padre Island. It was designed by the Bridge Division of the Texas Highway Department under the supervision of Wayne Henneberger. The Center for Highway Research, University of Texas at Austin, under the supervision of Prof. John E. Breen, assisted in the design and also built and tested a one-sixth scale model of the bridge to check design requirements and construction techniques.g

E LIRGS.

E PIER I

E PIER 2

203’.9”

2 6 5 .- 0 ‘,

I

E PIER 4

E PIER 3 265’.0

I

E PIER 5

265*-O”

t P I E R 6 265’-0”

E PIER 4

”

E 265,-O”

PIER 7 203’.9”

U N I T S N O . 7. 2 7 . 4 7 . 6 7 . 67,107. 1 2 7 AND 147 A R E C A S T IN P L A C E (DECK CLOSlNG UNITS)

ELEVATION

FIGURE 3.57. Bear River Bridge, elevation, f’ron~ ref. 8 (courtesy of the Prestressed Concrctc Institute).

I--- %

ROADWAY

39’-6”

2-6’!

I5’-0’

6’4

I’-()”

yI_ 4”

Ia’-0” I

-

I

FIGURE 3.58. Bear River Bridge, typical cross section, from ref. 8 (Courtesy of the Prestressed Concrete Institute).

HALF INTERIOR SPAN TENDON ELEVATION

HALF SECTION AT

MIDSPAN

TENDON

HALF SECTtON AT PIER DISTRIBUTION

FIGURE 3.59. Bear River Bridge, typical center-span tendon elevation and distribution, from ref. 8 (courtesy of the Prestressed Concrete Institute). 111

112


’

P

FIGURE

3.62. JFK Memorial Causeway, balanced cantilever construction (courtesy of J. E. Breen).

FIGURE 3.60. Bear Kiver Bridge, longitudinal prestress ducts in forms (courtesy of the Prestressed Concrete Institute).

FIGURE 3.61. Be,u

Ki\ cl Bi idge, crcc tion b v barge-mounted crane (courtesy of the Prestressed Concrete Institute).

The structure consists of thirty-six 80 ft (24.4 m) long approach spans of precast, prestressed bridge beams and the 400 ft (122 m) total length segmental bridge spanning the Intercoastal Waterway. The segmental portion of this structure has a center span of 200 ft (61 m) with end spans of 100 ft (30.5 m). The segments were precast, transported to the site, and erected by the balanced cantilever method of construction using epoxy joints, Figure 3.62. The precast, segmental superstructure consists of constant-depth twin box girders with a 2 ft (0.61 m) cast-in-place longitu-

dinal closure strip, Figure 3.63. Segments are 10 ft (3.05 m) in length and in cross section, are 8 ft (2.44 m) in depth, and have a nominal top flange width of 28 ft (8.53 m). The top flange or deck is of constant dimension longitudinally but of variable thickness in a transverse direction. The bottom flange is of constant dimension transversely but varies longitudinally from 10 in. (254 mm) at the pier to 6 in. (152 mm) at 25 ft (7.62 m) from the pier center. Segments were cast with male and f-emale alignment keys in both the top and bottom flanges as well as large shear keys in the webs, Figure 3.64. Integral diaphragms were cast with the pier segments, Figure 3.65. Both matching faces of the segments were coated with epoxy, and temporary erection stressing at both top and bottom of the segments precompressed the joint before installation of the permanent post-tensioning tendons. The segments were erected by a barge-mounted crane. As each segment was erected, it was tilted 21 degrees from the in-place segment, so that a pair of hooks in the top of the segment being erected engaged pins in the segment previously erected. The new segment was then pivoted down by the sling until its shear key slipped into the mating shear key of the previously erected segment.g Figure 3.66 shows a permanent tendon being tensioned and the temporary working platform. The design concept on this project utilized prestressing tendons in the top flange for dead-load cantilever stresses; after closure at midspan, continuity tendons were installed for the positive moment, Figure 3.67. Research on the model testing of the bridge is documented in references 10 through 15 with particular emphasis in reference 14 on lessons learned during construction that might facilitate or improve similar projects.

Saint And& de Cubzac

Sym. B Q

28 ft. (8.53 m)

L

-m

6’-8” (2.03 m)

6 ft. (1.83 ml

2 z

l---l

113

Bridges, France

8

h .?

7 ft. (2.13 m)

al :- s

T-10” (2.39 m)

13 ft. (3.96 m) /-

I-

FIGURE 3.63. JFK Memorial Causeway, typical cross section. Bottom slab

thickness varies from 10 in. (254 mm) at pier to 6 in. (152 mm) at 25 ft (7.62 m) from pier center.

FIGURE 3.66. JFK Memorial C;IUSC\V;I~,

prestressing permanent tendon (courtesy of J. E. Breen).

FIGURE 3.64. JFK Memorial Causeway, precast seg-

ment in casting yard (courtesy of J. E. Breen).

FIGURE 3.65. J FK Xlemorial Causeway, construction

view showing pier segments with diaphragms (courtesy of J. E. Breen).

3.11 Saint And& de Cubzac Bridges, France Opened to traffic in December 1974 after a construction period of 29 months, this important structure crosses the Dordogne River north of Bordeaux on the South Atlantic Coast. A view of the finished bridge is shown in Figure 3.68. The main river crossing has a total length of 3800 ft (1162 m) with approach land spans of 190 ft (59 m) and main river spans of 312 ft (95.3 m), Figure 3.69. Two intermediate expansion joints located at the point of contraflexure in the transition spans separate the deck into three sections for concrete volume changes. The center section has a length of 1920 ft (585 m). The main piers have rectangular hollow box shafts supported by circular opendredged caissons 30 ft (9 m) in diameter. Approach piers have an I section. Another structure, constructid under the same contract, consisted of twin bridges 1000 ft (307 m) in length with typical 162 ft (49.5 m) spans in an

114


Cantilever (negative moment) tendons

8 Main pier

C$ Central span 100 ft (30.5 mJ -I

FIGURE 3.67. JFK Memorial Causeway, system of prestressing tendons.

FIGURE 3.68.

Saint Andre de Cubzac Bridge, view of the finished bridge over the Dordogne River. area north of the main crossing where poor soil conditions did not permit stability of an embankment. Altogether the deck area is 97,000 sq ft (29,500 m2), entirely of precast segmental construction. The typical cross section is a single box 54.4 ft (16.6 m) wide with transverse ribs both in the side cantilevers and between webs, Figure 3.69, to provide structural capacity to the deck slab under traffic loads. A casting yard located along the bank of the Dordogne River produced the 456 segments for both bridges (main crossing and north viaducts) in three casting machines (two for the typical segments and one for the special segments such as pier, hinge, or end segments). Moderate steam curing at 86°F (30°C) for 12 hours in a movable kiln enclosing the newly cast segment and its match-cast counterpart allowed a one-day cycle and proved very efficient in avoiding any geometric corrections.

Segments were placed in the structure by the beam-and-winch method either on land (for the northern viaducts or the approach spans of the main river crossing) as shown in Figure 3.70 or over water for the main spans as shown in Figure 3.71. This project was the occasion for a further improvement in the placing scheme by beam and winch, whereby the pier segments could be precast and placed with the same type of equipment as shown in principle in Figure 3.72. A provisonal tower prestressed against the pier side face allowed the pier segment to be installed upon the pier cap, with the beam and winch later used for cantilever placing. To keep the segment weight to a maximum of 110 t (100 mt) the pier segment, representing the starting base of each cantilever, had been divided into two halves placed successively, Figure 3.73. Figure 3.74 shows the lifting of the last closure segment.

3.12

Saint Cloud Bridge, France

A connection between the peripheral Paris Ring Road and the Western Motorway (A- 13) required the construction of a bridge over the Seine extended by a viaduct along the left bank leading to the Saint Cloud Tunnel, Figures 3.75 and 3.76. This structure has two traffic lanes in each direction. It will be duplicated later by a similar adjoining structure when the congested Saint Cloud Tunnel is duplicated. O r i g i n a l d e s i g n o f t h i s bridge contemplated a steel structure. However, an alternative design utilizing precast segments and

+

BORDEAUX

8'

I I

(0 a N

0,”

Al’iDRE

DE

CUBZAC

I)

I I

8 ---Hm

I

6,00 /

FIGURE

3.69.

1 ,

Saint Andrk de Cubzac Bridge, elevation and cross section.

3.71. Saint Andrk de Cubzac Bridge, beamand-winch segment placing over water. FIGURE

FIGURE 3.70. Saint And& de Cubzac Bridge, beamand-winch segment placing over land.

115

WlNCt B

01 FIGURE 3.72.

02

116

.

A

03

Saint Andrk de Cubzac Bridge, placing precast pier segments.

FIGURE 3.73. Saint Andre Cubzac Bridge, lifting second half pier segment.

E

M

Saint Cloud Bridge, France

the balanced cantilever method of construction, submitted by the contractor, permitted substantial savings and was accepted by the authorities. The bridge has a total length of 3618 ft (1103 m) with a constant-depth superstructure. It includes two sections: the bridge over the Seine, which is a 1736 ft (529 m) long curved structure; and a 1883 ft (574 m) long viaduct, which follows a straight layout along the bank of the Seine and then crosses the Place Clemenceau, on a 2260 ft (690 m) radius curve, by an access ramp to the Saint Cloud Tunnel. It includes 16 spans divided as follows (refer to Figure 3.76): Seine Bridge: 160.8,288.7,333.8,296.0,150.9,and two 219.5 ft spans (49, 88, 101.75, 90.25, 46, and two 66.9m) Common area: 66.4 ft (20.24 m) up to the expansion joint, and then 153.1 ft (44.66 m), total 219.5 ft (66.9 m) Viaduct: five219.5; 285.4,210.0,and 137.8ftspans (five 66.9; 87, 64, and 42 m) Architectural considerations led to the choice of a 11.8 ft (3.6 m) constant-depth three-cell box girder with slopingexternalwebs with nooverhangs, Figure 3.77. Segments are 7.4 ft (2.25 m) in length with a record width of 67 ft (20.4 m), their average weight varying from 84 to 143 tons (76 to 130 mt). Since the superstructure has a constant depth, the bending capacity is adjusted to the moment dis-

117

tribution by varying the bottom flange thickness, which decreases from 3 1.5 in. (800 mm) at the river piers to 7 in. (180 mm) at midspan. To accommodate the curvature of the bridge the segments in this area are cast, in plan, in a trapezoidal shape. A 4.5% superelevation is obtained by placing the units over the piers in an inclined position. Three-dimensional prestressing was used in the superstructure: the main longitudinal prestress, transverse prestress in the deck, and a vertical prestress in the webs to accommodate shear. After the closure joint at midspan was cast, additional longitudinal prestress tendons were installed to provide continuity. Superstructure segments were precast in a plant on the right bank of the Seine. Two casting molds were used for fabrication of the segments. Each mold had an external formwork and an internal retractable formwork. The adjacent, previously cast segment was used as a bulkhead to achieve a match-cast joint. For erection, segments were transported on a trolley to a cable-stayed launching gantry of unusual size and capacity. It was of high-yield steel construction, 402 ft (122.5 m) in length and weighing 250 tons (235 mt), with a maximum load capacity of 143 tons (130 mt). The constant-depth gantry truss was supported on central and rear legs, which were tunnel shaped to allow passage of the precast segments endwise. At the central support, a 52.5 ft (16 m) high tubular tower topped

,.,. 59 _ls .,’

- --

FIGURE 3.76. Saint Cloud Bridge, plan view.

COUPE TRANSVERSALE D’UN VOUSSOIR TYPE

FIGURE 3.77.

Saint Cloud Bridge, longitudinal and typical cross section.

120


with a saddle provided a large eccentricity to the three pairs of cable stays, which improved the negative-moment capacity at this support location. At the forward end of the gantry an additional leg was used as a third support point during launching and pier segment placing, Figure 3.78. The launching girder was moved forward on rails mounted on the completed superstructure, by

sliding on pads placed at the central and rear legs. The launching girder, in cross section, was triangular in shape. The base of this triangle included two structural steel I sections, which served as tracks for the segment transportation trolley. The diagonal bracing of the launching girder consisted of tubular steel members. The girder was fabricated in ten sections, approximately 39 ft (12 m)

FIGURE 3.78. Saint Cloud Bridge, segment placing.

PLACING OF PILE UNITS

AVANCEMENT DU +ORTlQUE D E LANCEMENT. MOVING

THE

MISE E N

TRUSS

PLACE

DES P L A C I N G T H E U N I T S I N CANTELIVER

FIGURE 3.79.

Saint Cloud Bridge, sequence of operations in moving launching girder.

Saint Cloud Bridge, France in length, so as to be transportable over the highways. These units were assembled at the job site by prestressing bars. The sequence of operations in moving the launching girder forward is illustrated in Figure 3.79 and included the following operations: Placing pier se
temporary front leg supported just in front of the pier. Launching of the gantry: The gantry slid on rails at the rear leg and rolled over an auxiliary support placed atop the pier segment. The central leg, during this travel, crossed the gap between the cantilever end and the pier unit. Placing typical segments in cantilever: In this phase the gantry was supported at two points: the central leg placed over the pier and the rear leg anchored

“2 3F6

FIGURE 3.80. the river.

121

Saint Cloud Bridge, sequence of operations of launching gantry over

122


FIGURE 3.81. Angers Bridge, longitudinal section. at the end of the last completed cantilever. The segments were lifted by the trolley at the rear end of the girder, moved forward, after a rotation of a quarter turn, and then placed alternatively at each end of the cantilevers under construction. As a result of the horizontal curvature of the structure, the transverse positioning of a segment was accomplished both by moving the segment transportation trolley sideways relative to the girder [possible side travel of 3 ft (0.9 m) on either side] and by moving the launching gantry itself sideways relative to its bearing support on the bridge. Thus, the construction of a cantilever required one, two, or three different positions of the gantry, according to the curvature radius and length of span, as shown in Figure 3.80. Work started in October 1971 and was completed in Dkcember 1973. Placing the 527 precast segments in the 3600 ft (1097 m) long superstructure took exactly one year. In terms of erection speed, a more interesting project was successfully carried out on a precast segmental bridge awarded,to Campenon Bernard. A unique set of circumstances arose where a bridge over the Loire River at Angers could be fitted to use simultaneously the dimensions and casting machines of Saint Andre de Cubzac Bridge, which had recently been completed, and the gantry of Saint Cloud Bridge. The 2577 ft (786 m) long structure rests on 10 piers and has 280 ft (85.1 m) typical spans, Figures 3.81 and 3.82, using a single box girder with ribbed

FIGURE structure.

3.82. Angers Bridge, view of the completed

deck slab units identical to the sections used at Saint Andre de Cubzac. The construction contract was signed in August 1974 and the superstructure was completed in May 1975. All segments were placed between January and May 1975, in a little less than five months, corresponding to an average erection speed of 26 ft (8 m) per day of finished deck. 3.13 Sallingsund Bridge, Denmark Sallingsund in Northern Jutland between Arrhus and Thisted is a site of great natural beauty. Construction of a bridge in such an environment was the object of careful study, which concluded, after an international competition, in the selection of a precast segmental structure, Figure 3.83, resting on piers of a unique design. This structure has two end spans of 167 ft (5 1 m) and 17 interior spans of 305 ft (93 m). There are 18 piers between the two abutments. The level of the roadway reaches 100 ft (30.5 m) above the water at the center span and 82 ft (25 m) at the abutments. The two center spans are navigation spans requiring 85 ft (26 m) vertical clearance over a width of 197 ft (60 m). The bridge deck accommodates two traffic lanes, approximately 13 ft (4 m) each, two cycle paths, and two sidewalks for a total width of 52.5 ft (16 m), Figure 3.84. The

FIGURE 3.83. Sallingsund Bridge, view of the completed structure.

t

L

f FIGURE 3.84. Sallingsund

Bridge, typical dimensions.

124

Precast Balanced Cantilever Girder

Bridges

superstructure consists of precast concrete box girder segments 11.7 ft (3.57 m) in length, with epoxy match-cast joints, which are prestressed together. Segment depth varies from 8.2 ft (2.5 m) at midspan to 18 ft (5.5 m) at the pier. The precast superstructure segments were match-cast by the short-line method (see Chapter 11). There are altogether 453 segments varying in weight from 86 t (78 mt) to 118 t (107 mt). The typical segment shown in Figure 3.85 has web corrugated shear keys together with top and bottom flange keys. Hinge segments equipped with a roadway expansion joint for thermal movement of the superstructure are placed every other span near the point of contraflexure. A hinge segment with its diaphragm is shown in Figure 3.86. Segments are placed in the structure in cantilever with a cable-stayed launching gantry. Transfer from the casting area and the storage yard to the construction site and the launching gantry is achieved by a low-bed dolly pushed by a tractor, Figure 3.87. The gantry shown in Figure 3.88 should look

FIGURE 3.85. Sallingsund Bridge, view of a typical segment.

FIGURE port.

3.87. Sallingsund Bridge, segment rrans-

FIGURE 3.86. Sallingsund Bridge, hinge segment with diaphragm.

FIGURE

3.88.

Sallingsund BI idge, I,~un&ing

g,~ntry.

B-3 South Viaducts, France

125

Figure 3.91 presents a plan of this project and shows a subdivision in accordance with the type of cross sections used. It includes the following main subdivisions: 1. 2.

The main viaduct VP 1-A through VP 1-J. The main viaduct VP 2-A and VP 2-B.

3. The viaducts ramps to the 4. The viaducts ramps to the

FIGURE 3.89. Sallingsund Bridge, elevation of main

piers in water.

and have 860,000 sq ft (80,000 m”) of bridge deck. The project is in a congested area that required the crossing of railway tracks, canals, and more than 20 roads; its diverse structural geometry contains curves, superelevation ranging from 2.5 to 6% and grades up to 5%. .

FIGURE 3.90. B-:l South Viaduct, overall view.

Vl and V2, which are access main viaduct VP 2. V3 and V4, which are access National Road RN3.

The original design for this project, prepared by the French authorities, was based on conventional cast-in-place construction of the superstructure in complete spans using movable formwork. The contractor proposed a more economical design based on the use of precast segments. The alternative design had advantages in erection, wherein parts were erected by a launching truss and parts by a mobile crane in conjunction with an auxiliary truss and winch. The use of precast units allowed a deeper and thus a more economical superstructure, because the space required for formwork did not have to be deducted in the clearance requirements over existing roads and other facilities. The superstructure has a constant depth of 6.5 ft (2 m), consisting of three different cross sections, Figure 3.91. Different width and transitions were accommodated by varying the width of the castin-place median slab connecting the top flanges of the precast segments. Only the V3 and V4 access ramps were of conventional cast-in-place construction. The webs of the precast segments have a constant thickness of 12 in. (310 mm), increased in some cases to 20 in. (500 mm) near a pier. Webs are stiffened by an interior rib, which also serves to anchor the longitudinal prestressing inside the box rather than in the web at the end of a segment. Where the webs are not thickened near a support, they are prestressed vertically by bars to accommodate shear forces. The top flanges of the segments are cantilevered 10 ft (3 m). In the case of segment types 2 and 3, Figure 3.9 1, the top flange cantilever between box sections is 9 ft (2.75 m). The top flange follows the superelevation of the roadway. The thickness of the cast-in-place longitudinal slab between box girders varies from 7.9 to 13.8 in. (200 to 350 mm), depending upon its width. The total superstructure is supported on neoprene or sliding bearings. Expansion joints are spaced at distances up to 1970 ft (600 m) and are


126

1 5 . 2 5 in T -- - - - - - 1

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TYPE

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VCkl5501RS.L~

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w 3,401-n

3,40m

SUD SOUTH

RN3 +

FIGURE

3.91.

B-3 South Viaduct, plan showing segment type location.

located in special hinge joints near a pier. Superstructure spans vary from 89.6 to 174 ft (27 to 53 m), with 90% of them being in the range of 111 to 125 ft (34 to 38 m). This project required 2225 precast segments, all manufactured by the short-line method (see Chapter 1 l), which involved the following operations:

8. Transfer of the segment, eight hours after curing, to a more permanent storage until required for erection. 9. Return of the mold bottom, after temporary storage, to the casting area for reuse.

1.

1. 2. 3.

2. 3. 4. 5. 6.

7.

Subassembly of mild steel reinforcing on a template. Storage of subassembly units. Assembly of complete reinforcement cages including tendon ducts. Placing of the cages in the forms. Concreting and curing of the segments. After concreting and curing, transportation of the segment by a dolly to a position where one end would act as a bulkhead for the casting of the next segment. At the same time its position was adjusted to conform to the proper geometric configuration of the superstructure. Transfer of the segment that had previously acted as the bulkhead to temporary storage for further curing.

Curing of the segments was accomplished with low-pressure steam in the following 4&-hour cycle: An initial l&hour curing period at 35°C. A two-hour temperature rise reaching 65°C. A one-hour curing period at a level of 65°C.

The short curing cycle can be accomplished if the following conditions are satisfied: use of a proper cement, preheating of the materials to 35”C, rigid forms, and proper supervision. Casting of a segment required nine hours, allowing two segments per day per form; the four forms used produced a total of eight segments per day. Erection of precast segments by the launching gantry shown in Figure 3.92 is schematically illustrated in Figure 3.93. After being rotated 90”, segments V2 and V’2 were placed at the same time by means of two trolleys suspended from the bottom chord of the launching girder, Figure 3.94.

B-3 South Viaducts, France

3.92. H-:5 in operation. FIGURE

127

V2 and V’2 were then attached to the previously erected segments by temporary prestressing. During the erection operation of V2 and V’2 a transport dolly delivered segment V’3, then V3, and so on. In this manner the erection of segments could be carried out without being delayed by transportation of the segments from the storage area. In addition, the threading and stressing of the permanent prestressing tendons were independent of the erection cycle, since the tendons were anchored in the internal ribs and could be prestressed inside the box girder. Where the span length was less than 125 ft (38 m), the pier segments were placed by the gantry in its normal working position. The pier segment position was adjusted from a platform fixed to the top of the pier to avoid delaying the placement of cantilever segments at the preceding pier. For the few

South Viaduct, launching gantry

The matching faces of the segments being erected and the previously erected segments, V 1 and V’l, were coated with epoxy joint material. Segments

(b)

FIGURE 3.93. B-3 South Viaduct, erection sequence. (a) Placing the units: The two trolleys bring the units V2 and V’2 which will be placed, after rotation at 90”, against the units VI and V’l. During this time, the lorry carries the units V’3, then V3, and so on. (b) Launching the truss: The rear and the central legs are lifted above the piers PO and Pl. ‘The truss is supported by trestles and trolleys in Pl and P2 and moves forward by the action of the trolley motors until the legs reach Pl and P2. Thus the truss has advanced along one span length and can place the pile-unit in P3 and the cantilevers from P2.


t.*

--

FIGURE 3.94. B-3 South Viaduct, placing two seg-

ments in balanced cantilever.

larger spans, the pier segment was placed after closure of the preceding completed spans and advancement of the launching gantry. The center leg was advanced out onto the last completed halfspan cantilever, but it remained in the proximity of the pier. Launching of the gantry to the next span was achieved by using the two segment transportation dollies temporarily fixed on the completed superstructure by two auxiliary steel trusses. The high degree of mechanization of the gantry together with the repetitive nature of the project allowed speedy erection. A typical 130 ft (39 m) span was erected and completed in two working days. To maintain the construction schedule 2nd minimize required erection equipment, the super-

structure segments were placed simultaneously by two different methods. The launching gantry previously described placed 57% of the segments and a mobile crane in conjunction with a movable winch frame erected the remaining ones. The latter method was used where access was available for a truck-mounted crane and the segment transportation dolly. The truck-mounted crane could easily be used along the centerline of the structure to place segments at outboard cantilever ends. However, its use became complicated in the midspan area, particularly when it was used to place the closure segments. To solve this problem, an auxiliary truss equipped with a winch was used in conjunction with the mobile crane. This truss was supported at one end over the pier where cantilever construction proceeded and at the other end over the last completed cantilever arm, which might or might not require a temporary support pier, Figure 3.95. The segments were lifted by a trolley-mounted winch traveling along the truss. This truss was also used to stabilize the cantilevers during erection, since it was fixed to the pier and the completed portion of the superstructure. After the pier segment was positioned by the mobile crane, the frame was launched with the trolley in a counterweight position at the rear of the frame. When the span exceeded 65 ft (20 m), the front of the frame was held by the crane, This structure exemplifies an innovative application of precast balanced cantilever segmental construction to a difficult urban site and shows its adaptability to almost any site conditions.

B-3 South Viaduct, auxiliary truss for segment assembly (crane placing). (1) Auxiliary truss, (2) winch for segment lifting, (3) precast segment, (4) possible tempnrary support (as required), and (5) concrete cantilever stability device. FIGURE 3.95.

Alpine

Motorway

3.15 Alpine Motorway Structures, France The new Rhone-Alps Motorway system in South East France includes 220 miles (350 km) of tollways, of which 60 miles (100 km) are an optional section, between the cities of Lyons, Grenoble, Geneva, and Valence in order to improve communications between Germany and Switzerland on one hand and South France and Spain on the other. The motorway is situated among the beautiful western slopes of the Alpine mountain range (see the location map, Figure 3.96). The first 160 miles (250 km) include the following structures: Ten viaducts varying in length between 500 and 1300 ft (150 to 400 m) Two hundred overpass bridges Fifty underpasses Such a project afforded an exceptional occasion to

Structures,

France

129

optimize the structures in terms of initial investment and low maintenance costs. The underpasses had to accommodate a variable and often considerable depth of fill to reduce the constraints of the longitudinal profile in this mountainous region. The ideal answer was found in the use of reinforced concrete arch structures, which proved extremely well adapted and had a cost approximately half that of conventional girder bridges. Apart from the first section of the motorway (East of Lyons), which had to be built immediately and therefore called for conventional solutions (cast-in-place prestressed concrete slab), and except for certain special situations (excessive skew, railroad crossing, and so on), a careful study showed that the remaining 150 overpass bridges should be of precast concrete segmental construction, which were 20% more economical than other methods and practically maintenance free. The study further showed that segmental construction

FIGURE 3.96. Alpine Motorway, location map.

130


should be extended to viaduct structures and that all segments for both overpasses and viaducts could be economically built in a single factory located near the center of gravity of the motorway network. The maximum carrying distance was no more than 75 miles (120 km) and the average was 40 miles (60 km). Figures 3.97 and 3.98 are views of a typical viaduct and a typical overpass in the motorway network. The two-span and three-span overpass bridges have spans ranging from 59 to 98 ft (18 to 30 m). A variety of standardized precast cross sections were developed for this project, depending upon span and width requirements. The first structures used single and double-cell trapezoidal box sections, although later on voided slab sections were preferred, as illustrated in Figure 3.99a. This solution proved aesthetically pleasing and very simple to manufacture and assemble. The viaducts had to satisfy a wide range of environmental requirements. It was found that span lengths could be limited at all sites to a maximum of 200 ft (60 m),

4

2.60

c

\ +..

I ---. 4.m

-.

-4

(.

FIGURE 3.99. Alpine Motorway, typical sections of overpass and viaducts. (a) Overpass segments. (b) Viaduct segments.

FIGURE 3 . 9 ’ 7 .

FIGURE 3.98.

Alpine Motorway, view of a viaduct

Alpine Motorway, view of an overpass.

which allowed a constant-depth superstructure with precast segments, Figure 3.996. Segment manufacture was carried out in a factory close to the new motorway with easy access to the existing highway system, which was used to haul all segments to their respective sites. The factory had two parallel bays, Figures 3.100 and 3.10 1, one for the overpass segments and one for the viaduct segments. Segments for the overpasses, Figure 3.100, were match-cast by the short-line method with their longitudinal axis in a vertical position. The bottom segment was a previously cast unit. The segment at the top was then match-cast against the segment on the bottom. After the unit being cast had reached the required strength, the bottom unit was removed for storage, and the en-

-

40'

PORTAL

(RAIit

fb)

FIGURE 3.100. Alpine Motorway, precasting factory.

132


FIGURE 3.101. Alpine Motorway, general view of precast factory and segment storage.

tire process repeated. Figure 3.102 is a view of a segment in a vertical match-casting position. Erection procedure for a typical three-span overpass structure was as follows: 1 . After the foundations and pier columns had been constructed, precast concrete slabs were placed on sand beds adjacent to the piers to form foundations for the steel falsework towers. The precast slabs and towers were reusable for subsequent bridges. The erection commenced with placement of the first segment on top of four partially extended 25-ton jacks, Figure 3.103~. 2. The second and third segments were placed and prestressed to the first segment, Figure 3.103b. The joints between the segments were epoxy coated as the segments were erected. The prestressing of the second and third segments to the first segment consisted of temporary bars above the top surface of the segments, and other temporary tendons within the segments near the bottom of the segments. The four 25-ton hydraulic jacks under the first segment were then replaced by four partially extended loo-ton hydraulic jacks positioned under segments two and three. The jacks were supported on teflon sliding bearings. 3. The remaining segments were then erected, forming cantilevers on each side of the falsework The prestressing of the towers, Figure 3.103~. segments consisted of temporary tendons positioned above the segments, as indicated in Figure 3.103. 4. The erection of the segments could take place simultaneously at both piers, or one could precede the other, Figure 3.103d. Observe that at this stage of erection each assembly of segments was independently supported on four large hydraulic jacks and hence could be raised, lowered,

FIGURE 3.102. Alpine hlororway, casting of segments.

vertical

match

or rotated if required to adjust its position with respect to its pier or to its counterpart at the opposite pier. This method eliminated the need for a castin-place closure joint at midspan of the central span. Through the adjustment of the hydraulic jacks, perfect mating of the two centermost match-cast segments could be achieved when the assemblies of segments were slid together as indicated. The time required to erect the superstructure was significantly reduced by avoiding the use of a cast-in-place closure joint. 5. At this point in the erection, the first group of permanent prestressing tendons were inserted in preformed holes through the segments, after which they were stressed and grouted, Figure 3.103e. 6. The process proceeded with the erectiomof the remaining segments, Figure 3.103f 7. After installation of precast match-cast abutments, a second group of permanent tendons was installed, and finally the temporary falsework and temporary prestressing was removed, Figure 3.103g.

Alpine Motorway Structures, France

133

SECMENlS 18 b 2ST

SPAN

I8 te 30 m

J

lb)

TEMPORARY TIE URS

FIGURE 3.103. Alpine Motorway Bridges, erection scheme for typical three-span overpasses. (a) Placing the first and second segments. (b) Transfer to loo-ton jacks. (c) First half completed. (d) Joining precast assemblies by sliding. (e) Threading and stressing cables. v) Placing the end segments. (g) Threading and stressing last cables. Overpass structures of two spans could be erected using the technique illustrated above for three-span structures, Figure 3.104. As would be expected, the longer spans required the use of additional falsework towers. An overpass bridge, foundations plus piers and superstructure, could be constructed in less than two weeks. Figure 3.105 shows a typical segment being placed in the over-

pass bridge with a mobile crane. Temporary prestress over the deck slab is shown in Figure 3.106. The viaducts required the manufacture of larger segments in the same precasting factory used for the overpass segments, but with casting proceeding in the usual short-line horizontal fashion. Three casting machines were used simultaneously to produce all viaduct segments.

134

Precast HYORAULIC

SLIDE

Balanced

Cantilever

JAcu5

Girder

Bridges

EMFORARY PRE5TR6’59

_

5ilDE

FIGURE 3.104. Alpine Motorway Bridges, erection scheme for two-span overpass

bridges.

Erecting segments in the various structures required the use of a launching gantry of an exceptionally light and elaborate design, allowing easy transportation and erection from site to site, Figure 3.107. A typical 200 ft (60 m) long cantilever in-

eluding 25 segments, one pier segment weighing 48 t (44 mt), and 24 typical segments weighing 36 t (33 mt) could be accomplished in six to eight working days, including launching the gantry to the following pier and achieving continuity with the preceding cantilever. The maximum rate of segment placing was 12 units in a single day. This project is another interesting application of mass-production techniques and the standardization of segmental construction. 3.16

Bridge over the Eastern Scheldt, Holland

The bridge over the Eastern Scheldt, otherwise known as the Oosterschelde Bridge, Figure 3.108,

FIGURE 3.105. Alpine Motorway, segment placing in overpass with crane.

FIGURE 3.106. Alpine Motorway, provisional prestress over deck slab.

135

Br-idge Over the Eastern Scheldt, Holland

3.107. ,\lpine >lotol viaducts with launching gantry.

FIGURE

wn, segment placing

in

time restraints for construction, and scarcity of labor, prefabrication was required to a very high degree. Since the precast pile elements would be large and heavy, it was decided that the pier and superstructure segments should be equally large and heavy, in the range of 400 to 600 tons.i6 A casting yard, Figure 3.110, capable of producing all the various precast elements for the structure was constructed near one end of the bridge. This facility provided all the advantages of yard production techniques and the potential for high quality control. The 14 ft (4.27 m) diameter cylinder piles have 14 in. (0.35 m) thick walls and were cast vertically in 20 ft (6 m) lengths. They were then rotated into a horizontal position where they were aligned,joints concreted, and the pile post-tensioned. In this manner piles were produced in required lengths up to 165 ft (50 m). The assembled pile was then transported by barge to the site, where a derrick picked it up at one end and rotated it into its verti-

600 tons

FIGURE 3.108. Bridge over the Eastern Scheldt,

overall view of the structure. is part of a project known as the Delta Works, which closed the mouths of many rivers and streams southwest of Rotterdam to protect the coastline from flooding. The bridge consists of fifty-five 300 ft (9 1.4 m) spans, a roadway width of 35 ft (10.7 m), and a vertical navigation clearance of 50 ft (15.2 m). Parameters considered in the choice of structural type and span were economics, foundation restraints, and ice loads. Substructure consists of three cylinder piles with a caisson cap and an inverted V pier, Figure 3.109. The superstructure was assembled from seven precast elements, one pier segment, and two each of three progressively smaller segments to produce one double cantilever span of 300 ft (91.4 m). The bridge design, therefore, consists of very large prestressed cylinder piles, precast pier elements posttensioned together, and precast superstructure elements erected and post-tensioned together to form a double cantilever system with a joint at each midspan location. Because of open-sea conditions,

Cytindrml hollow

FIGURE 3.109. Bridge over the Eastern Scheldt,

schematic of precast elements in the structure (courtesy of the Portland Cement Association).

FIGURE 3.110. Bridge over the Eastern Scheldt, view of precasting plant (courtesy of the Portland Cement Association).

136


cal position. Cylinder piles weighted from 300 to 550 tons (270 to 500 mt). The pier cap was also precast at the same yard, where it was post-tensioned circumferentially and vertically. The inverted V portion of the pier was also precast with provision for on-site post-tensioning to achieve final assembly.16 Figure 3.111 shows the bridge under construction. The temporary enclosures between each section are to protect the cast-in-place joint concrete against cold weather. Cast-in-place joints 16 in. (0.4 m) wide were used, with faces of the precast elements serrated to act as shear keys. The superstructure segments were all set from a traveling steel gantry, Figure 3.111, that extended over two and one-half spans at a time. Segments were barged to their final location, then hoisted in symmetrical order about each pier. The joints were concreted and the primary stressing completed be-

fore the next series of segments were hoisted into position. Erection sequence is depicted in Figure 3.112. An aerial view of various stages of construction is shown in Figure 3.113. A typical cycle for two spans of superstructure, not including the pier segment, involving the raising, concreting, and stressing of 12 segments, was three weeks.

3.17 Captain Cook Bridge, Australia

FIGURE 3.111. Bridge over the Eastern Scheldt, view of launching truss and enclosure for cast-in-place joints (courtesy of the Portland Cement Association).

This structure carries a six-lane highway over the Brisbane River in Brisbane, Australia, as part of the Riverside Expressway and South-West Freeway designed to relieve the city’s overloaded traffic system. The navigation requirements were for a 300 ft (91.4 m) wide horizontal clearance with a vertical clearance of 45 ft (13.7 m) across 200 ft (61 m) and 40 ft (12 m) at either extremity. However, a 600 ft (183 m) span became necessary because of the skew crossing. Adequate bearing rock, at a reasonable depth, was found at the south bank such that the pier could be founded on a spread footing. At the north end, because of the steeply rising bank, the anchor span was limited to a span of 140 ft (42.7 m) and the abutment was designed as a counterweight connected to the superstructure by a prestressed tie-down wall, Figure 3.1 14.17 Once the navigation span requirements had been met, the remaining span lengths were selected to meet design requirements, while the superstructure depth boundaries had to fall within a maximum allowable grade requirement of 3% and the flood level. The superstructure is a dual

FIGURE 3.112. Bridge over the Eastern Scheldt, schematic of erection sequence (courtesy of the Portland Cement Association).

FIGURE 3.113. Bridge over the Eastern Scheldt, aerial view of construction showing various phases (courtesy of the Portland Cement Association).

ELEVATION

FIGURE 3.114. Capt. Cook Bridge, plan and elevation, f‘rom ref.. 17.

138


structure of prestressed concrete segmental twocell boxes, Figures 3.115 and 3.1 16.17 Steel rocker bearings were used to support the superstructure at piers 1, 3, and 4, and largediameter single steel roller bearings were used at pier 2. Lubricated bronze bearings sliding on stainless steel were used at the north abutment and for the movable bearings at the suspended spans. Steel finger joints, allowing a 10 in. (250 mm) maximum movement, were provided at each slid-

II

H-=---

FIGURE 3.115. Capt. Cook Bridge, cross section at

pier 3, from ref. 17.

FIGURE 3.116. Crpt. Cook Bridge, two-cell box gir-

der segment being erected (courtesy of G. Beloff, Main Roads Department).

ing bearing location and rubber and steel finger joints at the remaining locations.” The box girder segments have a maximum depth of 32 ft (9.75 m) and a minimum depth of 6 ft (1.83 m). Segment length is 8 ft 8 in. (2.64 m). A 16 in. (0.4 m) cast-in-place, fully reinforced joint was used between segments. Maximum segment weight is 126 tons (114 mt). A total of 364 precast segments were required in the superstructure with the two segments over the tie-wall in the south abutment being cast in place.” The ContracEor chose to locate the precasting operation on the river bank near the south abutment. This casting yard consisted of a concrete mixing plant, steam-curing plant, three adjustable steel forms, segment tilting frame, and a gantry crane to transport the segments to a storage area along the river bank. Segments were designed so that the top flange and upper portion of the webs had a constant thickness. The depth and lower portion accommodated all variations, allowing the contractor to cast in two sets of adjustable forms. Segments were cast with their longitudinal axis in a vertical position for ease of concrete placement around the prestressing ducts. Separate interior forms were constructed for each box to permit variations in the bottom flange and web thickness and size of fillets. Aft.er casting and curing, segments were lifted into a tilting frame to realign the segment into its normal position ready for handling and storage.i7 A floating crane, designed and built by the contractor, was used for erection of the segments. It was essentially a rectangular pontoon with mounted A-frame lifting legs rising to 120 ft (36.6 m) with adequate clearance to service the finished deck level, while the stability was sufficient to transport the segments to the erection position, Figure 3.117. An extended reach was required to position segments on the first two spans in the shallow water near the bank.17 Segments on each side of the pier were supported on falsework anchored to the pier shafts, Figure 3.118. From this point additional segments, as they were erected, were supported on a cantilever falsework from the completed portion of the structure. This falsework was fixed under the completed girder and supported from deck level, Figure 3.119. When the capacity of the pier to carry the segment unbalanced load was reached, a temporary prop support on driven piles was constructed before cantilever erection could continue.. Segment erection then proceeded on each side until either the joint position of the suspended

Other

Notable

Structures

139

FIGURE 3.119. Capt. Cook Bridge, cradle support trusses and temporary support tower (courtesy of G. Beloff, Main Roads Department).

FIGURE 3.117. Cap. Cook Bridge, segment being transported by barge derrick to final position (courtesy of G. Beloff, Main Roads Department). span was attained or the closure gap in span 3 was reached. The completed structure was opened to

traffic in 1971, Figure 3.120.

3.18 Other Notable Structures In Sections 3.2 through 3.15 the historical de-

velopment of precast segmental bridges with match-cast joints has been illustrated by examples,

ranging from the first structure at Choisy-le-Roi to the largest applications such as the Rio Niteroi and Saint Cloud bridges. Emphasis has been placed on North American experience as well as on the advantages of precast segmental construction for urban structures (B-3 Viaducts) or repetitive applications (Alpine Motorways). Two particularly outstanding structures, deserving special mention because of their size and characteristics where pre-

cast segmental was used with conventional joints (not match-cast) were the Oosterschelde and Captain Cook Bridges (Sections 3.16 and 3.17). Before closing this important chapter, let us briefly give due credit to several other contemporary matchcast segmental bridges. 3.18.1

CALIX BRIDGE, FRANCE

This 14-span superstructure has a maximum span length of 512 ft (156 m) over the maritime

FIGURE 3.118. Capt. Cook Bridge, support for segments on each side of pier (courtesy of G. Beloff, Main Roads Department).

Cook Bridge, completed FIGURE 3.120. Capt. structure (courtesy of G. Beloff, Main Roads Department).


140

1.39

3.h2 13.42

e

FIGURE

L

3.121. Calix Viaduct, near Caen, France general dimensions. waterway and typical 230 ft (70 m) spans in the approaches on both banks. Dimensions are shown in Figure 3.12 1. The deck consists of two parallel box girders connected by a precast prestressed slab strip. All segments, with a maximum weight of 49 t (43 mt), were cast in a long bench and placed with a tower crane traveling between the box girders in the approaches. Segments were barged in for the main span, and a beam and winch system was used for hoisting them into place, Figure 3.122. 3.18.2 VAIL PASS BRIDGES, U.S.A.

FIGURE 3.122. Calix Viaduct, placing precast segments in superstructure.

These bridges are located on Interstate I-70 over Vail Pass near Vail, Colorado, in a beautiful setting at an altitude between 9000 and 10,000 ft (2700 and 3000 m) above sea level where winter conditions are critical and the construction period is very short. Dimensions are shown in Figure 3.123, and a view of one finished bridge appears in Figure 3.124. 3.18.3 TRENT VIADUCT, U.K.

Section near midspan

FIGURE 3.123. Vail Pass Bridge, cross-section general dimensions.

This structure carries the M-180 South Humberside motorway over the River Trent and consists of dual roadways of three lanes each, with a central median. Precast segmental construction was selected against a steel plate girder design with a reinforced concrete deck slab. The bridge is sym-


141

FIGURE 3.124. \‘A Pass bridge, a completed precast segmental structure (courtesy of International Engineering Company, Inc.). F I G U R E 3 . 1 2 6 . I‘rellt Bridge, finishing the deck.

metrical with four spans of 159, 279, 279, and 159 ft (48.5, 85, 85, and 48.5 m). Each roadway is supported by an independent superstructure of twin concrete box girders varying in depth from 16 ft (4.9 m) at the piers to 7 .ft (2.1 m) at midspan of the center spans. Principal dimensions are shown in Figure 3.125. Each box girder is made up of 91 precast segments 10 ft (3 m) long, varying in weight between 38 t (35 mt) to 82 t (75 mt). All segments were placed in balanced cantilever with a launching gantry shown in operation in Figure 3.126, with precast units being delivered on the finished deck. 3.18.4

L-32 TAUER,~AUTOBAHN

l a u n c h i n g ganrry

ft (33.5, twenty at 55, and 33.5 m). Box piers have a maximum height of 330 ft (100 m). The constantdepth superstructure of 12.5 ft (3.8 m) is made up of 722 segments match-cast in a job-site factory equipped with four casting machines, Figure 3.127. A launching gantry was used to place all segments in the two bridges in balanced cantilever, Figure 3.128. 3.18.5 KISHWAUKEE RIVER BRIDGE, U.S.A.

This dual structure carries U.S. Route 51 over the Kishwaukee River near the city of Rockford, Illinois. Dimensions are shown in Figure 3.129. Prestressing is achieved in the transverse and longitudinal directions by bar tendons. All segments were placed in the structure by a launching gantry, shown in Figure 3.130, which represents the first application of this method in the United States.

BRIDGE, AUSTRIA

This structure is located between Salzburg and Villach, Austria, as part of a new motorway connecting Germany and Yugoslavia. The 22-span twin bridge has a total length of 3820 ft (1167 m) distributed as follows: 110, twenty at 180, and 110 17.400

MOTORWAY CENTRAL L RESERVE 4 -

INSITU

JOINT \

t

WEST

I-

I-

EAST

NAVIGATION CNANNEL

Elevation

FIGURE 3.125. Trent Bridge, typical dimensions.

FIGURE 3.127. L-32 Tauernautobahn Bridge, casting machine.

FIGURE 3.128. L-32 launching gantry.

Tauernautobahn

Bridge,

3.18.6 KENTUCKY RIVER BRIDGE, U.S.A.

3.18.7 I-205 COLUMBIA RIVER BRIDGE, U.S.A.

This structure crossing the Kentucky River is located in Franklin County just south of Frankfort, Kentucky. It is a three-span structure with a 323 ft (98.5 m) center span and 228.5 ft (70 m) side spans. In cross section the superstructure consists of two rectangular boxes. It is,the first precast segmental bridge to be constructed in the United States using the long-bed casting method, Figure 3.131. A view during construction is shown in Figure 3.132.

This large project represents one of the major applications of precast segmental construction in the United States. The 5770 ft (1759 m) long structure carries Interstate I-205 from Vancouver, Washington, across the North Channel of the Columbia River to Government Island near Portland, Oregon. Twin structures carry two 68 ft (20.7 m) wide roadways with span lengths varying between 600 ft (183 m) and 242 ft (74 m). Typical dimensions of

170’-0’ I_--

ELEV 694.0

-TRANSVERSE POST-TENSIONIN

21

0 -!k-Ao.‘-L+fb)

._

fcJ

FIGURE 3.129. Kishwaukee River Bridge, superstructure elevation and cross sections. (a) Elevation. (b) Section at midspan. (c) Section at pier. (From ref. 18.)


143

FIGURE 3.131. Kentucky River Bridge, long-line casting bed.

FIGURE 3.130. Klrhwaukte River Bridge, v~elv during construction showmg launching truss. the main spans over the river are shown in Figure

3.133. Dimensions of the cross section, as designed, are shown in Figure 3.134. However, the contractor, under a value engineering option in the contract documents (see Chapter 12), elected to redesign the cross section to a two-cell box section, Figure 3.135. The contractor exercised the op-

FIGURE

3.133.

FIGURE struction.

3.132.

kcntut

k\ Rncxt 131 idgc. (In1 111% con-

tion allowed in the bidding documents to select his own construction method and proceeded with casting in place in conventional travelers the two cantilevers adjacent to the main navigation channel (piers 12 and 13), while all other spans are of precast segmental construction. Figure 3.136 shows a rendering of the structure.

I-205 Columbia River Bridge, elevation and plan.

144

.

Precast Balanced Cantilever Girder Bridges ll'-10" I

67'-10"

1 I

67'-11"

I

I

I

FIGURE 3.134. I-205 Columbia River Bridge, cross sections.

CROSS

FIGURE

3.135.

SECTION

OF

PRECAST

SEGMENTS

!

372’ ,

I-205 Columbia River Bridge, revised

cross section. 3.18.8

ZILWAUKEE BRIDGE, U.S.A.

This bridge is another important example of precast segmental construction in the United States. Located in central Michigan, this 8080 ft (2463 m) long structure carries dual four-lane roadways over the Saginaw River near Zilwaukee, Michigan. Principal dimensions are shown in Figure 3.137.

366’i

FIGURE

FIGURE 3.136. I-205 Columbia Rner Bridge, ren-

dering of the structure.

389’

3.137.

377’ !

392’

368’ j

372’ 1351’

Zilwaukee Bridge, typical dimensions.

The 5 1 spans vary m length from 155 ft to 392 ft (47 to 119 m). An additional three-span ramp carries some traffic onto the southbound high-level bridge. Navigation clearance is 125 ft (38 m) above the Saginaw River. For a total deck area of 1,180,OOO sq ft (110,000


I

11.70

FIGURE

3.138.

Ottmarsheim

Bridge,

145

general

dimensions.

m*) t h e r e a r e 1 5 9 0 l a r g e s e g m e n t s v a r y i n g i n length from 8 to 12 ft (2.4 to 3.65 m) with a maximum weight of 160 t (144 mt). Segments were produced in a production-line operation with short-line casting and placed in the structure in balanced cantilever with a large launching gantry accommodating two successive spans.

weighing a maximum of 50 t (45 mt) are designed to be placed in balanced cantilever with an auxiliary overhead truss (and winch system) in the approach spans to stabilize the deck over the flexible piers during construction.

3.18.9 OTTMARSHEIM BRIDGE, FRANCE

This very important project is a recent application of precast segmental construction to urban elevated structures. The constraints relating to location of piers and construction over highway and railway traffic are comparable to the conditions encountered at the B-3 South Viaducts in Paris, France. The principal project dimensions are shown in Figure 3.142. All segments will be placed in the twin bridge using two launching gantries, which incorporate the latest technological developments in safety and efficiency.

This bridge in East France close to Germany and the Rhine River at the Ottmarsheim hydroelectric plant is today the longest clear span of precast segmental construction and the first major application of lightweight concrete to this type of structure. Principal dimensions are shown in Figure 3.138. As shown in the longitudinal section, lightweight concrete was used only in the center portion of the two main spans over the navigable waterway and over the outlet channel of the power plant. Figure 3.139 is a view of the completed structure.

3.18.11 F-9 FREEWAY, MELBOURNE, AUSTRALIA

3.18.10 OVERSTREET BRIDGE, FLORIDA, U.S.4.

This structure crosses the lntracoastal Waterway near Panama City in Western Florida. Dimensions are shown in Figures 3.140 and 3.141. The main navigation span is 290 ft (88 mm) long between piers to avoid any construction in the water fender system during operation. Approach spans are 125 ft (38 m) long and rest on I-shiped piers bearing on precast piles. The main piers consist of twin I piers of the same design. The total length of structure is 2650 ft (808 m) divided as follows: 95, seven at 125, 207.5, 290, 207.5, seven at 125, and 95 ft (29, seven at 38, 63, 88, 63, seven at 38, and 29 m). Precast segments 10 ft (3 m) long and

FIGURE 3.139. Ottmarsheim Bl-idge,

completed

structure.

vic\v of’ the

2650’-0”

Overall Length of Bridge 21t

207’b!i”

;125,-O&l

25’-0225’.0’2

25’-0’~25’-0~125-0’~125-0’~i’-0r;l 1

Sand Cement 3 : Riprap. (Typ.1

FIGURE 3.140. Overstreet Bridge, blot-da,

elevation

L2ig”

t,

LOLO”

I

al

FIGURE 3.141. Overstreet Bridge, Florida, cross sections.

2’-6” -4-

References

References 1. Jean Muller, “Ten Years of Experience in Precast Segmental Construction,” Journal of the Prestressed Concrete Institute, Vol. 20, No. 1, January-February 1975. 2. C. A. Ballinger, W. Podolny, Jr., and M. J. Abrahams, “A Report on the Design and Construction of Segmental Prestressed Concrete Bridges in Western Europe- 1977,” International Road Federation, Washington, D.C., June 1978. (Also available from Federal Highway Administration, Office of Research and Development, Washington, D.C., Report No. FHWA-RD-78-44.) 3. Walter Podolny, Jr., “An Overview of Precast Prestressed Segmental Bridges,” Journal of the Prestressed Concrete Institue, Vol. 24, No. 1, January-February 1979. 4. J. Mathivat, “Reconstruction du Pont de Choisy-leRoi,” Travaux, Janvier 1966, No. 372. 5. Jean Muller, “Long-Span Precast Prestressed Concrete Bridges Built in Cantilever,” First International Symposium, Concrete Bridge Design, Paper SP 23-40, AC1 Publication SP-23, American Concrete Institute, Detroit, 1969. 6. Andre Bouchet, “Les Ponts en Beton Precontraint de Courbevoie et de la Grande-Jatte (Hauts-deSeine),” La Technique des T r a v a w , Juillet-Aout 1968. 7. “Bear River Bridge,” STUP Bulletin of Information, November-December 1972. 8. “Nova Scotia’s Bear River Bridge-Precast Segmental Construction Costs Less and the Money Stays at Home,” Bridge Bulletin, Third Quarter 1972, Prestressed Concrete Institute, Chicago. 9. “John F. Kennedy Memorial Causeway, Corpus Christi, Texas,” Bridge Report SR 162.01 E, Portland Cement Association, Skokie, Ill., 1974. 10. G. C. Lacey, and J. E. Breen, “Long Span Pre-

147

stressed Concrete Bridges of Segmental Construction State of the Art,” Research Report 12 l-l, Center for Highway Research, The University of Texas at Austin, May 1969. 1. S. Kashima and J. E. Breen, “Epoxy Resins for Jointing Segmentally Constructed Prestressed Concrete Bridges,” Research Report 121-2, Center for Highway Research, The University of Texas at Austin, August 1974. 2. G. C. Lacey and J. E. Breen, “The Design and Optimization of Segmentally Precast Prestressed Box Girder Bridges,” Research Report 121-3, Center for Highway Research, The University of Texas at Austin, August 1975. 13. R. C. Brown, Jr., N. H. Burns, and J. E. Breen, “Computer Analysis of Segmentally Erected Precast Prestressed Box Girder Bridges,” Research Report 121-4, Center for Highway Research, The University of Texas at Austin, November 1974. 14. S. Kashima and J. E. Breen, “Construction and Load Tests of a Segmental Precast Box Girder Bridge Model,” Research Report 121-5, Center for Highway Research, The University of Texas at Austin, February 1975. 15. J. E. Breen, R. L. Cooper, and T. M. Gallaway, “Minimizing Construction Problems in Segmentally Precast Box Girder Bridges,” Research Report 121-6F, Center for Highway Research, The University of Texas at Austin, August 1975. 16. Ben C. Gerwick, Jr., “Bridge over the Eastern Scheldt,” Journal of the Prestressed Concrete Institute, Vol. 11, No. 1, February 1966. 17. “A Proud Achievement-The Captain Cook Bridge,” Issued by the Commissioner of Main Roads-1972, Main Roads Department, Brisbane, Queensland, Australia. 18. “Prestressed Concrete Segmental Bridges on FA 412 over the Kishwaukee River,” Bridge Bulktin, No. 1, 1976, Prestressed Concrete Institute, Chicago.

4 Design of Segmental

4.1 4.2 4.3 4.4

4.5

4.6 4.7 4.8

INTRODUCTION LIVE LOAD REQUIREMENT?3 SPAN ARRANGEMENT AND RELATED PRINCIPLES OF CONSTRUCTION DECK EXPANSION, HINGES AND CO4.4.1 Hinges at Midspan 4.4.2 Continuous Su~tructures 4.4.3 Expansion of Long Bridge TYPF, SHAPE AND DIMENSIONS OF THE SUPERsTRu4.5.1 Box Sections 4.5.2 Sbape of Superst~cture in Elevation 4.5.3 Choice of Typical Cross Section 4.5.4 Dimensions of the Typical Cuss Section TRANSVERSE DISI’RIBUI’ION OF LQADS BETWEEN BOX GIRDERS IN MULTIBOX GIRDERS EFFECT OF TEMPFXATI-JRF, GRADIENTS IN BRIDGE suPFRsl-RucrUREs DESIGN OF LONGITUDINAL MEMBERS FOR FLEXURE AND TENDON PROFILES 4.8.1 Principle of Pre&ess Iayout 4.8.2 Draped Tendons 4.8.3 Shaight Tendons 4.8.4 Summary of Tendon Profiles and Anchor Locations 4.8.5 Special Problems of Continuity PresWss and Ancbonge Thereof 4.8.6 Iayout of Pmskess in Strucaups with Hinges and Expansion Joints 4.8.7 Redistribution of Moments and Stresses Through concrete creep

4.1

Introduction

Design of concrete highway bridges in the United States conforms to the provisions of The American Association for State Highway and Transportation Officials (AASHTO) “Standard Specifications for Highway Bridges.” For railway structures, specifications of the American Railway Engineers Association (AREA) should be consulted. For the 148

4.9 4.10

Bridges

4.8.8 Prediction of Preskess Losses ULTIMATE BENDING CAPACITY OF LONGITUDINAL MEMBERS SHEAR AND DESIGN OF GROSS SECITON

4.10.1 Introduction 4.10.2 Shear Tests of Reinforced Concx~te Beams 4.103 DifIiculties in Actual Structmw 4.10.4 Design of h@dinal Members for Shear 4.11 JOINTS BETWFEN MATCH-CAST SEGMENTS 4.12 DESIGN OF SUPERSTRUCl-URE CROSS SECl’ION 4.13 SPECIAL PROBLEMS IN SUPmUCIWRE DESIGN 4.13.1 Diapluagms 4.13.2 Superstructure over Piers 4.13.3 End Abutments 4.13.4 Expansion Joint and Hinge Segment 4.14 DEFLECITONS O F C A N T I L E V E R B R I D G E S A N D CAMBER DESIGN 4.15 FATIGUE IN SEGMENTAL BRIDGES 4.16 PROVISIONS FOR FUTURE PmING 4.17 DEhGN FXAMPLE 4.17.1 Longitudinal Beding 4.17.2 Redktribution of Moments 4.17.3 Stresses at Midspan 4.17.4 shear 4.17.5 Design of the Cross-Section Frame 4.18 QUANTITIES OF MATERIALS 4.19 POTENTL4L PROBLEM ARF.AS REFERENCES

most part, the provisions in these specifications were written before segmental construction was considered feasible or practical in the United States. Before discussing design considerations, the authors wish to emphasize that no preference for either cast-in-place or precast methods of construction is implied here. The intent is simply to present conditions that the designer should be

of

Span Arrangement and Related Principles aware of to produce a satisfactory design. Both concepts are viable ones, and both have been used to produce successful structures. In general, the segmental technique is closely related to the method of construction and the structural system employed. This is why segmental construction, either cast in place or precast, has been often identified with the cantilever construction used in so many applications. It is logical to take bridge structures built in cantilever as a basis for the design considerations developed in this chapter. Where other methods, such as incremental launching or progressive placement, require special design considerations, such problems are discussed in the appropriate chapters.

4.2 Live-Load Requirements In comparing practices in other countries to those in the United States, an important parameter to keep in mind is that of live-load requirements. Figure 4.1 illustrates the considerable differences among code requirements in various countries.’ For a simple span of 164 ft (50 m) and width of 24.6 ft (7.5 m), the German specification requires a live-load design moment 186% greater and the French requires one 290% greater than that of AASHTO. Some Canadian provinces use the AASHTO specifications but arbitrarily increase the live load by 25%.

4.3

Span Arrangement and Related Principles of Construction

In the balanced cantilever type of construction, segments are placed in a symmetrical fashion about a pier. The designer must always remember that construction proceeds with symmetrical cantilever deck sections centered about the piers and not with completed spans between successive piers.2 For a typical three-span structure, the side spans should preferably be 65 percent of the main center span instead of 80 percent in conventional castin-place structures. This is done to reduce to a minimum the length of the deck portion next to the abutment, which cannot be conveniently built in balanced cantilever, Figure 4.2~. Where span lengths must vary, as between a main span and an approach span, it is best to introduce an intermediate span whose length will average the two flanking spans, Figure 4.26. In this manner the cantilever concept is optimized. Individual cantilever sections are generally made continuous by insertion of positive-moment ten-

A A S H T O IRC 50 100 138 100 loo 138

M km)

CPC

290 177

f

5000 P

t

A

//

4 0 0 0 l- q M a x . M

8

I

0

France-

DIN l07i

10

20

30

40

/

/

/

/

50

149

The depth-to-span and width-to-depth ratios for segmental construction presently advocated in the United States have been adopted from European practice. The lighter live loads used in the United States should permit further refinements in our design approach.

AASHTO 100% DIN 1072 186 173

P(m)

Construction

60

70

80

90

loo

ah)

Span

FIGURE 4.1. Maximum live-load moment (simple span) (F. Leonhardt, New Practice in Concrete Structures, IABSE, New York, 1968).

Design of Segmental Bridges

0.65-07OL

065-O

70L

(a)

LI

,

I ‘ 2 (LITL2)

L2

..I

(b)

Section A-A

FIGURE 4.4. End restraint at abutment. fc) FIGURE 4.2. Cantilever construction showing choice of span lengths and location of expansion joints. dons upon closure. It is preferred not to have any permanent hinges at midspan. Continuous decks without joints have been repeatedly constructed to lengths in excess of 2000 ft (600 m) and have proved satisfactory from the standpoint of maintenance and riding quality. For very long viaduct-type structures, intermediate expansion joints are inevitable to accommodate volume changes. These joints should be located near points of contraflexure, Figure 4.2c, to avoid objectionable slope changes that occur if the joint is located at midspan. This consideration will be discussed in more detail in Section 4.4. In many cases it may not be possible to provide the desirable optimum span arrangement. Thus, the end span may be greater or less than the optimum span length desired.2 In the case of a long end span, the superstructure might be extended over the abutment wall to provide a short additional span. As shown in Figure 4.3, a conventional

bearing (1) is provided over the front abutment wall. A rear prestressed tie (2) opposes uplift and permits cantilever construction to proceed outward from the abutment to the joint t’Jl), where a connection can be effected with the cantilever from the first intermediate pier. Figure 4.4 shows an alternative scheme with a constant-depth section, as opposed to a haunched section, where the deck has been encased within the abutment wing walls for architectural purposes. For the normal end span, a special segment is temporarily cantilevered out so as to reach the first balanced cantilever constructed from the next pier, Figure 4.5. Alternatively this portion could be cast in place on falsework, if site conditions permit. In a short-end-span situation, cantilever construction starts from the first pier and reaches the abutment on one side well before the midspan section of the adjacent span, Figure 4.6. An uplift reaction must be transferred to the abutment during construction and in the completed structure. Consequently, the webs of the main box girder deck are cantilevered over the expansion

FIGURE 4.3. End restraint in abutment.

Deck Expansion, Hinges and Continuity

151

FIGURE 4.5. Conventional bearing on abutment.

FIGURE 4.6. Anchorage for uplift in abutment. joint into slots provided in the main abutment wall, Figure 4.7. The neoprene bearings are placed above the web cantilever rather than below to transfer the uplift force while allowing the deck to expand f-reely. Interesting examples of such concepts are given in the three following bridges: Givors Bridge over the Rhone River, France, shown in Figure 4.8. The main dimensions are given with the typical construction stages of the superstructure. Tricastin Bridge over the Rhone River, France (Section 2.15.11). No river piers were desired for the structure, which dictated a main span of 467 ft (142.50 m), and there was no room on the banks to increase the side spans so as to avoid the end uplift. Two very short side spans of only 83 ft (25.20 m) provide the end restraint of the river span. The uplift is transferred to the abutments, which are earth tilled to provide a counterweight, Figure 4.9. The magnitude of the uplift force has been re-

duced by the use of lightweight concrete in the center of the main span. Puteaux Bridges over the Seine River, near Paris (Section 2.15.10). A few bridges have even been built in cantilever entirely from the abutments. The Reallon Bridge in Frarice is one such structure, Figure 4.10, where very special site conditions with regard to bridge profile and shape of the valley were best met with this concept. Another set of circumstances may be encountered when it is not possible to select the desired span lengths to optimize the use of cantilever construction. Such was the situation of the bridge over the Seine River for the Paris Ring Road, where a side span on the left bank could not be less than 88 percent of the main river span over the river, while very stringent traffic requirements governed the placement pattern of precast segments on the right bank, Figure 4.11.

4.4

Deck Expansion, Hinges and Continuity 4.4.1 HINGES AT MIDSPAN

Prestressing

FIGURE 4.7.

Longitudinal section.

units

Historically, the first prestressed concrete bridges built in cantilever were provided with a hinge at the center of the various spans. Such hinges were designed to transfer vertical shear between the tips of two adjacent cantilever arms (which could develop under the live loading applied over one arm only in half the span length) while enduring a free expansion of the concrete deck under volume changes (concrete creep and seasonal variations of temperature). Continuity of the deflection curve

RIM CiMCHE 6

s

FIGURE 4.8.

152

3

2

Givors Bridge over the Rhone River, France, span dimensions and typical construction stages. (1) Construction of left bank river pier segment. The eight segments either side of the pier are erected, and pier stability is assured by temporary props. (2) The connection between deck and abutments is made. Temporary props are removed and the seven remaining segments are placed in cantilever. (3) The above operation is repeated on the right bank. The central pier segments are poured. Two segments are erected on either side of each pier, supported by scaffolding. (4) The last segment is placed in the central span, continuity is achieved between the two cantilevers, and the scaffolding is removed. (5) The remaining 16 segments on either side of the central piers are placed. (6) The 110 m spans are completed by pouring the closure segments and tensioning the continuity prestress. The superstructure is now complete.

1

RP.‘E

CRXTE

Elevation

Section

A-A

’

’

I

I Plan

FIGURE 4.9.

Tricastin Bridge over the Rhone River, France.

FIGURE 4.10. Reallon Bridge, France.

PHASE

1

construction of central cantilever 1

2 n

PHASE 2

construction of right bank cantilever 1 3

PHASE 3

*

\

1 e

2 1 &

3 -1

j

c

closure of central and right bank cantilever

\

154

&z---4

1

2

!

il d

f

3 n ”

5

PHASE 4

joining of right bank cantilever with abutment

PHASE 5

construction of left bank cantilever

PHASE 6

closure of left bank and

PHASE 7

joining of left bank cantilever with abutment

central

cantilever


155

I

Cc) FIGURE 4.11. Paris Belt (Downstream). (0) Typical construction stages. (b) Segment assembly-right bank. (c) Segment assembly-left bank.

was thus obtained in terms of vertical displacement but not insofar as rotation at the hinge point was concerned. Remember that in this type of structure the deck is necessarily fixed at the various piers, which must be designed to carry the unbalanced moments due to unsymmetrical live-load patterns over the deck. On the other hand, these structures are simple to design because they are statically determinate for all dead loads and prestressing, and the effect of live load is simple to compute. Because there are no moment reversals in the deck, the prestressing tendon layout is simple. Some disadvantages were accepted as the price of simplicity of design: The deck has a lower ultimate capacity as compared with a continuous structure, because there is no possible redistribution of moments. Hinges are difficult to design, install, and operate satisfactorily. There are many expansion joints, and regardless of precautions taken in design, construction, and operation they are always a source of difficulty and high maintenance cost. The major disadvantage, revealed only by experience, related to the exceeding sensitivity of such structures to steel relaxation and concrete creep.

Because of the various hinges at midpoints of the spans, there is no restraint to the vertical and angular displacements of the cantilever due to the effect of creep. Steel relaxation and the corresponding prestress losses tend to make matters worse, while concrete creep is responsible for a progressive lowering of the center of each span. With time, there is an increasing angle break in the deck profile at the hinge. The magnitude of the deflection has been reported to be in excess of one foot (0.03 m). The difficulties experienced with this type of construction are such that most government officials in Western Europe will no longer permit its use.3 4.4.2 CONTINUOUS SUPERSTRUCTURES

Further research concerning the exact properties and behavior of materials for such structures having a midspan hinge would enable more accurate prediction of the expected deflection and thus better control. A far more positive approach is to eliminate the fundamental cause of the phenomenon by avoiding all permanent hinges and achieving full continuity whenever possible. To show the relative behavior of a continuous structure and one with hinges at midspan, a numerical application was made for the center span of the Choisy-le-Roi Bridge in two extreme cases:

156

Design TABLE

4.1.

of Segmental Bridges

Comparison of Crown Deflections (Hinged versus Continuous Structure) Cast-in-Place Hinged Structure E

( lo6 psi)

No. Load Stage 1 2 3 4 5 6 7 8 9 10

Girder weight Initial prestress Cumulative 5% Deviation of prestress Continuity prestress Superimposed load Finished structure (initial) Concretecreep and losses Finished structure (final) Live loads

4.3 4.3 4.3 6.4 2.1 6.4

?

(in.) 1.80 - 1.50 0.30 0.30 0.60 1.10 1.70 0.90

Precast Continuous Structure 0

E

(in. X 103/in.)

( lo6 psi)

(in.)

2.4 -2.0 0.4

5.1 5.1 5.1 6.4 6.4

1.50 -0.90 0.60

23% 0.4 0.8 1.4 2.2 1.1

6J

?

-0.30 0.10 0.40 -0.10 0.30 0.30

2.1 6.4

(in.

X

103/in.)

2.0 - 1.2 0.8 7%

0 0 0.8 0 0.8 0

Explication of symbols: E = modulus of elasticity for each particular loading stage y = vertical deflection at crown o = total angular break at crown (expressed in thousandths of inch per inch) Derivation of results: girder weight and initial prestress (3) = (1) + (2) (7) = (3) + (5) + (6) finished structure (initial stage) finished su-ucture (final stage) (9) = (7) + (8)

Cast-in-place cantilever with a hinge at midspan, and Precast

segmental

continuous

construction.

Results comparing the two structures are shown in Table 4.1 and in Figures 4.12 through 4.14. The study shows no significant difference between the two types of structures with respect to the theoretical behavior of the cantilever method under combined dead load and initial prestress, Figure 4.12. In fact, the angle change at midspan is even slightly less for the hinged structure, because

the prestress offsets a greater percentage of dead-load moments, 83 percent instead of 58 percent.

f CIilCREl;t

; E d

ClEEP

11.5

I

CISI II race nim swuciure

ClnIInw

rrecw Slrrcw

FIGURE 4.13. Comparison of deflection caused by creep (hinged versus continuous structure).

3

=

= ,=

I.0

LIYE’ LILI

-

2’ CISI II me nln181

struclnrc

cloIIoIIII I

rrecna

Wrrlrre

FIGURE 4.12. Comparison of deflection under dead load and prestressing (hinged versus continuous structure).

E =

m 11 rince llnltd Strwri

I

Clntlnrl~:

rrecul ltrrclrre

FIGURE 4.14. Comparison of deflections caused by live load (hinged versus continuous structure).


157

7% in the continuous structure. Therefore, the continuous structure is three times less sensitive to possible deviations from the assumed material properties. Live-load deflections of the continuous structure are three times more rigid than the hinged structure, Figure 4.14. The deflection of a typical span of the Oleron Viaduct in France is compared with a continuous span and with a crown hinged span in Figure 4.15. From these data it is obvious that the fullest use of continuity and the elimination of hinges at midspan whenever possible is beneficial to the structural behavior of the bridge, to safety and comfort of traffic, and to the structure’s aesthetic appearance. In practice, the continuity of the individual cantilever arms at midspan is obtained by another set of prestressing tendons, usually called continuity

When the effect of concrete creep is considered, however, there is a significant difference between the two types of structures, Figure 4.13. The hinged structure has a vertical deflection of 1.1 in. (28 mm) and a corresponding total angle break of 0.0028 in./inch. This value is twice that shown in Table 4.1 and Figure 4.13 for the angle change of one cantilever, the value of 2.8 being the total angle break of the two abutting cantilevers. The continuous structure indicates a camber of 0.1 in. (3 mm), and no angle break will ever appear because of full continuity. Further, the effect of deviation of actual prestress load from the design prestress load points out an important difference in the sensitivity of the two systems. Assuming the actual prestress in the structure to differ from the design assumption by 5%, the corresponding maximum deflection is increased by 23% in the hinged structure but only

\

I

/’ \(

’

0 , 6 I. % Y

I

2 6 0 I-L

Comparisons between live-load deflections for continuous or hinged structures.

FIGURE 4.15.

158


prestressing, which is installed along the span in a continuous structure. Details of the design aspects of this prestress will be discussed in Section 4.8. 4.4.3 EXPAMSIO,V

OF LOAVG

BRIDGES

When the continuity of the superstructure is selected as optimum for the behavior of the structure, one must keep in mind that proper measures should be concurrently taken to allow for expansion due either to short-term and cyclic volume changes or to long-term concrete creep. The piers may be made flexible enough to allow for such expansion or may be provided with elastomeric bearings to reduce the magnitude of horizontal loads to acceptable levels when applied to the substructure. This important aspect of the overall bridge design concept is considered in Chapter 5. Several structures are currently made continuous in lengths of 1000 to 2000 ft (300 to 600 m) and in exceptional cases even 3000 ft (900 m). For longer structures, full continuity between end abutments is not possible because of the excessive magnitude of the horizontal movements between superstructure and piers and related problems. Therefore, intermediate expansion joints must be provided. For long spans they should not be placed at the center of the span, as in the early cantilever bridges, but closer to the contraflexure point to minimize the effect of a long-term deflection. Such a concept was developed initially for the Oleron Viaduct and is currently used on large structures such as the Saint Cloud Bridge in Paris, Sallingsund Bridge in Denmark, and the Columbia River and Zilwaukee Bridges in the United States. Detailed computations were made in the case of the Oleron Viaduct to optimize the location of the expansion joint in a typical 260 ft (80 m) span, Figure 4.15 shows the shape of the deflection curve for a uniform live loading with the three following assumptions:

Maximum deflection under live load is reduced in the ratio of 2.2 to 1. Maximum angle break under live load is reduced in the ratio of 3.0 to 1. For dead-load deflections the difference is even more significant, such that there is no substantial difference between the actual structure and a fully continuous one. The variation of the angle break at the hinge point versus the hinge location along the span length is shown in Figure 4.16. There seems to be little doubt that the structure is improved by selection of a proper location for the hinge and the expansion joint. Theoretically, the ideal hinge position is between points ,4 and B, which are the contraflexure points for dead and live loads. From a construction standpoint, such a location f-or the hinge complicates the erection process, for the hinge must be temporarilv blocked and subsequentlv released when the span is complete and continuitv is achieved. We will consider this subject in detail after examining the layout of longitudinal prestress in cantilever bridges (Section 4.8.6). It was recently discovered, in the designing of the Sallingsund Bridge, that the optimum location

Fully continuous span Span with a center hinge Span with an intermediate hinge located at 29 percent of the span length from the adjacent pier (actual case) The advantages of having moved the hinge away from the center toward the quarter-span point are obvious:

LOCATION OF HINGE BETWEEN MID- SPAN AND

PIER

FIGURE 4.16. Variation of angle break at the hinge with hinge location along the span.

159

Type, Shape, and Dimensions of the Superstructure

of the hinge to control the deflections under service-load conditions does not simultaneously permit achievement of the overall maximum capacity under ultimate conditions. This question will be discussed later in this chapter. The preceding discussion of hinge location applies particularly for very long spans or for slender structures. For moderate spans with sufficient girder depth it has been found that careful detailing of the prestress in the hinged span can allow the hinge to be maintained at the centerpoint for simplicity (spans less than 200 ft with a depth to span ratio of approximately 20). Such was the case for the cantilever alternatives of the Long Key and Seven Mile Bridges in Florida. 4.5

Type, Shape, and Dimensions of the Superstructure 4.5.1 BOX SECTIONS

The typical section best suited for cantilever construction is the box section, for the following reasons: 1. Because of the construction method, deadload moments produce compression stresses at the bottom fiber along the entire span length, and maximum moments occur near the piers. The typical section therefore must be provided with a large bottom flange, particularly near the piers, and this is achieved best with a box section. The efficiency of the box section is very good, and for a given amount of concrete provides the

(0) Longitudinal section

least amount of prestressing steel. The efficiency of a section is usually measured by the following dimensionless coefficient: r2

p=C&2

with the notations as given in Figures 4.17 and 4.18, where some basic formulas are presented. The efficiency would be p = 1 if the concrete were concentrated in thin flanges with webs of negligible thickness. On the other hand, a rectangular section has an efficiency of only l/3. The usual box section efficiency is p = 0.60, which is significantly better than that of an I girder. 2. Another advantage of the large bottom flange is that the concrete area is sufficiently large at ultimate load to balance the full capacity of the prestressing tendons without loss in the magnitude of the lever arm. ’ 3. The elastic stability of the structure is excellent both during construction and under service conditions, because the closed box section has a large torsional rigidity. 4. In wide bridge decks where several girders must be used side by side, the large torsional stiffness of the individual box girders allows a very satisfactory transverse distribution of live loads without intermediate diaphragms between piers. 5. Because of their torsional rigidity, box girders lend themselves to the construction of curved bridge and provide superstructures maximum flexibility for complicated tendon trajectories.

6) Typical tramverSe

section

FIGURE 4.17. Typical characteristics of a box section: Total section height: h; crosssection area: A; moment of inertia: I; position of centroid; c,, c2; radius of gyration: r given by rp = Z/A; efficiency ratio: p = r%,c,; limits of central core: r*/c, = PC,; r%, = pc2; for the usual box girder: p = 0.60.


160

h

dl

px ,c2/ I

-F

Cl

Y

(a)

(b) The optimum selection of the proportions of the box section is generally a matter of experience. A careful review of existing bridges provides an excellent basis for preliminary design. The various parameters that should be considered at the start of a design are: Constant versus variable depth Span-to-depth

ratio

Number of parallel box girders Shape and dimensions of each box girder, including number of webs, vertical or inclined webs, thickness of webs, top and bottom flanges

h

Ph -

I Cc)

FIGURE 4.18. ‘rypical prestress requirements of a box girder. (u) For maximum negative moment over the pier (LX + LL): total moment = M; required prestress = F = M/z with z = c, - cf, + cp; usually over the piel- z = 0.75 12. (b) For maximum positive moment at midspan (LX + IL): total moment = ‘M; required prestress = F = M/i with z = cp - cf2 + c ,; usually at midspan z = 0.70h. (c) For variable moments (LL): total moment variation = A M (sum of positive and negative L.C. moments); required prestress = F = hM/ph (p = 0.60). All these factors are closely related to each other, and they also depend largely upon the construction requirements-for example, the size of the project that will require a large investment in sophisticated casting equipment. 4.5.2 SHAPE OF SUPERSTRUCTURE IiY ELEVz4TlOh’ Constant depth is the easiest choice and affords the best solution for short and moderate spans, up to 200 ft (60 m). However, constant depths have been used for aesthetic reasons for spans to 450 ft (140 m), such as the Saint Cloud Bridge in Paris and the

161

Type, Shape, and Dimensions of the Superstructure

Pine Valley and Columbia River Bridges in the United States, Figure 4.19~. When the span increases, the magnitude of dead-load moments near the piers normally requires a variation of structural height and a curved intrados. When clearance requirements allow, a circular intrados is the easier and more aesthetically pleasing choice, although in some cases (such as the Houston Ship Channel Bridge) a more complex profile must adjust to the critical corners of the clearance diagram. Between the constantdepth and the curved-intrados solutions, Figure 4.19, intermediate options may be used, such as: Increase thickness at pier , Ii?!’

A ,_ ,,,.,,

The semiconstant depth, where the concrete required in the bottom flange near the piers is placed outside the typical section rather than inside the box (constant dimension for the interior cell). This solution has been used on two bridges in France and is aesthetically satisfactory, Figure 4.196. Straight haunches (bridge for the Ring Road in Paris). In this case caution must be exercised to insure compatibility of the local stresses induced by the abrupt angle change of the bottom soffit at the start of the haunch, where a full diaphragm is usually needed inside the box, Figure 4.19~.

_.

l/15< h<1/30 optimum l/18 to l/20 ,,

-Yw“’

.““.

”

_.,

1116

1/16
II22
1/30
< l/50

Circular intrados or third-degree parabola

Cd)

FIGURE 4.19 Longitudinal profile for segmental bridges. (k) Constant depth.

(b) Semiconstant depth. (c) Straight haunches. (d) Variable depth.

162

Design of Segmental Bridges 4.53

CHOICE OF TYPICAL CROSS SECTION

Web spacing is usually selected between 15 and 25 ft (4.5 and 7.5 m) to reduce the number of webs to a minimum, simplifying construction problems while keeping transverse bending moment in the top and bottom flanges within reasonable limits. A superstructure up to 40 ft (12 m) in width is thus normally made up of a single cell box girder with two lateral cantilevers, the span of which is slightly less than one-fourth the total width (7 to 8 ft for a 40 ft width). For wide bridges, multicell box girders may be used: Three webs, two cells: as in the B-3 South Viaduct and the Deventer Bridge Four webs, three cells: as in the Saint Cloud Bridge and the Columbia River Bridge Alternatively, large lateral cantilevers and a large span length between webs are accepted with special provisions to carry the deck live loads transversely: Transverse flange stiffeners as in the Saint Andre de Cubzac, Vejle Fjord, and Zilwaukee Bridges Side boxes as in the Chillon Viaduct Alternatively several boxes may be used side by side to make up the superstructure. Figures 4.20 through 4.24 give the dimensions of a few structures selected at random from various countries throughout the world. 4.5.4 DIMENSIONS OF THE TYPICAL CROSS SECTION

Three conditions must be considered in determining the web thickness: Shear stresses due to shear load and torsional moments must be kept within allowable limits Concrete must be properly placed, particularly where draped tendons occur in the web Tendon anchors, when located in the web, must distribute properly the high prestress load concentrated at the anchorages Following are some guidelines for minimum web thicknesses: 8 in. (200 mm) when no prestress ducts are located in the web

10 in. (250 mm) when small ducts for either vertical or longitudinal post-tensioning tendons occur in the web 12 in. (300 mm) when ducts for tendons (twelve 3 in. diameter strands) occur in the web 14 in. (350 mm) when an anchor for a tendon (twelve 4 in. diameter strands) is anchored in the web proper Most codes underestimate the capacity of twoway slabs, such as the roadway slab or top flange of a box girder bridge, whether prestressed transversely or mild-steel reinforced. There is a great reserve of strength due to the frame action between slabs and webs in the transverse direction. The minimum slab thickness to prevent punching shear under a concentrated wheel load is approximately 6 in. (150 mm). However, it is recommended that a slab thickness of not less than 7 in. (175 mm) be used to allow enough flexibility in the layout of the reinforcing steel and prestressing ducts and obtain an adequate concrete cover over the steel and ducts. Recommended minimum top flange thickness versus the actual span length between webs should be: 7 in. (175 mm) Span less than 10 ft (3 m) 8 in. (200 mm) Span between 10 and 15 ft (3 to 4.5 m) 10 in. (250 mm) Span between 15 and 25 ft (4.5 to 7.5 m) Over 25 ft (7.5 m), it is usually more economical to substitute a system of ribs or a voided slab for a solid slab.

Early bridges used very thin bottom flanges in order to reduce critical weight and dead-load moments. A 5 in. (125 mm) thickness was used in bridges, such as the Koblenz Bridge in Germany. It is very difficult to prevent cracking of such thin slabs due to the combined effect of dead load carried between webs and longitudinal shear between web and bottom flange. For this reason, it is now recommended that a minimum thickness of 7 in.

FIGURE 4.20. Typical dimensions of some cast-in-

place segmental cantilever bridges in France. Year of construction and maximum span length (ft): (a) Moulin a Poudre (1963), 269. (6) Morlaix (1973), 269. (c) Bordeaux St. Jean (1965), 253. (d) Givors (1967), 360. (e) Oissel (1970), 328

(fl)

-I--=-+

(b)

(4

t s,

(e) 163

164


FIGURE 4.20 (Continzx~) (f) Viosne (1972), 197. (g) J o i n v i l l e (twin deck) (1976), 354. (h) Gennevilliers (1976), 564.

(175 mm) be used, regardless of the stress requirements. Where longitudinal ducts for prestress are distributed in the bottom flange, a minimum thickness of 8 to 10 in. (200 to 250 mm) is usually necessary, depending on the duct size. Near the piers, the bottom slab thickness is progressively increased to resist the compressive stresses due to longitudinal bending. In the Bendorf Bridge, 680 ft (207 m) span, the bottom flange thickness is 8 ft (2.4 m) at the main piers and is heavily reinforced to keep the compressive stresses within allowable limits. After this brief review of the various conceptual choices for dimensioning the deck members, con-

sideration should be given to the design of such members with particular emphasis on the following points: Distribution of load between box girders in multibox girder bridges Effect of temperature gradients in the structure

4.6

Transverse Distribution of Loads Between Box Girders in Multibox Girders

We noted earlier that wide decks can conveniently consist of two or even three separate boxes trans-

(1))

3 4.60

0

10.92

I 1

1 ’

I

! -t

5.50 g I 10.60

’

10.60

1 7

,

FIGURE 4.21. Typical dimensions of some precast segmental cantilever bridges in France. Year of construction and maximum span length (ft): (a) Choisy-le-Roi (1965), 180; (b) Courbevoie (1967), 197; (c) Oleron Viaduct (1966), 260; (d) Seudre (1971), 260; (e) B-3 South Viaduct (1973), 157; cf) St. Andre de Cubzac (1974), 312; (g) St. Cloud (1974), 334; (h) Ottmarsheim (1976), 564. 165

%I

900

,96,

%1 1

9cKl

(4

c_- 9 50

(e)

t

I

c

.-

-

(h) 166

196 1

Cc)

(4

(ft): (a) Koblenz, Germany (1954), cast in place, 374; (b) Bendorf, Germany (1964), cast in place, 682; (c) Chillon, Switzerland (1970), precast, 341; (d) Sallingsund, Denmark (1978), .precast, 305; (e) Vejle Fjord, Denmark (1979), cast in place, 361. 167

(b)

(c) FIGURE 4.23. Typical dimensions of some segmental cantilever bridges in Europe. Year of construction and maximum span length (ft): (a) Felsenau, Switzerland (1978), cast in place, 512; (6) Tarento, Italy (1977), cast in place, 500; (c) Kochertal, Germany (1979), cast in place, 453. 168

Transverse Distribution of Loads Between Box Girders in Multibox Typical Cross Section

!22'-6"! !22'-;q6, t

4

, r t

t

4

20'4

1 r

3a'-6"

4

,

17'-5"

,

'

38'-6"

T

, T

36' 59'-3"

1 r

4

m I 1

381

1

T

FIGURE 4.24. Typical dimensions of some segmental cantilever bridges in the Americas. Year of construction and maximum span length (ft): (n) Rio Niteroi, Brazil (1971), precast, 262: (h) Pine Valley, U.S.A. (1974), cast in place, 450; (c) Kipapa, U.S.A. (1977). cast in place, 250; (n) Kishwaukee, U.S.A., precast, 250; (e) Long Key, U.S.A., precast, 118;(r) Seven Mile, U.S.A., precast, 135; (y) Columbia River, U.S.A., cast in place and precast, 600: (h) Zilwaukee, U.S.A., precast, 375; (i) Houston Ship Channel, U.S.A., cast in place, 750.

versely connected by the top flange. A detailed analysis was made of such decks with regard to the distribution of live load between the various boxes. It was found that in normal structures of this type, the combined effect of the flexural rigidity of the roadway slab acting transversely as a rigid frame with the webs and bottom slab of the various box

Girders

169

girders, on one hand, and the torsional rigidity of such box girders on the other hand, would result in a very satisfactory transverse distribution of live loads between box girders. There is no need for diaphragms between girders as normally provided for I-girder bridgers. Comprehensive programs of load testing of several bridges, including accurate measurements of deflections for eccentric loading, fully confirmed the results of theoretical analysis. This analysis has been reported in various technical documents, and only selected results will be presented in this section. The first bridge analyzed in this respect was the Choisy-le-Roi Bridge. A knife-edge load P is considered with a uniform longitudinal distribution along the span, Figure 4.25. When this load travels crosswise from curb to curb, each position may be analyzed with respect to the proportion of vertical load carried by each box girder, together with the corresponding torsional moment and transverse moment in the deck slab. These analyses have made it possible to draw transverse influence lines for each effect considered, such as longitudinal bending m o m e n t s ( o v e r t h e s u p p o r t o r a t midspan), torsional moments, or transverse moments. For longitudinal moments it is convenient to use a dimensionless coefficient, Figure 4.25c, which represents the increase or decrease of the load carried by one box girder in comparison with the average load, assuming an even distribution between both girders. Numerical results show that the transverse distribution of a knife-edge load placed on one side (next to the curb) of a twin box girder produces bending moments in each box that are 1.4 and 0.6 times the average bending moment. For the same configuration, a typical deck with I girders would have an eccentricity coefficient of approximately 4 compared with 1.4 for the box girders. There are, however, two side effects to such an encouraging behavior, which relate to torsion stresses and transverse bending of the deck slab. Torsional Moments in the Box Girder An unsymmetrical distribution of live loads in the transverse direction tends to warp the box girders and cause shear stresses. It is their high torsional rigidity which produces a favorable distribution of loads between girders. However, the maximum torsional moments usually occur when only one-half the structure (in cross section) is loaded, and the resulting stresses do not cumulate with the shear stresses produced by the full live-load shear force.

170

Design of Segmental Briees Span length, L

P (A) 4

I

41

I (z

4

I + ho (at midspan) Center of span h, (over support)

(

2d c---h *

2d’ 4 67)

FIGURE 4.25. Principle of transverse distribution of loads between box girders. (a) Dimensions. (b) Influence line of the shear in the connecting slab. (c) Transverse influence line of longitudinal moment. (d) Transverse bending influence line at section A. Transuer Moments in the Deck Slab The deck slab cannot be considered as a continuous beam on fixed supports because of the relative displacements on the two boxes due to unsymmetrical loading. Figure 4.25d shows the consequence. If the slab were resting on fixed supports, the influence line for the moment in a section such as (A) would be the typical line (1). Because the box girders undergo certain deflections and rotations, the effect is to superimpose the ordinates of another line such as (2). Numerically, the difference is not as great as may be expected at first sight, because line (1) pertains to the effect of local concentrated truck loads while line (2), being the result of differential movements between box girders, pertains to the effect of uniformly distributed loads. In summary, deck moments are increased by only 20 to 30% over their normal values if flexibility of the box

girders is ignored. As a matter of practical interest, actual numerical values for several bridges in France with either two or three box girders that have all shown excellent performance for more than 10 years are presented in Figures 4.26 and 4.27.

4.7

Effect of Temperature Gradients Bridge Superstructures

in

Experience has shown the sensitivity of long-span cantilever bridges to concrete creep. This resulted in the preference for continuous rather than hinged cantilevers. However, two more problems arose from this significant change in design approach, both being the immediate result of continuity. These problems are (1) effect of temperature gradient in bridge decks and (2) redistribution

Effect of Temperature Gradients in Bridge Superstructures

Spans

Bridge

2d

2d'

(ft)

(ft) 29.5

6.6/18.0

13.1

26.2

11.1/18.0

15.4

33.9

9.2115.7

14.1

23.3

5.9114.'

14.6

24.9

5.2/10.7

0

Givors

7360' 1 ,- (1lOiy

15.'

0

D/S Paris Ring Parkway

-300'by

0 co 0

U/S Paris Ring Park-

'

I

(90m)

7

295'

r

way Corde

Juvisy

l(79m)T 5

,

260'

T

220' yp __- -.-

@

Eccen. Coeff.

ho/h1

Choisy-leRoi

\

y

180'

7

1.23

t

1 (55m)

11.1

22.3

1

co;;fant

1 1.28

i

t

:

t

?d' I 2d I

1

FIGURE 4.26. Transverse distribution of loads between box girders, numerical values for several two-box girders. of internal stresses due to long-term effects (steel relaxation and concrete creep). The importance of these two new problems was discovered experimentally. All structures are designed, according to the provisions of the various codes, for changes of temperature that are assumed to apply to the entire section. Significant bending moments in the superstructure occur only as a result of the frame action with the piers where a rigid connection is achieved between sub- and superstructure. Actual

measurements on existing structures confirm this assumption. The average concrete section undergoes a progressive shortening due to shrinkage and concrete creep superimposed naturally with the usual seasonal temperature variations, Figure 4.286. The total concrete strain of 120 X 10m6 in./in. was very moderate for a period of four years. Daily readings, on the same bridge, of strains and magnitude of reactions over the abutment

172


I

1 2

Calculated deflection (E = 6.4 X lo6 psi)

I

Measured deflection

Measured deflection Calculated deflection (E = 6.9 X lo6 psi)

0 ,

I

I

1

I

II /I I ---

,

4.27.

1

deflection

--

I-

FIGURE

..~ Measurec

Calculated deflection (E = 7.4 X lo6 psi)

Transverse distribution of loads between box girders.

brought to light a factor that had previously been ignored. This was the differential exposure of the bridge deck to the sun on warm summer days. This situation is aggravated for bridges crossing a river, where the bottom flange is kept cool by the water and the usual black pavement placed over the top

flange concentrates the sun’s radiation. Within a 24-hour period the reaction over the abutment could vary as much as 26%, Figure 4.28~. The equivalent temperature difference between top and bottom flanges reached 18°F (10°C). The maximum stress at the bottom flange level, due

Design

of Longitudinal Members for Flexure and Tendon Profile

173

The effect is usually computed by assuming the gradient to be constant throughout the bridge superstructure length, which is not necessarily the case. Figure 4.29 shows the result for the case of a typical span built-in at both ends (this is the case of a long structure with many identical spans). The stress at the bottom fiber depends only upon the following two factors: Variation of height between span center and support (ratio hi/h,) Position of the center of gravity within the section (ratio c,lh,) The lowest stress is obtained for a symmetrical section and a constant-depth girder. The stress increases rapidly when the variation in depth is more pronounced. For normal proportions the effect of gradient is increased by 50% in variable-depth girders compared to constantdepth girders (240 psi versus 160 psi for a 9°F gradient and a modulus of 5 x lo6 psi).

4.8

Design of Longitudinal Members for Flexure and Tendon Pro$le

Cc)

FIGURE 4.28. Champigny Bridge, observed values of concrete strains and deck reactions. (a) Typical dimensions. (b) Long-term shortening of bridge deck due to concrete creep superimposed with temperature variations. (c) Daily temperature variations as exemplified by change in reactions over abutments. only to this temperature gradient, reached 560 psi (3.9 MPa), a value completely ignored in the design assumptions. Various countries of Western Europe have now incorporated special provisions on temperature gradients as a result of this knowledge. In France, the following assumptions are required: 1. Add the effect of a 18°F (10°C) temperature gradient to the effect of dead loads and normal volume changes (such as shrinkage, creep, and maximal temperature differences). The effect of gradient is computed with an instantaneous modulus of elasticity (usually 5 million psi). 2.

Add the effect of a 9°F (5°C) temperature gradient to the combined effect of all loads (including live load and impact) and volume changes, again using an instantaneous modulus of elasticity.

4.8.1 PRINCIPLE OF PRESTRESS LAYOUT The longitudinal prestress of a cantilever bridge, whether cast in place or precast, consists of two families of tendons: 1.

As construction in cantilever proceeds, the increasing dead-load moments are resisted at each step of construction by tendons located in the top flange of the girder and symmetrically placed on either side of the pier, Figures 4.30 and 4.31~. These are known as cantilever tendons. 2. Upon completion of individual cantilevers, continuity is achieved by a second family of tendons essentially placed at the center of the various spans, Figure 4.316. Because girder load moments are small, except through long-term redistribution, because of the construction procedure, the continuity prestress is designed to resist essentially the effect of: a. Superimposed loads (pavement, curbs, and the like). b. Live loads. C.

Temperature

gradient.

174


I

I

2.0

1.5

ELEVATION

OF

SPAN

2.5

3.0

ho -L

SECTION AT CENTER

FIGURE 4.29. Effect of thermal gradient on box girder decks. d. Subsequent redistribution of girder load and cantilever prestress. Tensile stresses are large at the bottom flange level, but seldom will continuity prestress gain the full advantage of the available eccentricity because of the stress conditions at the top flange level. Usually this prestress is divided into tendons, B 1 or B2, located in the bottom flange, and a few tendons such as B3 which overlap the longer cantilever tendons, Figure 4.3 lb. For the best selection of prestressing methods, it is essential to use prestressing units of a capacity large enough to reduce the number of tendons in the concrete section, particularly in very long spans. On the other hand, there must be a sufficient number of tendons to match with the number of segments in the cantilever arms. Also, units with an excessive unit capacity will pose serious problems for the transfer of concentrated high loads, particularly for cast-in-place structures, where concrete strength at the time of prestress is always a critical factor within the construction cycle. In practical terms, prestress bars are as well adapted to short and medium spans as strand ten-

dons (such as twelve 3 in. diameter strands). For very long spans (above 500 ft) large-capacity tendons (such as nineteen 0.6 in. diameter strands) with a final prestress force of about 700 kips afford a very practical solution for cantilever prestress. For continuity prestress the size of tendons is governed by the possibility of locating the tendon anchors in such areas and with such provisions as to allow a proper distribution of the concentrated load to the surrounding concrete section. Units such as twelve 3 in. diameter or twelve 0.6 in. diameter are usually well adapted with careful detailing for this purpose. 4.8.2

DRAPED

TENDONS

In early applications, both families of prestress were given a draped profile in the web of the box section to take advantage of the vertical component of prestress to reduce the shear stresses. In such a configuration there is a considerable overlapping of tendons in the web, because the cantilever prestress is anchored in the lower part of the web and the continuity prestress is anchored at the top flange level; see the layout in Figure 4.31~. For a constant-depth section and for segments of equal

Diagrams of moments in a cantilever

4.30. Typical cantilever moments and prestress. When placing unit 8, the increase of bending moment is represented by the hatched area and the resultant curve is transferred from position 7 to position 8. Additional sets of cantilever prestressing tendons are placed each time a pair of segments is erected. This procedure allows the magnitude of prestress to follow very closely the various steps of construction. FIGURE

length, it is easy to completely standardize the layout of prestress in various segments. Mechanization of the casting operations is a very desirable feature, all prefabricated reinforcing cages being identical, with ducts always at the same locations. A substantial amount of repetition may still be obtained in variable-depth members as seen in Figure 4.32, which represents a typical span of the Oleron Viaduct. The two disadvantages of such a prestress layout are: Cantilever tendon anchors are located in the web and it is difficult to prevent web cracking, particularly in cast-in-place structures, except through the use of thicker webs and smaller tendons. Continuity tendons extend above deck level at both ends. The installation of the anchor with the block-out for stressing is difficult in the casting

form, and good protection against water seepage to the tendons in the finished structure is a critical factor. 4.83 STRAIGHT TENDONS

Tendons are in this configuration located in the upper and lower flange of the box girder and anchored near the web in their respective flanges. There is no draped profile for the tendons within the web and consequently no reduction of shear stresses due to a vertical component of prestress. This is a disadvantage of this scheme, which may often require vertical prestress to maintain shear stresses within allowable limits. On the other hand, the two advantages are: Simplicity in both design and construction

176

Design of Segmental Bridges (AI )

span

Average

length

of

L

cantilever

tendons

0.52 L

fd

Average

length

of

contlnuit?

tendons

I

: 0 35 - 0.50 L

(b)

.4 : cantilever tendons B : continuity tendons

0 A

Q A

FIGURE 4.3 1. Typical layout of longitudinal prestress. (a) Cantilever tendons. (b) Continuity tendons. (c) Standardized layout of tendons for constant-depth segments.

Significant reduction in friction losses of the prestress tendons for both curvature and wobble effects, and consequent savings on the weight and cost of the longitudinal prestress of at least lo%, all else being equal The Rio Niteroi Bridge (described in Section 3.8) used straight tendons, Figure 4.33. Typical characteristics of the deck are as follows:

Span length

262 ft

Width of a box

42 ft

Two webs at

14.2 in. each 42 (12 4 in. diam. strands) 14 (12 f in. diam. strands) 1 in. diameter bars

Longitudinal prestress Longitudinal prestress Vertical

cantilever continuity

prestress

Cantilever prastress 30412 x 1.2” 6) +8-(12x .315”9)

15

I<

13

I?

1’1 ,o

Continuity prestress--14-(12x l/2” @I + 4(12 x .315” $1

Detail B Detail A Transverse

prestress

r

Longitudinal

FIGURE 4.32. Oleron

prestress

Viaduct, longitudinal prestress.

distribution


178

Vertical bars 25 mm 9 (typ. )

TOP

PRESTRESS 12 strand 12.7 mm # cablas

FIGURE 4.33. Rio-Niteroi Bridge, typical prestress layout.

Critical stresses near the pier are: Longitudinal compression Vertical compression Maximum shear stress Diagonal stresses

850 psi 400 psi 580 psi - 110 psi (tensile), and 1360 psi (compressive)

Typical details of tendon profiles and anchorages are portrayed for Linn Cove Viaduct in North Carolina, U.S.A., in Figures 4.34, 4.35, and 4.36. 4.8.4

SUMMARY OF TENDON PROFILES AND ANCHOR LOCATIONS

In the two preceding configurations, tendons were anchored in the following manner: 1. For cantilever prestress: a. On the face of the segment in the fillet between top flange and web. b. On the face of the segment along the web. c. In a block-out near the fillet between top flange and web, but inside the box. 2. For the continuity prestress: a . At the top flange level. b. In a block-out near the fillet between web and bottom flange.

c.

In a block-out in the bottom flange proper away from the webs.

Configurations lc, 2b, and 2c all permit prestressing operations to be performed safely and efficiently inside the box, Figure 4.37, permitting such operations to be removed from the critical path of actual placement or construction of the segments. Only those tendons required for balancing the self-weight of the segments need to be installed at each step of construction. The balance of the required prestressing may thus be installed later, even after continuity is achieved between several cantilever arms. Tendons for the additional prestress may then be given a profile comparable to that used in cast-in-place bridges with a length extending over several spans. The practical limit to this procedure is excessive sophistication and related high friction losses in the tendons. 4.8.5 SPECIAL PROBLEMS OF CONTINUITY PRESTRESS AND ANCHORAGE THEREOF

Tendons for continuity prestress may not, or even should not, always be located in the fillet between web and bottom flange. They may be located in the bottom flange proper. When a variable-depth member is used, the bottom flange has a curvature in the vertical plane, which must be followed by the prestress tendons. Unless careful consideration is

Design

of Longitudinal Members for Flexure and Tendon Profile 5 s&u e 7 “: Z-‘/1 .’

\ c 8’ .I. e ; ” FIGURE 4.34.

3

SW 0

7”,1!4’

Linn Cove Viaduct, typical cross section showing prestress ducts.

given to that fact at the concept and detailed design stages, difficulties are likely to develop; we may see this by looking at Figures 4.38 and 4.39, which show the free-body diagrams of stresses in the bottom flange due to the curvature, together with a numerical example. Curvature of a tendon induces a downward radial load, which must be resisted by transverse bending of the bottom flange between the webs. Longitudinal compressive stresses in the bottom flange similarly induce an upward radial reaction in the flange, counteracting at least in part the effect of the tendons. Unfortunately, when the full live load and variable effects, such as thermal gradients, are applied to the superstructure, the lon-

gitudinal stresses vanish and consequently the partial negation of the effect of tendon curvature is lost. Therefore, the effect of tendon curvature adds fully to the dead-load stresses of the concrete flange. The corresponding flexural stresses are four to five times greater than the effect due to dead load only, and if sufficient reinforcement is not provided for this effect, heavy cracking is to be expected and possibly failure. Practically, the situation may be aggravated by deviations in the location of the tendon ducts in the segments compared to the theoretical profile indicated on the drawings. At the point between segments, ducts are usually placed at their proper position; but if flexible tubing is used with an insufficient number of sup-

F

ANCHORAGE

/

A .?:o*

I i

I

__-----------t

FIGURE 4.35. 180

Linn Cove Viaduct, top flange prestress details.

:‘s HOLES FOR TEMPORARY PRESTRESSING 1’1)‘8 ~-~--~

BARS

D U C T 3 ‘/I$’ 0

FIGURE 4.36.

I .--__~__--.-.--p.p.

Linn Cove Viaduct, bottom flange prestress details.

181

182


PARTIAL

CROSS-SECTION

FIGURE 4.37. B-3 South Viaduct, prestressing operations in box girder. porting chairs or ties, the duct profile will have an angle break at each joint. In addition to the increased friction losses, there is a potential danger of local spalling and bursting of the intrados of the bottom flange, Figure 4.40. Rigid ducts properly secured to the reinforcement cage and placed at the proper level over the soffit of the casting machine or traveler will avoid this danger. Another item concerning potential difficulties in continuity prestress relate to the projection of the anchor block-out in the bottom flange and where anchor blocks are not close to the fillet between web and bottom flange. When this method is used in conjunction with a very thin bottom flange (a

Assumed

COMPRESSIVE STRESSES

FREE BOOY DIAGRAM

FIGURE 4.38. Secondary stresses due to curved tendons in the bottom flange.

flange as thin as 5 or 6 in. has been used in early bridges), it is almost impossible to distribute the concentrated load of the anchor block in the slab without subsequent cracking. For a 7 or 8 in. flange it is recommended that no more than two anchor

Longitudinal

Radius

(12 x l/2"+

1,000'

tendons) Typ.

a. = 15.67' 4

lo'-0"

I_

lo'-0"

FIGURE 4.39. Secondary stresses due to curved prestressing tendons, numerical example. Assumed longitudinal radius = 1000 ft. Weight of bottom slab = 100 psf. Effect of compressive stresses: unloaded bridge,f, = 2000 psi, compressive radial load: f,tlR = (2000 x 8 x 12)/1000 = 200psf; loaded bridge, 0 psi. Effect of prestressing tendons: stranded tendons (twelve f in. dia strands) at 10 in. interval with a 280 kip capacity, corresponding radial load: F/R = 280,000/[( 10/12)1000] = 336, say 340 psf. Total loads on bottom slab: (1) during construction, load = 100 psf; (2) unloaded bridge, load = 100 - 200 + 340 = 240 psf; (3) loaded bridge, load = 100 + 340 = 440 psf, moment = we2112 = 9 kips ft/ft, stress in bottom slab: f = M/S = (9000 x 12)/[( 12 x 64)/6] = 840 psi.

Design of Longitudinal Members for Flexure and Tendon Profile

tendons may be made continuous through the expansion joint or equipped with couplers.

Pamal elevation

d. Resume normal cantilever segment placing and prestressing to the center of the span, with tendons crossing the joint. e. Achieve continuity with previous cantilever by pouring closure joint and stressing continuity tendons. Layout of these tendons includes anchors in the special hinge segment to transfer the shear forces in the completed structure. f.

FIGURE 4.40. Ef‘fect

of’ misalignment of’ continuity

prestress.

blocks for (I2 f in. diameter strands) tendons be placed in the same transverse section in conjunction with additional reinforcing to resist bursting stresses. Wherever possible, the anchor blocks for continuity tendons should be placed in the fillet between the web and flange where the transverse section has the largest rigidity. 4.8.6 L,-iYOI’T OF PRESTRESS I.Y STRUCTURES IZ’ITH HI.Z’GES ,4SD EXP,4.\‘SION JOINTS

Section 4.4.3 explained how the expansion joints in the superstructure should be located preferably near the contraflexure point of a span rather than at midspan as in previous structures. However, there is a resultant complication in the construction process, because cantilever erection must proceed through the special hinge segment. A typical construction procedure and the related prestress layout are presented in Figure 4.4 1. For the geometry of the structure in this figure, the construction proceeds as follows: a.

b. C.

Place the first five segments in balanced cantilever and install cantilever prestress for resistance against dead load. Place the lower half of the special segment and the corresponding tendons. Install the upper half of the special hinge segment with permanent, or provisional bearings, and provisional blocking to permit transfer of longitudinal compressive stresses. Cantilever

Remove temporary blocking at hinge. Release tension in cantilever tendons holding segments 7, 8, and 9 or cut tendons across the hinge after grouting. 4.8.7 REDISTRIBUTION OF MOME,VTS AND STRESSES THROUGH CONCRETE CREEP

In a statically indeterminate structure the internal stresses induced by, the external loads depend upon the deformation of the structure. In prestressed concrete structures such deformations must include not only short-term but also longterm deformation due to relaxation of prestressing steel and concrete creep. In conventional structures such as cast-in-place continuous superstructures, the effect is not significant if all loads and prestress forces are applied to the statical design of the completed structure, which is the common case of construction on scaffolding. The behavior of cantilever bridges, particularly cast-in-place structures, is quite different, because the major part of the load (the girder load often represents 80% of the total load in long spans) is applied to a statical concept that is different from the completed design. As soon as continuity is achieved, the structure tends to resist the new situation in which it has been placed; this is one aspect of a very general law in mechanics whereby consequences always oppose their cause. A very simple example is presented in Figure 4.42, which will provide the basis for a better appreciation of the problem. Assume two identical adjacent cantilever arms built-in at both ends and free to deflect at the center. The self-weight produces a moment:

at both ends with a corresponding deflection and rotation at the center of y and o.

184

Design of Segmental Bridges I Cantilever &$qqq 5 3 4 2 tendons

1111

67)

Tendon

Cantilever tendons for construction

FIGURE 4.41. Construction procedure and prestress in a span with an expansion joint.

If the load is applied for a short time, the value of E to take into account is Ei (instantaneous modulus). Assuming that continuity is achieved between the cantilevers as shown in Figure 4.42c, there cannot be an angle break at the center, but only a progressive deformation of the completed span. After a long time the concrete modulus has changed from its initial value Ei to a final value E,, which may be approximately 2.5 times less than Ei . Because the external loads are unchanged and the structure is symmetrical, the only change in (he state of the structure is an additional constant moment M, developing along the entire span and increasing progressively with time until the concrete creep has stabilized. At all times the magnitude of this moment adjusts in the structure to maintain the assumed continuity at the center.

y2, takes The additional deflection at midspan, place in a beam with fixed ends under the effect of its own weight and only because of the progressive change of the concrete modulus from the value Ei to the value E,. Considering the concrete strain at any point of the structure, the total strain q is the sum of two terms: Ef = E, + Ep

where cr = strain before continuity is achieved, E2 = strain after continuity is achieved. Hooke’s law relating stress and strain at a particular point in time states: E, =g 1

185

Design of Longitudinal Members for Flexure and Tendon Profile

In other words, the effect of concrete creep is to place the final stresses in the structure in an internal state (either of moments, shear forces, deflections, or stresses) intermediate between: The initial statical design with free cantilevers, and The completed design with continuity. Assume, for example, EfIEi = 0.40. Thus:

f= 0.4Of, + O.SOf, The relationship is equally true for moments, shear forces, or deflections. Moments over the support are: In the free cantilevers, M = M, In the continuous structure, M = 3M,

MO-MI Ml

Cd)

The final moment is therefore:

c

M, - M, = 0.40M, + 0.6O(fA4,)

= 0.80M,

and

M, = 0.20 M,. FIGURE 4.42.

crete

creep.

Redistribution of stresses through con-

At midspan,

moments are:

In the free cantilevers, M = 0 Similarly there is a relationship between the additional strain e2 and the corresponding stressfi produced at the same location by the same loads applied in the continuous structure. One may write: E$

f2 =EC

where E,, the creep modulus, is given by: 1 -= -1 - 1 E, Ef Ei or 1 l 2=f2 --+ i Ef 1 1 Thus:

The corresponding total stress in the structure then becomes:

In the continuous structure, M = MO/3 and the actual final moment:

M, = 0.60 +- = 0.20M, The above derivation applies not only to external loads but also to the effect of prestressing. Continuity prestress applied to a continuous structure gives little internal redistribution of moments except in multispan structures, where the spans react with one another according to the actual construction procedure. Cantilever prestress, which acts to offset an appreciable part of the dead-load moments, tends to reduce the distribution of moments due to external loads, Figure 4.43. Up to now the concrete modulus has been assumed to take only the two values Ei and E, (short-term and long-term values). In fact, because construction of a cantilever takes several weeks (or even several months in the case of cast-in-place structures), account must be taken of the concrete strains versus the age and the duration of loading.


186

a = L/2 I MGL Pe MO=

ia 1 dx n a dx { T

M

= Girder Load Cantilever Moment = =

Cantilever MGL

-

Prestress

Moment

Pe Moment Inducing Redistribution

MO

= Moment at

6 under M in continuous beam

I

= Moment of Inertia (variable)

FIGURE 4.43. Computation of moment redistribution due to dead load and cantilever prestress.

Such relationships are presented for normalweight prestressed concrete and average climate in Figure 4.44. Concrete strains are presented for convenience as a dimensionless ratio between the actual strain and the reference strain of a 28-day-old concrete subjected to a short-term load. We see that short-term strains vary little with the age of the concrete at the time of loading except at a very early age. However, long-term strains are significantly affected by the age of the concrete. For example, a three-day-old concrete will show a final strain 2.5 times greater than a three-monthold concrete. This is particularly important for cast-in-place structures with short cycles of construction (two pairs of segments cast and prestressed every week, which has now become common practice). Two other factors play an important role in the redistribution of stresses in continuous cantilever bridges:

1. Relaxation of prestressing steel and prestress losses. Because the stress in the prestressing steel varies with time (a part of that variation being due precisely to the concrete creep), the internal moments that produce the deformation of the structure and therefore originate the redistribution of stresses varv continuallv. This factor is important because’the resultant moments in the cantilever arms (dead load and prestress) are given bv the difference of two large numbers, and a variation on one usual+ has an important effect upon the result, Figure 4.43. 2. Change of the mechanical properties of concrete section. For the sake of simplicity gross concrete section is usually adopted computation of bending stresses. In fact, section to be used should be:

the the for the

a. The net section (ducts for longitudinal prestress deducted from the concrete sec-

Design of Longitudinal Members for Fkxure and Tendon Profile

187

Because it is difficult for some engineers to depend fully upon computer solutions in approaching a design problem, it is desirable to have orders of magnitude of the moment redistribution for preliminary proportioning and dimensioning of the structure. The following guidelines are based on experience and judgment.

L -mp”pDAYS

-v MONT,,-3

YEARS

1. Consider the case of a symmetrical span made up of two equal cantilevers fixed at the ends and built symmetrically. Compute girder load moments of the typical cantilever and prestress moments using the final prestress forces and the transformed concrete sections with n = 10 (average). 2. Compute the moment at midspan due to the difference of the above two loading cases (Figure 4.43). More generally, compute in the final structure the moments in the various spans due to the difference between cantilever girder load and moments and final prestress moments, including the restraint due to piers if applicable. 3. Reference is made now to the formula given previously and repeated here for convenience:

FIGURE 4.44. Concrete strains versus age and dura-

tion ot loading. Note that strain is given as a dimensionless ratio beuqeen the actual strain and the reference str-ain of a 2%dav-old concrete subjected to shot-t-term hid.

b.

tion) for effect of girder load and prestress up to the time of tendon grouting. The transformed section (with incorporation of the prestress steel area with a suitable coefficient of transformation) after grouting, where the coefficient of equivalence n = E,JE,, ratio of the modulus of steel and concrete, should be taken as a variable with time, from 5 to 12 or even 15.

The above discussion indicates the complexity of the problem with respect to the material properties and indicates the unreliable results of the early designs. The only acceptable solution is the global approach, whereby a comprehensive electronic computer program analyzes step by step the state of stresses in the structure at different time intervals and whenever any significant change occurs, thus following the complete history of construction. Such programs are now available and have proven invaluable in helping us understand the behavior of segmental bridges. They provide efficient tools for the final design of the structure.

w h e r e f = final stress (or moment or shear load in the structure at any point), ,ft = stress at the same point obtained by adding all partial stresses for each construction step using the corresponding statical scheme of the structure, f2 = stress at the same point assuming all loads and prestress forces to be applied on the final structure with the final statical scheme, Ei = initial or intermediate modulus of elasticity (short-term or for the duration of loading before continuity), E, = final modulus (long-term). Using different assumptions on the construction sequence of bridge decks and the corresponding strains as given by Figure 4.44, we find that the average value of EfIEi would vary from 0.50 to 0.67. It is recommended that the conservative value of 0.67 be used in this approximate method. Thus the actual moment due to redistribution should be 0.67, the value computed under paragraph 2. This moment must be added to the effect of live load and thermal gradient at midspan.


188

tween cantilevers of different ages, and the redistribution of support moment may thus vary in wide proportions, Figure 4.45. To keep on the safe side, it is not recommended that the reduction in support moment be taken into account in designing the prestress forces. It is interesting at this stage to give some orders of magnitude of moment redistribution by considering some fundamental formulas given as reference in Figure 4.46. It has been assumed: That the secondary moment due to the stressing of continuity tendons is 6% of the total moment over the support, That the distance, n, between the center of gravity of the cantilever tendons and the top slab is equal to 0.05h.

MOMENTS DUE 10 REDISTRIBUTION ft-kips

That the center of gravity, depending upon the section dimensions, may vary between (c,lh = 0.4 and c,lh = 0.6) and (c,lh = 0.6 and c,lh = 0.4). +1260

a

BOTH CANTILEVER5 OF SAME AGE (BUILT IN

@

CANT(l) CANT(2)

BUILT O-100 DAYS BUILT 100-200 DAYS

@

CANT(2)

ONE

YEAR

OLDER

THAN

100

That the efficiency factor is p = 0.60.

DAYS)

From the data indicated above and in Figure 4.46, the percentage of prestressing steel,p, may be determined as follows:

CANT(I)

FIGURE 4.45. Variation of redistribution moment in cantilever construction with the construction procedure.

4. Correspondingly, the support moment (over the piers) is decreased by the same amount. In fact, the construction of cantilevers in successive stages is such that continuity is achieved in each span be-

r2/c2 -I

1

I

i

r2/c1

assuming a final stress in the tendons of 160 ksi assuming a maximum compressive stress in the bottom flange of 2000 psi: P = AJ&, = 2 0 0 0 +

T Cl

r\”

limit of the central core r2 s;

r2 -= PC1 CZ

= PC2

p = efficiency factor c2

average

stress =

2000

g

.fi \ ~200~ psi

FIGURE 4.46. Approximate moment redistribution (moments over support). Total moment: MT = MGL + MSL + ML,,where MGL = girder load moment, MSL = superimposed load moment, ML, = live-load moment (including impact). Assumed secondary moment due to continuity prestress: 0.06 M,. Final prestress force: P = 0.94M,l[r + (r*/c,)] = 0.94MJ(e + pc2). Prestress moment (1): Pe = 0.94M,l[l + (pcJ~)]. Momentinducing redistribution: MGL - Pe, given by (2): (MGL - Pe)IM, = M,,lM, - 0.94/[1 +

hJe)l.

189

Design of Longitudinal Members for Flexure and Tendon Pro@

AS P = A,--8;)) For a symmetrical section, cr = 0.5h, andp would, thus, be equal to 0.63%, a reasonable and common value. The transformed percentage area of the steel with n = 10 is equal to: np =

sections is plotted versus the position of the centroid with or without transformed area. It is interesting to study the effect of an accidental variation in the prestress load due to excessive friction in the ducts. Assume, for example, a reduction of 5% in the prestress load for the case c,lh = 0.5 (symmetrical section over the support) and M,,lM, = 0.80. The intial values of (M,, - Pe)IM, are changed as follows:

0.125 +

0.500 -

Prestress

95%

Percent Variation

0.236 0.265

0.264 0.292

1.12 1.10

100%

All mechanical properties of the section change to make the denominator of equation (2) in Figure 4.46 increase and, consequently, the momentinducing redistribution increase also. This fact, which was completely overlooked for many years, is clearly seen in Figure 4.47, where the percentage of moment-inducing redistribution in the various

0.600 -

Prestress

I

I

Gross area

Transformed area

The combined effect of tendon grouting and of added friction losses increases the redistribution of moments by 25%.

I

-.--GrossArea Transformed

I

I

Area

Cl lh

0.35

I 0.40

I 0.45

I 0.50

I 0.55

I 0.60

0.65

czlh

0.65

0.60

0.55

0.50

0.45

0.40

0.35

Figure 4.47.

Moment redistribution, numerical values over support.


190 4.8.8

PREDICTION OF PRESTRESS LOSSES

The prediction of losses in prestressed concrete has always been subject to uncertainty. This is due to the high stress levels used for the prestressing steel, the variable nature of concrete, and its propensity to creep and shrink. As recently as 1975, AASHTO made a major revision to its code to provide improved methods for predicting prestress losses. The Structural Engineers Association of California has an excellent report on creep and shrinkage control for concrete in general. The report concludes that special attention should be given to material selection and proportioning. For creep and shrinkage calculations many European engineers recommend the guidelines of the Federation Internationale de la Precontrainte, Comiti Europeen du B&on (FIP-CEB). The design computations for segmental prestressed concrete bridges are very involved for the construction phase. Every time a segment is added or a tendon is tensioned, the structure changes, and it must be reanalyzed. As the segment ages, the concrete and prestressing steel creep, shrink, and relax. Thus, each segment has its own life history and an elastic modulus that depends upon the age and composition. To accurately compute all of these effects by hand, throughout the life of the structure, would be very difficult, particularly during the construction phase. Comprehensive computer programs such as “BC” (Bridge Construction) and others have been recently developed and are now available to aid the design engineer. In addition to construction analysis, these programs will check the completed bridge in accordance with AASHTO specifications. It is possible to revise them to satisfy other codes or loadings, such as AREA. Not only are all prestress losses properly evaluated and taken into account, but redistributions of moments due to concrete creep and steel relaxation are automatically incorporated in the design analysis. 4.9

Ultimate Bending Capacity of Longitudinal Members

Basically, the design approach of segmental bridges is one of service load. It is important, however, not to lose sight of the ultimate behavior of the structure to ensure that safety is obtained throughout.

In simply supported structures, the ultimate capacity is very simply analyzed by comparing in the section of maximum moment: The total design load moment including girder load and superimposed load (DL) and live load (LL ) The ultimate bending moment of the prestressed section M, Depending on the governing codes and the usual practice in various countries, this comparison may be done in various ways: Apply a load factor on DL and LL and a reduction factor for materials on M, Apply a single factor K on (DL + LL) and compare with M, Apply a single factor K on LL only and compare DL + KLL with M, In all cases, the designer must first compute the ultimate capacity of the section considering the concrete dimensions and characteristics of prestressing tendons (and possible conventional reinforcement). From previous studies it may be shown that the ultimate moment of a prestressed section is computed very simply by considering a dimensionless factor called the weight percentage of prestressing steel, q (see Figure 4.48). To account for the fact that the concrete characteristics are less reliable than those of the prestressing steel, which are well known and very constant, fs is usually taken equal to the guaranteed minimum tensile strength, whereas,fi is assumed to be only 80% of the 28-day cylinder strength. Considering now the case of- segmental superstructures, which are most generally continuous structures, one may take the conventional approach of considering the various sections of the member (for example, support section and midspan sections in the various spans) as independent from one another in much the same way as for simple members. Such simplification overlooks the capacity of the redundant structure to redistribute, internally, the applied loads, which seems to be a conservative assumption. In fact, it is not always as conservative and safe as it looks, as will be shown by an example computed numerically for a typical span of the Rio Niteroi Bridge. For such a span the design moments are as follows (in foot-kips x 1000):

Ultimate Bending Capacity of Longitudinal Members

191

As prestressing steel FIGURE 4.48. Ultimate moment of a prestressed section. (1) Dimensionless coefficient, q’ = (A,lbd) f&/f:), whereA, = area of prestressing steel, 6 = width of section,d = effective depth of section (distance between centroid of prestress and extreme compression fiber), f,i = ultimate tensile strength of prestressing steel,fi = ultimate compressive strength of concrete. (2) Value of ultimate moment: for q’ < 0.07, M, = 0.96A&d; for 0.07 < q’ < 0.50, M, = (1 - O.Gq’)A,J-?‘d.

Support Girder load Superimposed load

116 10

Total dead load (DL) Total live load (LL)

126 29

155 Total (DL + l.L) Live-load moment in simple span: 37

Midspan 0 5 5 2-2 27

The ultimate moments have been computed for all sections for both positive and negative bending. The envelopes of ultimate moments are shown in Figure 4.49. Neglecting any moment redistribution, the situation would be the following over the support and at midspan: Section Moment

IV” DL LL M, = or .M, =

Support

Midspan

256 126 29 1.65(DL + LL) DL + 4.5 LL

79 5 22 2.93(DL + LL) DL + 3.4LL

The picture is substantially different when looking at redistribution due to plastic hinges. Assuming an overall increase of both dead and live load

simultaneously (loading arrangement A), we ob-

tain the overall safety factor by comparing the sum of ultimate moments over the support and at midspan: 256 + 79 = 335 and the sum of simple span moment due to DL and LL: DL: 126 + 5 = 131 LL 37

Total

iii

The overall safety factor is thus: K = $ =2.0

approximately 20% higher than for the support section considered alone. In fact, it is more important and more realistic to consider only an increase of the live load, which is the only variable factor in the structure. Proceeding as before, the safety factor on LL only would be: K

=

335

-

37

13’

=

j

j

.

However, this is not the actual safety factor of the structure, because there exists a more aggressive loading arrangement than that where all spans are live loaded. In the case where the live load is applied to only every second span [arrangement

r


192

260

!-

260

Elevation

I

I

I

I Live-load arrangement (A)!

I

I

260

Support

I FIGURE

4.49.

*

I

I

I

-

I

!

, support

I 1

I i

I

I

i-----f

Ultimate bending capacity of’ a continuous deck.

(b) in Figure 4.491, the first plastic hinge will appear at the center of the unloaded spans with a negative moment (tension at the top fiber) and the support moment reaches the following limiting value: Ultimate negative moment at midspan:

38

Actual dead-load moment in simple span: 126 + 5 = 131 169 This value of 169 is substantially lower than the ultimate moment at that support section considered by itself (M, = 256). The failure appears when the second plastic hinge appears at the center of the loaded span under positive moment (tension at the bottom fiber). The limiting value of the safety factor K is such that: 169 + 79 = 131 + K . 37 and K = 3.2

In such structures a very important characteristic must be emphasized. At the time of ultimate load failure, due either to negative moment in the unloaded spans or positive moments in the loaded spans, the maximum moment over the support has only slightly increased above the value at design load ( 169 against 155) and is far below the ultimate moment of the section (256). Three interesting consequences may be derived from this fact: 1. Because the overall safety of the structure is not dependent upon the ultimate moment near the supports, it is not necessary to dimension the bottom flange of the concrete section in this area to balance the ultimate capacity of the prestressing tendons. 2. The global safety factor of the structure depends directly on the capacity of the sections near midspan for both positive and negative moments. The capacity for positive moments is given by the continuity tendons placed in the

Shear and Design of Cross Section

3.

bottom flange for service-load conditions. The capacity for negative moments depends upon the tendons placed at the top flange level to overlap the cantilever tendons of the two indiiidual cantilever arms. The magnitude of this overlap prestress does not appear as a critical factor when designing the structure for service loads, yet it plays an important role in the ultimate behavior of the structure. At ultimate load, it was shown that the areas of the members close to the supports are subjected to moments only slightly in excess of design load moments and in most cases below cracking moments. No early failure due to combined shear and bending is anticipated.

In long structures where hinges and expansion joints are provided in certain spans, the same design principles may be applied to analyze the ultimate capacity. Hinges represent singular points through which the moment diagrams must go regardless of the loading arrangement under consideration. It was found that the optimum location of the hinge with regard to ultimate safety is somewhat different from the location allowing the best control of long-term deflections. It may be of interest therefore to move the hinge slightly toward the center of the span, which has a further advantage of simplifying construction. 4.10

Shear and Design of Cross Section 4.10.1

I,\‘TRODUCTIO,V

Designing prestressed concrete members for shear represents a challenging task for the engineer, because there are many differences of opinion and large variations in the requirements of the various codes. In particular t h e AC1 c o d e a n d t h e AASHTO specifications differ in several ways from the FIP-CEB and other European codes. It is common practice in many countries to design reinforced concrete and prestressed concrete members for shear by allowing the concrete to carry a proportion of the shear loads while stirrups (formerly in conjunction with inclined bars) carry the rest. A complete agreement has not yet been reached on this aspect of design for shear: The French codes (CCBA, for example) allow nothing to be taken by the concrete and the total shear to be carried by the transverse steel, which is certainly an overconservative approach. Obviously,

193

the beneficial effect of longitudinal compression (either in columns subject to axial load or in prestressed members) is taken into account. The recent FIP-CEB code allows some proportion of the shear to be carried by the concrete. AC1 code allows a larger proportion of shear to be carried by the concrete with a consequent savings in stirrup requirements. 4.102 SHEAR TESTS OF REINFORCED CONCRETE BEAMS

Tests were recently carried out in France in order to increase the knowledge of this phenomenon, both on simply reinforced concrete and on prestressed members.4 Static tests on reinforced concrete I beams showed that the steel stress in stirrups increases linearly with the load and is three times smaller than it would be if the concrete carried no shear, Figure 4.50. In this respect, all codes are fully justified in taking the concrete into account as a shear-carrying component. However, dynamic testing on the same beams showed a very different behavior. A cyclic load was applied between one-third and two-thirds of the ultimate static load for one million cycles, whereupon the beam was statically tested to failure, Figure 4.51. Before cracking, the elastic behavior of the homogeneous member kept the steel stress in the stirrups very low. However, before 10,000 cycles, a crack pattern had appeared that remained to the end of the test and became more and more pronounced with a continuous increase of the inclined crack width. Crack opening reached &r in. (1.5 mm) at the end of the dynamic test. Most probably stirrup rupture took place about 600,000 cycles, although the ultimate static capacity of the

FIGURE 4.50. Static test of reinforced concrete Ibeam steel stress in stirrups.


lbre de cycles ~(log N) , 1, + aa 106

FIGURE 4.51. Dynamic test of reinforced concrete

I-beam web cracking and variation of steel stress in stirrups. beam after dynamic testing was substantially the same as for the other beams, which were tested only under static loads. Such tests show that the conventional approach of designing web reinforcement for static loading with a large part of the shear carried by the concrete may not provide adequate safety in the actual structures as soon as web cracking is allowed to develop. 4.103

DIFFICULTIES IN ACTUAL STRUCTURES

Another source of information is afforded by the behavior of existing structures. Fortunately, examples of difficulties imputable to shear in cantilever box girder bridges are scarce. The authors are aware of only two such contemporary examples, which are summarized here for the benefit of the design engineer. The first example relates to a box girder bridge deck constructed by incremental launching and

shown in Figure 4.52. Permanent prestress was achieved by straight tendons placed in the top and bottom flanges, as required by the distribution of moments. During launching an additional uniform prestress was applied to the constant-depth single box section, which produced an average compressive stress of 520 psi (3.60 MPa). Near each pier there was a vertical prestress designed to reduce web diagonal stresses to allowable values. During launching a diagonal crack appeared through both webs between the blisters provided in the box for anchorage of top and bottom prestress. The corresponding shear stress was 380 psi (2.67 MPa), and there was no vertical prestress in that zone. The principal tensile stress at the centroid of the section was 200 psi (1.40 MPa), which is far below the cracking strength of plain concrete. In fact, the webs of the box section were subjected to additional tensile stresses due to the distribution of the large concentrated forces of the top and bottom prestress. The truss analogy shown in Figure 4.52 indicates clearly that such tensile stresses are superimposed on the normal shear and diagonal stresses due to the applied dead load and may therefore produce cracking. This could have been prevented by extending the vertical prestress in the webs further out toward midspan. The second example concerns a cast-in-place variable-depth double box girder bridge with maximum span lengths of 400 ft. Because the bridge was subsequently intended to carry monorail pylons, two intermediate diaphragms were provided at the one-third and two-thirds points of each span, as shown in Figure 4.53. Prestress was applied by straight tendons in the top and bottom flanges and vertical prestress in the webs to control shear stresses. Diagonal cracking was observed in the center web only near the intermediate diaphragms with a maximum crack opening of 0.02 in. (0.6 mm). Repair was easily accomplished by adding vertical prestress after grouting the cracks. A complete investigation of the problems encountered revealed that cracking was the result of the superposition of several adverse effects, any one of which was almost harmless if considered separately: (1) The computation of shear stresses failed to take into account the adverse effect (usually neglected) of the vertical component of continuity prestress in the bottom flange of a girder with variable height. (2) The distribution of shear stresses between the center and side webs was made under the assumption that shear stresses were equal in all three webs. In fact the center web


,,i:‘,: t:“:“, \ I’ ,A

NO

VERTICAL

PRESTRESS

IN

THAT

ZONE

195

TOP PRESTRESS /

TYPICAL tIAlF (ROSS SKTlOIi A- A FIGURE

4.52.

Example of web cracking under application of’ high prestress forces.

f

FIGURE 4.53. Example of web cracking in a 400 ft span. (a) Typical cross section. (6) Partial longitudinal section. carries a larger proportion of’ the load, and shear

stresses were underestimated for this web. (3) The vertical web prestress was partially lost into the intermediate diaphragms, and the actual vertical compressive stress was lower than assumed. (4)

Present design codes do not provide a consistent margin of safety against web cracking when vertical prestress is used. This margin decreases significantly when the amount of vertical prestress increases. In the present French code, the safety

196


factor against web cracking is 2 when no vertical prestress is used and only 1.3 for a vertical prestress of400 psi. (5) At present, vertical prestress is usually applied with short threaded bars, and even when equipped with a fine thread they are not completely reliable unless special precautions are taken under close supervision. Even a small anchor set significantly reduces the prestress load, and it is not unlikely that the actual prestress load is onlk three-fourths or even two-thirds of the theoretical prestress. It should, however, be emphasized that the difficulties mentioned above have led to progress in this field, and the increase in knowledge has ensured that these examples remain rare exceptions. Practically all existing box girder bridges have performed exceptionally well under the effects of shear loads and torsional moments.

The essential aspects of this important problem are: Dimensioning of the concrete section particularly in terms of web thickness Design of transverse and/or vertical prestress and of conventional reinforcement The twvo

major considerations are:

FIGURE 4.54. Computation of net applied ~IGIIload. (0) \Terrical comporicrit of’ pr-esrr.css. (h) k:f fvc t 01 inclined bottom flange (Resal effec-t). (0 Net \hr;rt- ~OI-cc. Shear f’orce due to applied loada = I.: dcdrrct \cr-tic al component of‘dr-aped tendons = - 1 P 5irr a,; aclcl \-erticdl component of continuitv tendons = * 2 t’ 4irr a?: ticduct Red effect = - f,v.tlB tan p: total ih ner applied shear t’k~ = I’,,. are obtained bv considering stresses on sections perpendicular to the top flange (\vhich is usually the orientation of joints betlveen segmenta) and projecting the loads on the section for determining shear stresses. The total . net shear force is the sum of the following terms: Shear force due to applied loads.

At the design stage (or, in modern code language, serviceability limit state) prevent or control cracking so as to avoid corrosion and fatigue of reinforcement.

Reduction due to vertical component tendons where used.

c\t the ultimate stage (or load factor design concept state or ultimate limit state) provide adequate safetv.

Reduction due to the inclined principal compressive stresses in the bottom flange (usuallv called the Resal effect after the engineer who fi;st studied members of variable depth). Because the direction of the principal stresses in the web is not fullv determined, it is usual to neglect the added reduction of shear force derived from web stresses.

For the box sections used in cantilever bridges the behavior under shear must be investigated: In the webs. At the connections between web and top flange (including the outside cantilevers) and web and bottom flange. Figures 4.54 and 4.55 show a suggested method to compute shear loads and shear stresses. Modern computer programs analyze the box girder cross sections perpendicular to the neutral axis and take into account all loads projected on the neutral axis and the section. Equivalent results

of

draped

Increase due to inclination ofcontinuitv tendons in the bottom flange for variable-depth girders.

Shear stresses may further be computed from shear force and torsion r;loment using the conventional elastic methoils. Tests have shown that the presence of draped tendon ducts in the webs, even if grouted after tensioning, changes the distribution of’ shear stresses. To take this effect into account, it is suggested to compute all shear stresses using a net web thickness that is the actual thickness minus one-half the duct


197

t

b

b’ Gross web thickness d diameter of duct (a)

FIGURE 4.55. Computation of shear stress. Typical box section: net web thickness = h = b’ - Id; shear stress due to shear force V, net applied shear load = v = V,;Q/[(Xb).I], where Q = statical moment at centroid, b = net web thickness, I = gross moment of inertia, V,, = net applied shear load; shear stress due to the torsion moment = v = CI(P.b.3, where C = torsion moment, b = net web thickness, S = area of the middle closed box. Note: check the shear stress at centroid level.

diameter. Ducts for vertical prestress need not be taken into account because they are smaller and parallel to the vertical stirrups, which compensates for the possible small effect of the prestress ducts. Web-thickness dimensioning depends upon the magnitude of shear stress in relation to the state of compressive stress. In the case of monoaxial compression (only longitudinal prestress and no vertical prestress) the diagonal principal tensile stress must be below a certain limit to insure a proper and homogeneous margin of safety against web cracking with its resulting long-term damaging effects. Figure 4.56 suggests numerical values based on the latest state of the art that are believed to be realistic and safe. Numerical values for.allowable shear stresses under design loads are given in Figures 4.5’7 and 4.58 for 5000 and 6000 psi concrete. Web thickness must therefore be selected in the various sections along the span to keep shear stresses within such allowable values. It may be that construction requirements or other factors make it desirable to accept higher shear stresses. It is necessary in this case to use vertical prestress to create a state of biaxial compression. Figure 4.566 indicates the corresponding procedure. The vertical compressive stress must be at least 2.5 times the excess of shear stress above the value for monoaxial compression. When vertical prestress is used, the beneficial effect of increasing the length of the horizontal component of the potential crack in the web created by

the horizontal compression due to prestress is partially lost. In fact, if both horizontal and vertical compressive stresses are equal,f, = fU, the direction of the principal stress is given by /3 = 45” as in

Y

fx

v

,‘p

fx

-'\ I tu7 (a)

q1 l-&’u 1;” (b)

FIGURE 4.56. Allowable shear stress for mono- and biaxial compression in box girders. (a) Monoaxial compression: allowable shear stress = v = 0.05f:. + 0.2Of,; corresponding diagonal tension = fP given by v2 = fPcfs + f,). (6) Biaxial compression: allowable shear stress = zl = 0.05f:. + 0.2Of, + 0.4Of,; corresponding diagonal tension = fp given by v’ = cfs +f,) (fU +fP).

198


stresses higher than a limiting value of lo* be accepted prior to careful investigation based on specific experimental research. In this respect, a very interesting case arose for the construction of the Brotonne Viaduct in France (described in Chapter 9), where an exceptionally long span called for minimum weight and consequently high concrete stresses. The most critical condition for shear stresses developed in the 8 in. (0.20 m) webs near the piers of the approach spans, where a maximum shear stress of 640 psi (4.5 MPa) was accepted together with an unusually low longitudinal compression stress of 500 psi (3.45 MPa). Vertical prestress was used in this case. The chart for a 6000 psi concrete, Figure 4.58, would give: In monoaxial state withf, = 500 psi, V = 400 psi. FIGURE 4.57. r\llo\
shear stresses forf:. = 5000

psi.

In biaxial state withf, = 550 psi, V = 620 psi, which is substantially equal to the actual shear stress of 640 psi. A test was conducted to study the behavior of the precast prestressed web panels in the normal design load stage and up to failure, Figure 4.59. Results are shown in Figure 4.60. The ultimate capacity of the web was very large and probably far in excess of the needs. It is believed that web

FIGURE 4.58. Allowable shear stresses forf:.

= 6000

psi. ordinary reinforced concrete. If a higher vertical stress is used, a crack with p > 45” could develop, with a *consequent reduction of the horizontal length over which concrete and reinforcement must carry the total shear. To prevent such a situation, it is deemed preferable to use a vertical compressive stress not greater than the longitudinal compressive stress, fu < fz. Finally, considering present knowledge on the behavior of prestressed concrete beams under high shear stresses, it is not recommended that shear

PLAN

TRANSVERSE

FIGURE

4.59.

cast web panels.

VIEW

SECTION

Brotonne Viaduct, test set-up for pre-

Joints Between Match-Cast Segments Strrssrs at &sign stug~ (approach viaduct): Horizontal compressive stress Vertical compressive stress Shear stress

500 psi 550 psi 6 4 0 psi

Rrsulb of test at rufduw: Normal L o a d 630 t 84o’t Ultimate shear Horizontal compressive stress Vertical compressive stress Shear stress (elastic theory) uniform

840x 1.3 630 1650 psi 5 8 0 psi 3 3 0 0 psi 2 2 0 0 psi

.JoinI destroyed and multiple keys sheared off. Panels intact. FIGURE 4.60. Brotonne Viaduct, results of precast \\,eb panel tests.

cracking control can be obtained only by proper stress limits at the design load level. When designing longitudinal bridge members for shear, another important factor remains to be considered, which has sometimes been overlooked t& inexperienced designers. It concerns longitudinal shear stresses developing between the webs and the top and bottom flanges as shown in Figure 4.6 1. When web stresses have been verified at the level of the centroid, it is not necessary to make a detailed study at other points of the web [such as levels (d) and (e)], although the principal tensile stress near the pier may be slightly higher at point (d) than at the center of gravity. On the other hand, to keep the integrity of the box girder, it is verv important to verify that shear and diagonal stresses in sections (a), (b), and (c) are within the // /

/

FIGURE 4.61. Longitudinal shear between web and flanges.

199

same allowable values as set forth previously for the webs and that a proper amount of reinforcing steel crosses each section. This leads to the design of transverse reinforcement in the cross section to resist shear stresses. According to the provisions of the AC1 Code and the AASHTO specifications, the web shear steel requirements are controlled by the ultimate stage. The net ultimate shear force is given by the following formula, based on the current partial load factors: V, = 1.3OV,,, + 2.17V,, + V, w h e r e V, = net shear force at ultimate stage, VDL = actual shear force due to the effect of all dead loads, including the reduction due to variable depth where applicable, VLL = shear force due to live loads including impact, VP = unfactored vertical component of prestress where applicable. Effects of temperature gradients and volume changes are usually small in terms of shear load and may be neglected except in rigid frames. On the contrary, shear due to moment redistribution and secondary effects of continuity prestress must be included. A partial safety factor on material properties is applied to the ultimate load state.

4.11 Joints Between Match-Cast Segments Joints between match-cast segments are usually filled with a thin layer of epoxy to carry normal and shear stresses across the joint. In the early structures, a single key was provided in each web of the box girder to obtain the same relative position between segments in the casting yard and in the structure after transportation and placing. This key was also used to transfer the shear stresses across the joint before polymerization of the epoxy, which has substantially no shear strength before hardening. Figure 4.62 summarizes the force system in relation to a typical segment both during erection and in the completed structure. Provisional assembly of a new segment to the previously completed part of the structure is usually achieved by stressing top (and sometimes bottom) longitudinal tendons, which induce forces F, (and F2). The resultant F of F, + F2 resolves with the segment weight W into a resultant R. The vertical component of R can be balanced only by a reac-

200


FIGURE 4.62. Typical segment in relation to the force system. (0) Provisional assembly of segment(s). (b) Segment(s) in the finished structure. tion such as R, given by the inclined face of the key, while the balance of the normal force is R, which produces a distribution of longitudinal compressive stresses. In the finished structure, all normal and shear stresses are naturally carried through the joints by the epoxy material, which has compressive and shear strengths in excess of the segment concrete. A series of interesting tests were performed for the construction of the Rio-Niteroi Bridge in Brazil to verify the structural behavior of epoxy joints between match-cast segments. A l-to-6 scale model was built and tested to represent a typical deck span near the support and the corresponding seven segments as shown in Figure 4.63. ELEVATION

FIGURE

4.63.

A crack pattern developed in the web when the test load was increased above design load, as shown in Figure 4.64. The epoxy joints had no influence on the continuity of the web cracks, and the behavior of the segmental structure up to ultimate was exactly the same as that of a monolithic structure. Failure occurred for concrete web crushing when the steel stress in the stirrups reached the yield point. The corresponding shear stress was 970 psi (6.8 MPa) for a mean concrete cylinder strength of 4200 psi (29.5 MPa). The first bending crack had previously occurred for a load equal to 93 percent of the computed cracking load, assuming a tensile bending strength of 550 psi (3.9 MPa). Other tests were performed DETAIL OF JOINT

Rio-Niteroi Bridge, partial elevation and joint detail.

201

Joints Between Match-Cast Segments

FIGURE

4.64.

Rio-h’itcroi Bridge. ~vel) crack pattern at ultimate in model test.

in order to study the transfer of diagonal principal compressive stresses across the segment joints as shown in Figure 4.65. Prismatic test specimens

I

P

fd PRlSMATlC

Efficiency PRISMATIC WITH KEYS

P

P

were prepared, some with and some without shear keys across the joint, and tested for various values of p, the angle between the principal stress and the neutral axis of the girder. In the case of the Rio Niteroi Bridge the value of /3 is between 30 and 35”. For a reinforced concrete structure p = 45”. A preliminary test showed that the epoxy joint had an efficiency of 0.92 as compared to a monolithic specimen with no joint (ratio between the ultimate load P on the prismatic specimen with an epoxy joint and with a monolith specimen). For various directions of the joint the results are as follows:

I

I

t

1

P

P (b)

FIGURE 4.65. Rio-Niteroi Bridge, test specimens for web. (a) Crack pattern in web and related test specimen. (b) Actual test specimens.

0” 0.94

15” 0.92

30” 0.98

45” 0.95

60” 0.70

It can be seen that for values of p smaller than 45” (which covers the entire field of prestressed concrete members) the compressive strength is hardly affected by the presence of the inclined joint. All these tests confirmed earlier experimental studies to show that epoxy joints are safe provided that proper material quality together with proper mixing and application procedures are constantly obtained. Several early incidents in France, and some more recently in the United States, have shown that these conditions are not always achieired. The logical step in the development and improvement of epoxy joints was therefore to relieve the epoxy of

Design of Segmental Bri&es

202

any structural function. The multiple-key (or castellated-joint) design embodies this concept and provides for simplicity, safety, and cost savings. Webs and flanges of the box section are provided with a large number of small interlocking keys designed to carry all stresses across the joint with no structural assistance from the resin. Figure 4.66 shows the comparison between the structural behavior of an early joint with a single web key and a joint with multiple keys, assuming that the epoxy resin has improperly set and hardened. It is now recommended that multiple keys be used in all precast segmental projects, as shown in Figure 4.67. With the current dimensions used for depth and height of multiple keys, the overall capacity of the joint is far in excess of the required minimum to transfer diagonal stresses safely up to the ultimate load state.

Continuous transfer of shear stresses

(It)

Design of Superstructure Cross Section

4.12

The typical cross section of a box girder deck is a closed frame subjected to the following loads, Figure 4.68: Girder weight of the various components (top and bottom flanges, webs) Superimposed loads essentially applied to the top flange (barrier, curbs and pavement) and sometimes to the bottom flange, as when utilities are installed in the box girder Live loads applied on the deck slab A typical box girder element limited by two parallel cross sections, Figure 4.686, is in equilibrium because the applied loads are balanced by the difference between shear stresses at the two limiting sections. To design the typical cross section the assumption is usually made that the shape of the section remains unchanged and that the closed frame may be designed as resting on immovable supports such as A and B. Bending moments are created in the various sections of the frame due to the applied loads. Maximum moments occur in the deck slab due to live loads in sections such as (a), (b), and (f).

Cb)

FIGURE 4.66. Joint between match-cast segments, comparison between single- and multiple-key concepts.

Cd) FIGURE 4.67.

Precast segment with multiple keys.

FIGURE 4.68. Design of deck cross section. (a) Typical loading on cross section. (b) Free-body diagram.

Special Problems in Superstructure Design

Because the webs are usually much stiffer than the flanges and the side-deck slab cantilevers and the center-deck slab between webs are built into the webs, most of the deck-slab moments are transferred to the web, with a maximum value in section (d) at the connection between web and top flange. In bridges where transverse or vertical prestress or both are used, the design of the deck cross section is not greatlv affected by the fact that moments and normal forces computed in the frame superimpose their effects on the shear stresses due to longitudinal bending mentioned in Section 4.10. The case is more critical when only conventional transverse reinforcing steel is used in both flanges and webs. A common method, based on experience, is to compute the steel area required on either face at critical sections such as (a) through (e), shown in Figure 4.68, for the following: 1. 2.

Shear stresses in the longitudinal members. Transverse bending of the frame.

The minimum amount of steel should not be less than the larger of the following: item 1 plus one-half of item 2, item 2 plus one-half- of item 1, or 0.7 times the

4.13

,

Special

sum

of item 1 and item 2.

Problems

in

Superstructure

Design

All design aspects covered in the preceding sections pertain to the design of deck members for bending and shear regardless of the local problems encountered over the piers or abutments and at intermediate expansion joints when required. This section will now deal with such local problems, which are of great practical importance. 4.13.1 DIAPHRAGMS

It was mentioned in Section 4.6 that the combined capacities of the deck slab in bending and the box girder in torsion allow a very satisfactory transverse distribution of live loads between girders in the case of multiple box girder decks. It has therefore been common practice to eliminate all transverse diaphragms between box girders except over the abutments. Diaphragms inside the box section are still required over the intermediate piers in most projects.

203

4.13.2 SUPERSTRUCTURE OVER PIERS

The simplest case is exemplified in Figure 4.69, where a deck of constant depth rests upon the pier cap with bearings located under the web of the box girder. The reaction is transferred directly from the web to the bearings, and there is need only for a simple inside diaphragm designed to transfer the shear stresses, due to possible torsion moments, to the substructure. A more complicated situation arises when the bearings are offset with regard to the webs, Figure 4.70. Reinforcing and possibly prestressing must be provided in the cross section immediately above the pier to fullfill the following functions: Suspend all shear stresses carried by the web under point A, where a 45” line starting at the bearing edge intersects the web centerline (hatched area in the shear diagram). Balance the moment (R * d) induced by the bearing offset. Looking at other schemes, we find that decks of variable depth pose several challenging problems. Figure 4.71 shows an elevation of a box girder resting on twin bearings designed to improve the rigidity of the pier-to-deck connection and consequently reduce the bending moments in the deck, which will be described in greater detail in Chapter 5. When the loading arrangement is symmetrical in the two adjacent spans, the transfer of the deck reaction into the piers through the four bearings is just as simple as for the case shown in Figure 4.69. Matters look very difficult for an unsymmetrical loading condition either in the completed structure, Figure 4.71, or during construction, Figure 4.72. Let us assume that the total deck reaction is transferred to the pier through one line of bearings only (for example, R, in Figure 4.71, for an excess of load in the left span). The compression C, carried by the bottom flange at the right is no longer balanced by the corresponding reaction R,, and an abrupt change in the system of internal forces results in a large vertical tensile force T,, which has to be suspended on the total width of the box section by special reinforcement or prestress. In long-span structures, these local effects are of no small magnitude. Taking the example of a 40 ft (12 m) wide box with a 20 ft (6 m) wide bottom flange and a span of 300 ft (90 m), the load carried by the bottom flange will probably be around 3000 t (2720 mt) and the angle change above the right bearing

Design of Segmental Bridges t ~~~~ t-

bearings

SECTION

FIGURE

FIGURE

4.70.

4.69.

A-A

S E C T I O N c-c

Pier segment for deck of constant depth and simple support.

Deck over piers with offset bearings.

about 10 percent. The corresponding unbalanced load is therefore 300 t (272 mt), and this is more than enough to split the pier segment along the section between the web and the bottom flange if proper consideration has not been given to the problem with respect to design and detailing. The situation may be even more critical during construction, Figure 4.72, if the unbalanced mo-

ment induces uplift in one of the two bearings. The load of the anchor rods (2) has to be added to the unbalanced load resulting from the angle change of the bottom flange. The diaphragm systems shown in Figures 4.71 and 4.72 are of the A type where both inclined diaphragm walls intersect at the top flange level. Any unsymmetrical moment that produces a tension force in the top flange T and a compression force in the bottom flange may thus be balanced by normal loads such as F, and C,, Figure 4.7 1, with no secondary bending. In this respect, then, it is a satisfactory scheme. Detailing may, however, be difficult because of the concentration of reinforcement or prestress tendon anchors in the top flange area, which usually is already overcrowded with longitudinal tendon ducts. A simple and more practical design, although less satisfactory from a theoretical point of view, is to provide vertical diaphragms above the bearings. This is the logical choice when the deck is rigidly connected with a

205

Dejections of Cantilever Bridges and Camber Design 4.13.4

EXPANSION JOINT AND HINGE SEGMENT

The expansion joints required at intermediate points in very long structures need a special segment to transfer the reaction between the two sides of the deck. When the expansion joint is located close to the point of contraflexure there is no provision for any uplift force, even with a load factor on the live loading. The hinge segment is therefore made up of two half-segments, as shown in Figure 4.75: The bearing half (reference A), which is connected by prestress to the shorter part of the span The carried half (reference B), connected by prestress to the longer part of the span Measures are taken to continue cantilever construction through the hinge segment until closure is achieved at midspan; see Section 4.8.6. Inclined diaphragms provide an efficient way to suspend or transfer the reaction through the bearings into the flanges and webs on both sides of the box section, Figure 4.75. One of the largest structures incorporating a hinge segment of this type is the Saint Cloud Bridge, described in Section 3.12. A typical detail of this segment is shown in Figure 4.76.

Neoprene bearings -

?

I

FIGURE 4.71. Deck of variable depth, permanent

deck-to-pie]- bearing arrangement.

.

box pier and where the pier walls are continued in the deck, as shown in Figure 4.73. Here again the transfer of all symmetrical loads between deck and pier is simple, and design difficulties arise for unsymmetrical loading. .4t the connecting points A and R, Figure 4.73, between the top flange and the vertical diaphragms, the part of the top flange tension load T such as T, induces into the diaphragm another tension load T,, and both loads result in an unbalanced diagonal component T,, which must be resisted both by the webs and bv special provisions such as stiffening beams. 4.13.3 E.VD ABUTME,VTS

A special segment will be provided at both ends of the bridge deck with a solid diaphragm to transfer torsional stresses to the bearings, as shown in Figure 4.74. The expansion .joint is, therefore, adequately supported by the end diaphragm on one side and the abutment wall on the other side.

4.14

Dejections of Cantilever Camber Design

Bridges

and

Each cantilever arm consists of several segments, fabricated, installed, and loaded at different points in time. It is important therefore to predict accurately the deflection curves of the various cantilevers so as to provide adequate camber either in the fabrication plant for precast segmental construction or for adequate adjustment of the form travelers for cast-in-place construction. When the structure is statically determinate, the cantilever arm deflections are due to: The concrete girder weight The weight of the travelers or the segment placing equipment The cantilever prestress After continuity between individual cantilevers is achieved, the structure becomes statically indeterminate and continues to undergo additional deflections for the following reasons:

206


e

FIGURE 4.72. Temporarv pier and deck connection.

Continuity prestress Removal of travelers or segment placing equipment Removal of provisional supports and release of deck to pier connections Placing of superimposed loads Subsequent long-term deflections due to concrete creep and prestress losses will also take place. Compensation for the following three types of deflections must be provided for by adequate camber or adjustment:

1. Cantilever arms. 2. Short-term continuous deck 3. Long-term continuous deck. It has already been mentioned that the concrete modulus of elasticity varies both with the age at the time of first loading and with the duration of the load (see Section 4.8.7). Deflections of types 2 and 3 above are easily accommodated by changing the theoretical longitudinal profile by the corresponding amount in each section to offset exactly all future deflections. A more delicate problem is to

Dejections of Cantilever Bridges and Camber Design

FIGURE

4.73.

Pier segment with vertical diaphragms.

I 2.2 5

50

SECTION

207

accurately predict and adequately follow the deflections of the individual cantilever arms during construction. It is necessary to analyze each construction stage and to determine the deflection curve of the successive cantilever arms as construction proceeds, step by step. A simple case with a five-segment cantilever is shown in Figure 4.77. The broken line represents the envelope of the various deflection curves or the space trajectory followed by the cantilever tip at each construction stage. By changing the relative angular positions of the various segments by small angles, such as -LY,, -(Y*, and so on, the cantilever should be assembled to its final length with a satisfactory longitudinal profile as shown in Figure 4.78, for the simple case considered. The practicalities of this important problem are covered in Sections 11.4 and 11.6.

t

b

S E C T I O N c-c

AmA

I ,

-?

2.25

SECTION B-B

I +

FIGURE 4.74. Outline of end segment over abutment.


208

3.43 SECTIOS

A-A

S E C T I O N C-C

6 30

bl.CTIOS

B-B

FIGURE

4.75.

Hinge segment \vith espmsion

joint.

COUPE A.A 20.40

&

COUPE B.B 1

FIGURE

A 4.00 r

4.76.

Saint

B P

ELEVATION J

CIOL~CI

4.00

Br-idge, hinge segment \cith expansion joint

It is interesting to compare the relative importance of deflections and camber for cast-in-place and precast construction. Figure 4.79 shows values for an actual structure, where computations have been made for the two different methods. The cal-

culational assumptions given in Figure 4.79 indicate that in most cases the difference would b e even more significant if a cast-in-place cvcle of less than one week were emploved and if precast segments were stored for more than two weeks. Hove-

Segments N’

ENVELOPE OF DE’YECTION CURVES

FIGURE

4.78.

Choice and control of camber.

FIGURE

4.77.

Deflections of a typical cantilever.

t Cl _z

CROWN

45.00

SUPPORT

d+5

d+b

db24

d+24

c

I de7 d+25

i

d*8 dt25

~

d+9

~ de26

/

.

I

1

I

d+lO

d+ll

d+26

d+27

dtl2 ‘dc27 I I I I

1

E 2 L -.-

ASSUMPTIONS

I /

PRECAST _ __ __ __ __

:

in

:

g

Casting : one segment per day Placing : two segments par day Segments at least 2 weeks old for

E Y LL x

3 -

placmg

CAST-IN-PLACE : , -----_---_______ Casting : one segment per week P r e s t r e s s i n g : 30 days after casting

- 4

-

-L-

FIGURE

4.79.

Comparison of deflections between precast and cast in place structures. 209


210

ever, one would normally expect a cast-in-place cantilever arm to resist deflections two or three times greater than the precast equivalent. 4.15

Fatigue

in

Segmental

Bridges

Basically, prestressed concrete resists dynamic and cyclic loadings very well. Eugene Freyssinet demonstrated this fact fifty years ago. He tested two identical telegraph poles under dynamic loading. One was of reinforced concrete and the other of prestressed concrete; both were designed for the same loading conditions. The reinforced concrete member failed after a few thousand cycles, while the prestressed concrete member sustained the dynamic load indefinitely (several million cycles). Fatigue in concrete itself has never been a problem in any known structure, because a variation of compressive stress in concrete may be supported indefinitely. When reference is made to fatigue in prestressed concrete, it is always inferred that fatigue problems arise in the prestressing steel or conventional reinforcing steel as a result of cracking due either to bending or to shear. If cracking could be avoided in prestressed concrete structures, the fatigue problem would be completely eliminated.

Figure 4.80 shows the resistance to fatigue of prestressing strands currently used in prestressed concrete structures. The diagram shows the limit of stress variation causing fatigue failure versus the mean stress in the prestressing steel. For convenience, both values are expressed as a ratio with respect to the ultimate tensile strength. For a steel stress of 60% of the ultimate the acceptable range of variation is 28% of the ultimate for a number of cycles between lo6 and 10’. Using, for example, 270 ksi quality strand, this variation is therefore 222,000 psi or a total range of 44,000 psi. Because dynamic loading on a bridge is of a short-term nature, the concrete modulus is high and the ratio between steel and concrete moduli is of the order of 5. Consequently, the maximum concrete stress in an untracked section that would cause a fatigue failure would be 44,000/5 = 8800 psi, a value which is probably ten times the stress variation under design live loads in highway box girder bridges. An untracked prestressed concrete structure is therefore completely safe with respect to fatigue, regardless of the magnitude of live loads. A limited amount of cracking, although considered unadvisable f-rom a corrosion point of view, is not critical if kept under control. Tests and experience show that a grouted prestressing tendon can transfer bond stresses up to

.6 -

4-

Stress variation causing failure

fs ? Afs

2-

FIGURE

4.80.

Resistance to fatigue of prestressing strands.

Fatigue in Segmental Bridges

500 psi to the surrounding concrete. Taking the example of a typical (twelve 3 in. diameter strand) tendon with an outside diameter of 2.5 in. (64 mm), a stress variation of 40,000 psi in the steel produces a tendon force variation of 73,000 lb (33 mt), and the bond development length across a x 2.5 x 7r) = 18 in. (0.46 crack is then 73,000/(500 m), see Figure 4.81. The ,corresponding crack width l is equal to the elongation of the prestressing steel between points A and B with the triangular stress diagram-that is, 40 ksi over an average length of 18 in., or

40 ' = EL = 26,000 x 18 = 0.028 in. (0.7 mm) A safe crack width limit of 0.015 in. (0.4 mm) can be accepted to eliminate the danger of fatigue in the prestressing steel. In fact, instances of fatigue in segmental structures are extremely few and far between. An isolated case has been reported of a bridge in Dusseldorf, Germany, where failure occurred as a result of fatigue of prestressing bars. The cast-inplace structure was prestressed with high-strength bars coupled at every construction joint. After ten years of service, a joint opened up to # in. (10 mm) and caused bar failures at the couplers. An investigation revealed that a bearing had frozen and prevented the structure from following the longitudinal movements due to thermal variations. This accidental restraint induced high tensile stresses in the concrete and caused cracking, which first appeared in the construction joints precisely where bar couplers were located. The live-load stress level in the prestressing steel increased from 850 psi (6

I nw,” f J

07./d’

FIGURE 4.81. Fatigue in prestressing steel across a cracked section.

211

MPa) for the previously untracked section to 14,000 psi (96 MPa) for the cracked section and induced failure in the bars. A recommendation was made as a result of this fatigue problem that couplers should be moved at least 16 in. (0.40 m) away from the construction joints and that reinforcing steel should be provided through the joints if practical. Another sensitive factor relating to fatigue in web reinforcing steel was mentioned in Section 4.10.2 for reinforced concrete test beams. No such danger would exist in prestressed concrete if shear and diagonal stresses were kept within the limits that control web cracking. In conclusion, fatigue in prestressed concrete is not a potential danger if design and practical construction take into account a few simple rules: 1.

Avoid bending cracks in girders by allowing no tension or only a limited amount at either top or bottom fibers for normal maximum loads, such as the combination of dead loads, prestressing, and design live loads including moment redistribution and half the temperature gradient.

2. Avoid web cracking by keeping diagonal tensile stresses within allowable limits by proper web thickness and possibly vertical prestress. 3. Design and maintain bearings and expansion joints that allow free volume changes in decks. Temperature stresses that cannot be controlled can give rise to enormous forces that may either tear the deck apart or destroy the piers and abutments. In this respect, elastomeric bearings, which work by distortion and cannot freeze, are safer than friction bearings, which are more easily affected by dust and weathering of the contact surfaces. Insofar as crack control in segmental structures is concerned, it is usually felt in Europe that excessive concrete cover over the reinforcing steel and prestress tendons does not prevent corrosion but merely increases the crack width.3 For example, the typical 2 in. (50 mm) cover commonly used in bridge decks in the United States is considered extreme in Europe. The 4 in. (100 mm) cover for concrete exposed to sea water would be a complete surprise to European engineers. Several examples of common practice in segmental bridges are given as a simple comparative reference in Table 4.2.

212


TABLE

4.2. Concrete Cover to Reinforcing Steel and Prestress Tendons in Europe

Concrete (in.) Germany 1) to 2 1t lf France 1 1t 2 Netherlands lb 1;A 2 to 2;R

4.16

cover

Description Reinforcing steel Outside exposure, tendons Inside exposure, tendons Transverse reinforcing steel Longitudinal reinforcing steel or tendons (normal atmosphere) Corrosive atmosphere (salt water) Reinforcing steel and tendons (normal exposure) Lightweight concrete Salt water exposure

Provisions for Future Prestressing

For larger segmental bridges, it may be necessary to modify the prestress forces after construction. An example would be a bridge built using cantilever construction where positive-moment (continuity) tendons are added after erection. Or, as discussed in Section 4.8.6, some tendons may be released to articulate a joint. In addition to these adjustments immediately after construction, addi-. tional prestressing may be required at a later date to correct for unanticipated creep deflection or for additional loads such as for a new wearing surface. In Europe on some bridges spare tendon ducts are provided for this reason. A reasonable assumption would be to provide for 5 to 10% of the total prestress force for possible future addition. Since the tendon anchorages for the spare ducts are inside the box girder and generally located at the web-flange fillet, they are readily accessible. If future prestressing is needed, it is only necessary to insert the required tendon in the duct, jack it to its designed load, anchor and grout it. Since all this work can be done inside a box girder, it is not necessary to interrupt traffic, and the workmen are fully protected.3

4.17 Design Example The Houston Ship Channel Bridge now under construction in Texas, U.S.A., is an outstanding example of segmental construction and represents

the longest box girder bridge in the Americas as of this writing. Typical dimensions were given in Section 2.14. This section will deal with some design aspects of this prestressed concrete segmental bridge. 4.17.1 LONGITUDINAL BENDING Each of the four identical cantilever arms is made up of: Ten segments 8 ft long (maximum weight 4 15 kips) Six segments 12 ft long (maximum weight 464 kips) Thirteen segments 15 ft long (maximum weight 457 kips) Longitudinal tendons are as follows: Cantileuer t e n d o n s : 4 2 ( n i n e t e e n 0 . 6 i n . d i a strands) + 50 (twelve 0.6 in. dia). Twelve additional bars used during construction are incorporated in the permanent prestress system. Continuity tendons in side spans: dia).

20 (twelve 0.6 in.

Continuity tendons in center span: dia).

40 (twelve 0.6 in.

A typical layout of the cross section was given in Figure 2.82. The main loading combinations considered in the design are summarized in Table 4.3. The lonTABLE 4.3. Houston Ship Channel Bridge, Main Design Load Combinations

Loading Case

Description

(1)

(G) + (P) + (E)

(2)

CD) (D) (D) CD)

(3) (4) (5)

+ + + +

Allowable Tension on Extreme Fiber, Top or Bottom (ksf)

(P) + CL + 1) (P) + (L + I) + l(AT) + (T) (I’) + t(L + I) + (AT) + (T) (f’) + (W

0 2: 25 25

Notations: (C) girder load, (D) total dead load including superimposed dead load, (L + I) live load plus impact, (P) prestress, (E) construction equipment, (AT) temperature gradient of 18°F between top and bottom fiber, (T) temperature and volume changes, (W) wind load on structure. Concrete strength and stresses:rC = 6000 psi = 864 ksf (42.1

MPa). Basic allowable compressive stress: 0.4fi = 346 ksf (16.8MPa).

213

Design Example

gitudinal bending of the box girder has been analvzed using the BC program, which considers the effects of the creep, shrinkage, and relaxation at each construction phase. Figure 4.82 shows the diagram of prestress forces due to cantilever and continuitv tendons at two different dates: After completion of the structure and opening to traffic (780 days after start’of deck casting) After relaxation and creep have taken place (4000 davs) Significant values of the prestress forces are given in ‘Table 4.4. The variation of stresses in the center and side spans is shown in the following diagrams for the corresponding loading cases: Figures 4.83 and 4.84, all dead loads and prestress at top and bottom fibers Figures 4.85 and 4.86, live load and temperature gradient at top and bottom fibers It is easily shown from these diagrams that all stresses in the various sections are kept within the allowable values mentioned in Table 4.3. The

WEISYT

FIGURE

O F O N E TRLwLm

:

130 I(. (mmt)

4.82. Houston Ship Channel Bridge, typical segment layout and longitudinal prestress.

TABLE 4.4. Houston Ship Channel Bridge, Significant Values of Prestress Forces

Prestress Force (kips) Maximum cantilever prestress in side span Maximum cantilever prestress in center span Maximum continuity prestress in side span Maximum continuity prestress in center span

Day 780

Day 4000

Percent Loss

54,710 51,310

6.2

54,390 49,280

9.4

9,540

8,760

8.2

18,130 16,780

7.5

maximum compressive stress at the bottom fiber level appears in the section located 124 ft from the pier and is equal to 335 ksf under the combined effect of all dead and live loads and prestress. 4.17.2 REDISTRIBUTION OF MOMENTS

The exceptional size of the structure gives rise to a moment redistribution of particular importance. The BC program allows a complete analysis of the behavior of the structure under the separate and combined effects of loads and prestress; also the effect of concrete creep and steel relaxation can be considered separately. Figure 4.87 shows the variation of stresses at top and bottom fibers along the center span between days 780 and 4000, which correspond to bridge opening date and the time when materials will have stabilized (concrete creep and shrinkage having taken place and prestress having reached its final value). The magnitude of the variation is remarkable, particularly at bottom flange level where it exceeds IO0 ksf (700 psi or 4.90 MPa). To isolate the effect of concrete creep on moment and stress redistribution, a section near midspan may be analyzed where cantilever prestress is neglibile. Results for the section located at a distance of 352 ft from the pier are summarized in Figure 4.88. The redistribution moment is equal to 52,000 ft-kips. It is interesting to compare this result, obtained through the elaborate analysis of the BC program, with the result of the approximate method outlined in Section 4.8.7. Figure 4.89 shows the moments in a typical cantilever under girder load and final prestress. The prestress moment has been computed using a reduced eccentricity obtained by


214

!

fc

TOP

TOP FIBER

o(-)

HIDSPAN

4.83. Houston Ship Ship Channel Channel Bridge, Bridge,top topfiber fiber prestress prestress for (LX.) (LX.) tt (P) (P) atat time time 780 days and 4000 days. Stresses at top fiber of the deck. Dead load at time 780 days \vhen the bridge is just opened to traffic and at time 4000 days. FIGURE

transforming the steel area in the concrete section. Therefore, the prestress moment is equal to: Pe( 1 - 7zP) where e = geometric eccentricity, n = 10, transformed coefficient, p = percentage of prestress steel in the section (varying between 0.5 and 0.7%). The total midspan moment produced in the continuous span with fixed ends under the combined effect of girder load and final prestress is equal to 84,000 ft-kips. Therefore, the actual redistribution moment obtained by the BC program is equal to: 52 000 ) = 62% of the total moment 84,000 The recommendation given in Section 4.8.7 to take a ratio of 2/3 gives a satisfactory approximation.

4.173 STRESSES AT MlDSPAlV

Because of the moment redistribution the bottom fiber near midspan is subjected to increasing tensile stresses while the top fiber is always under compression. It is therefore sufficient to consider the state of stresses at the bottom fiber after creep and relaxation. The results are shown in Table 4.5. It is instructive to compare the relative magnitude of the various factors influencing the stresses at midspan (stresses in ksf at bottom fiber): 1. Live load 2. Moment redistribution (difference between 250 for CL and 159 for prestress) 3. Temperature gradient 4. Temperature fall

44 91

48 18

215

Design Example TABLE 4.5. Houston Ship Channel Bridge, Stresses at Midspan Bottom Partial

Stresses (ksf) .Moment redistribution due to GL Slomenr r e d i s t r i b u t i o n d u e 10 presr ress Uoment redistribution due to (GL) + (P) All dead loads and all final prestress (from BC program including moment redistribution)

Fiber

Cumulative

are light in comparison with those used in other countries, particularly in France and Great Britain. These two factors tend to increase the importance of moment redistribution in relation to the effect of loads computed in the conventional manner.

+250

4.17.4 S H E A R

-159 + 91

Live load + impact ~Teniperarure gradient, AT = 18°F Temperature tall,

-66 -

The variation of shear stresses along the center span under design loads is given in Figure 4.90 together with the corresponding longitudinal compressive stress at the centroid. The most critical section is located 187 ft from the pier centerline. The numerical values in this section are as follows:

44

1.

48 18

T = -40°F Loading combination (‘L),” (D) + (P) + (L + 1) Loading combination (4),’

-22 Max +22 ( 2 5 )

(D) + (P) + $(L + I) + AT + T “See loading combinations in Figure 4.85. Tombination differential is

of Maximum AT + T (maximum temperature improbable in winter).

2. 3.

Vertical dead-load shear force: 4350 kips. Resal effect: the compressive stress at the centerline of the bottom slab is 192 ksf and the angle with the horizontal is 0.055 radians. Bottom slab area: 53.5 sq ft. Resal effect: 192 x 53.5 x 0.055 = 570 kips. Net dead-load shear: 3780 kips. Live-load shear force: 430 kips. Corresponding shear stresses in this section: I/Q = 14 ft web thickness b=4ft

The influence of the temperature fall (effect 4) is imputable to the frame action between deck and piers and would not appear in a conventional deck resting on its piers with flexible bearings. Considering only the other three factors combined, as in loading combination (4) of Table 4.3, the maximum tensile stress at the bottom fiber of the midspan section is: 9 1 +44+48= 1 5 9 k s f 2The live-load stress is only 44 ksf or 44/l 59 = 28 percent of the total. In all good faith, a design engineer would have completely overlooked effects 2 and 3 only a few years ago and consequently underdesigned considerably the continuity prestress. The situation has now completely changed, and the knowledge of materials together with the powerful tool of the computer allows segmental structures to be designed safely and realistically. It is as well to remember that the Houston Ship Channel Bridge is of exceptional size (which tends to increase the importance of dead load and moment redistribution) and that American live loads

Total shear stress under design load (no load factor) : V = 3780 + 430 = 4210 kips Shear

stress:

4. 5.

4210 v = ~ = 75.2 ksf 14 x 4 Longitudinal compressive stress:f; = 160 ksf Vertical prestress. The contract specifications called for a vertical prestress for the entire deck giving a minimum compressive stress of:

6.

3q = 232 psi = 33.5 ksf Verification of allowable shear stress. Using the formula proposed in Section 4.10.4: u = 0.05fi + 0.2Of* + 0.40fy

the allowable shear stress is: Vlll,, = 0.05 x 864 + 0.20 x

160 + 0.40 x 33.5 = 88.6 ksf while the actual shear stress is only 75.2 ksf


216

fc

B O T T O M Fl0ER

BOTTOM

w

FIGURE 4.84. Houston Ship Channel Bridge, bottom fiber stresses for (DL) + (P) at time 780 days and 4000 days. Stresses at bottom fiber of the deck. Dead load at time 780 days when the bridge is opened to traffic and at time 4000 days.

7.

Principal stresses at design loads for the state

of stress:

u = 75.2,

fJp =

160,

and

fu =

33.5 ksf

The two principal stresses are 3 (tension) and 195 (compression). The angle of the principal stress with the horizontal is given by: tan p = 0.466

If vertical prestresses were not used, the principal stresses would become: -30 (tension) and 190 (compression) 8.

Principal stresses at ultimate stage. For the load factors 1.30 + 2.17L, including the effect of prestress, the ultimate shear force is: V, = 5710 kips

Corresponding shear stress: VU = 102 ksf Principal

pression).

stress: - 23 (tension) and 217 (com-

Direction of the principal stress given by: tan p = 0.56

Web shear cracking at this level of stress would be unlikely. Assuming that the concrete carried none of the ultimate shear across the potential crack shown in Figure 4.91, the total shear load should be resisted by the vertical tendons and the conventional stirrups acting on a length equal to:

‘x Q

1 -=&=25ft tan/3 .

The unit force per foot of girder is therefore:

2 17

Design Example

TEMPERATURE t 18-F

L IVE

\

/

MAXI MINI

//

TOP FIBER

\

GRADIENT/

LOAD

\

/

\ \ L IVE

\

375 FIGURE

4.85.

375 FT

FT

LOAD MAxI

4

Houston Ship Channel Bridge, top fiber stresses for (L + I) and (AT =

18°F). 5710 - = 228 kips/lineal ft 25

shear force per unit length of girder to be carried across the crack is:

The ultimate capacity of tendons and stirrups is: Tendons in three webs Stirrups-O.88 in.Vineal per web at 60 ksi

ft

220 kips/lineal ft 158 kips/lineal ft 278 kips/lineal ft

1 x 5710 - x 0.5 = 240 kips/lineal ft 0.85 0.14 The corresponding amount of steel (grade 60) would be for each web:

L,2!&

The condition V,/C#I < V, becomes: 228 - = 268 < 378 kips/lineal ft 0.85 and is easily met. If no vertical prestress had been used, the slope of the shear crack would be: tan /3 = 0.487 Using the limiting value tan /3 = 0.5 instead of the actual value (as explained in Section 4.10.4), the

3

1.33 in.*/lineal ft

This amount of steel would still be reasonable (0.7%). 4.17.5

DESIG,V

OF THE CROSS-SECTION FRAME

Owing to the magnitude of the project, particular attention was given to this problem. Five finite element analyses were performed to analyze: The local effects in the transverse frame,


218

$BOTTOM

L I V E

L O A D [::,:

40 KSF

1

LIVE LOAD

“Axi 1 I

L I V E L O A D MAxi

,/’ \ TEMPERA&7E

\\

\

/I 375

FIGURE = 18°F).

4.86.

G R A D I E N T (+18-F) ,-j I

FT

375

FT

H I DSPAN

\

‘.

- A - - - -

Houston Ship Channel Bridge, bottom fiber stresses for (L + I) and (AT

The possible differential deflections three webs of the box section,

between

the

The relative behavior of sections close to the piers or at midspan, The effect of diaphragm restraint near the pier. The dimensions of the cross section at midspan are given in Figure 4.92 with the nine critical sections where moments and axial loads were computed for as many as fourteen loading combinations. A typical set of results is shown in Figure 4.93 for the midspan section. For the section located 187 ft from the pier centerline (already considered for maximum shear stresses), the moments and axial loads are substantially the same as for the midspan section. Excluding the vertical prestress, the most critical loading arrangement gives the following

values at the upper section of the outside web (section e of Figure 4.92). Moment 1 I .9 kip-ft/ft Axial load 5.4 kip/ft The steel section required at design stage for grade 60 steel stirrups is 0.34 in.2/lineal ft. Applying the recommendations of Section 4.10.4 for the simple case of a section without web prestress, the requirements for steel on both faces of the web would be: For shear of the longitudinal member: 3

x

1.33 = 0.67 in.2/lineal ft

For bending of the transverse member: 0.34 in.2/lineal ft

Quantities of Materials

A

f

219

TOP GIRDER LOAD

STRESS VARIATION AT TOP FIBER

Af

D z

B O T T O M PRESTRESS

STRESS

VARIATION

A T BOTTOM

8

FIBER

c_-----v_

Af

FIGURE 4.87. relaxation.

BOTTOtl

GIRDER LOAD

Houston Ship Channel Bridge, variation of stresses due to creep and

.The minimum area should thus be the higher of the f’ollowing values: 0.67 + 1 x 0.34

= 0.84 in.*/lineal ft

1 x 0.67 + 0.34

= 0.67 in.*/lineal ft

0.i(0.67 + 0.34)

= 0.71 in.2/lineal ft

In the actual structure, the stirrups in this section are #6 bars at 12 in. centers, giving on each face a steel area of 0.44 in.* together with the minimum vertical prestress of 44.2 kips/lineal ft (average compressive stress of 230 psi). ‘I‘he ultimate capacity of the section reinforcement is theref-ore:

With vertical prestress: 378/3 = 126 kips/lineal ft Without vertical prestress: kips/lineal ft

2

x

0.84

x

60 = 101

4.18 Quantities of Materials Before closing this chapter, it is interesting to give some statistical results concerning the quantities of materials required in segmental box girder bridges. Unit quantities have been computed by dividing the’ total quantities for the bridge superstructure by the deck area, using the total width of the prestressed concrete structure. The

220

Design of Segmental Bridges Stresses, ‘Top Fiber (ksf)

Srresses,

Bottom Fiber- (list)

780 Days

4000 Dqt

780 Dny.\

4000 lkJ\

Cantilever Prestress Girder + superimposed dead load

- 6.36 - 56.93

130.32 -266.50

-20.20 61.89

- 161.08 293.50

Total Variation from 780 davs to 4000 davs

-63.29

- 136.18

4 1.69

132.42

Loading Case

+9o.i3

-72.89

No& I: .I‘ensile stresses are positive. Note

2: This moment is the difference between girder load, 142,000 tt-kips, and cantilever prestress, 90.000 t’t-kips.

f, = (I =

72.89

Corresponding moment variation: AM = (f, ffd +

4774 FZ4)

= (72.89 + 90.73) F

-in

AM = 52,000 ft-kips

II 4

2 cd II c? A I+? fz= ++ 90.73 fz= FIGURE 4.88. Houston Ship Channel Bridge, analysis of section at 352 ft from pier.

average concrete quantity per span foot varies with the span length. For each structure considered, the span length used is the average span of the various two-arm cantilevers. The longitudinal prestressing steel is given in pounds per cubic yard of deck concrete versus the same span length. It is assumed that prestressing tendons are made up of strands with 270 ksi guaranteed ultimate strength. From the charts given in Figures 4.94 and 4.95, it may be seen that the average quantities of materials ma) be represented by the following approximate formulae: Concrete (ft3/ft2)

= 1.0 + (L/250)

Longitudinal prestress (lb/ft”) spans up to 750 ft)

2- 1.0 + (L/60) (for

4.19 Potential Problem Areas As with any type of construction with any material, problems arise that require the attention of not only the designers, but contractors and subcontractors as well. No matter how good the design, if FIGURE 4.89. Houston Ship Channel Br-idge, computation of moment redistribution.

rapid

Potential Problem Areas

I60 .----

I

221

fx(I

- . .\ \ ‘3

140 60

120 IO0

40

EF / PG

80 60

i .20

40 20 20

I 24

FIGURE 4.90. Houston Ship Channel Bridge, variation of web shear stress and average compressive stress in center span under design load.

the structure is not properly constructed, there will be problems. Conversely, no matter how diligent the contractor, if the design details are poor, problems will result. Obviously, if the design and the construction are poor, problems are compounded.

8 d 1 110"

I’

13s’

7

i

I

FIGURE 4.91. Houston Ship Channel Bridge, shear and principal web stresses in section 187 ft from Pier (under design loads).

di FIGURE 4.92. Houston Ship Channel Bridge, design of transverse frame at midspan.


222 Section M, dead load M, Prestressing

M, DL + PIT

M, live load with 1 M 1 maxi M, DL + PIT +

+I

LL

N, dead load N, transverse prestressing

N, DL + PIT

-5?67 16.59 10.92 -5.24

b -6.29 13.22 6.93 -6.68

-0.92 1.45 5.03

d -6.05 8.01 1.96 -8.82

4.11

- 1.75

7.98

-9.51

11.87

0.06 50.75

-0.53 51.06

-0.65 51.26

-0.59 51.35

50.8 1

50.53

50.61

2c37

h 2.14

-2.88 ~ 1.25

2.14 0.35

0.06 - 5.23 -0.78

-4.55

-4.50

2.59

-6.25

4.24 -0.31

6.08 -0.31

0.55 -0.29

50.76

3.93

5.77

0.26 -

0.37 -

0.24

50.76

1.10 5.36

5.77

0.26

0.37

0.24

N, live load

N,DL +PIT+LL +I

50.8 1

50.53

50.61

0.08

0

i -5.29

Web vertical prestress is not included.

,Vote:

ab 4: II i F4 \-=- - -------. \ I

\

f -\ B---------

c 1 9

SIGN

1

3.

II‘ensile cracks behind tendon anchorages, particularly for high-capacity continuity tendons in the bottom flange of box sections.

4.

Transverse cracking or opening of Joints, or adjacent thereto, due to the combination of several factors such as:

Compressive axial forces are positive. Positive bending moments cause tension at the broken line face.

a.

Underestimation of moment redistribution due to concrete creep. b. Thermal gradients in the box section.

I

c. Warping of segments due to improper curing procedures.

CONVENTION

FIGURE 4.93. Houston Ship Channel Bridge, moments and axial forces in transverse frame at midspan.

Problems are generally associated with quality control, poor design details, or a lack of understanding as to how the structure will behave, either through ignorance or because a particular phenomenon is unknown to the current state of the art, or a combination of all these factors. The following list of problem areas, as they are known to the authors, is presented so that those involved in designing and building segmental bridges may take adequate measures and precautions to avoid these problems. 1.

f -3.15 0.22 -2.93 - 1.25

-:96

1e22 3.01 4.23 5.88

Improper performance of epoxy due to mishandling of mixing and application procedure, particularly in rain and cold weather. The consequences are largely reduced by the use of adequate shear keys in webs and in both top and bottom flanges of the box section.

2. Grout leakage between adjoining ducts at joints between segments, particularly in precast segmental construction. Conformity of the ducts at the joints is a desirable feature if practical. The use of tendons outside the concrete eliminates this problem.

5.

6.

Several such points have been already a d dressed in this chapter; others are discussed in Chapter 11. Should the recommendations given be followed both in design and construction methods and in supervision, no more difficulties of this nature are to be expected. Laminar cracking in deck slab or in bottom flange due to wobble and improper alignment of ducts at the joints between ad-jacent segments. Such incidents have been experienced more often in cast-in-place construction than in precast construction. However, care should always be taken insofar as deck alignment is concerned in all segmental projects. Freezing of water in ducts during construction, especially those anchored in the deck slab (vertical prestressing tendons or draped continuity tendons).

7.

Excessive friction in ducts due to wobble. Proper alignment will reduce friction factors in segmental construction to those currently observed in conventional cast-in-place posttensioned construction.

8.

Improper survey control in segment manufacture for precast segments as well as in the field for cast-in-place segments.

i

I 100

200

300

400

500

600 AVERAGZ

FIGURE

15

I

4.94.

Average quantities of

deck

700

concrete.

T

FIGURE

600

SPAN L (ft)

/

4.95.

AVERAGE SPAN L(ft)

Average quantities of longitudinal prestressing steel.

223

224

,

Design of Segmental Bridges References

1. F. Leonhardt, “New Trends in Design and Construction of Long Span Bridges and Viaducts (Skew, Flat Slabs, Torsion Box),” International Association for Bridge and Structural Engineering, Eighth Congress, New York, September 9-14, 1968. 2. Jean Muller, “Ten Years of Experience in Precast Segmental Construction,” Journul of the Prestressed Concrete Institute, Vol. 20, No. 1, January-February 1975.

3. C. A. Ballinger, W. Podolny, Jr., and M. J. Abrahams, “A Report on the Design and Construction of Segmental Prestressed Concrete Bridges in Western Europe- 1977,” International Road Federation, Washington, D.C., June 1978. (Also available from Federal Highway Administration, Offices of Research and Development, Washington, D.C., Report No. FHWA-RD-78-44.) 4. “Effets de I’effort tranchant.” Federation Internationale de la Precontrainte, London, 1978.

Foundations, Piers, and Abutments

5.1 5.2

INTRODUCTION LOADS APPLIED TO THE PIERS

5.6.2 5.6.3

5.3

5.2.1 Loads Applied to the Finished Structure 5.2.2 Loads Applied During Construction SUGGESTIONS ON AESTHETICS OF PIERS AND

River Piers and Foundations for Choisy-le-Roi, Courbevoie, and Juvisy Bridges, France Piers and Foundations of Chillon Viaducts, Switrer-

5.6.4

land Main Piers and Foundations of the Magnan Viaduct,

5.6.5

France Main Piers and Foundations for the Dauphin Island

5.6.6

Bridge, U.S.A. Deformation and Properties of Piers with Flexible

5.6.7

Legs Elastic

ABUTMENTS 5.3.1 structure Layout 5.3.2 Aesthetics of Piers 5.4

5.3.3 Aesthetics of Abutments MOMENT RESISTING PIERS TIONS

AND

THEIR

FOUNDA-

5.4.1 Main Piers for the Brotonne Viaduct, France 5.4.2 Piers and Foundations for the Sallingsund Bridge,

5.7

5.7.1 5.7.2

5.5.3

Upstream Paris Belt Bridge, France Properties of Neoprene Bearings Notations

5.5.4

Deformations of Neoprene Bearings Deformation of Piers with a Double Row of Neop rene Bearings

5.5.5 5.5.6

Properties of Piers with a Double Row of Neoprene Bearings Influence of Thickness and Arrangement of Neoprene Bearings on the Variation of Force in a Three-Span Structure

5.6

5.7.3

Scope and General Considerations

5.5.2 Description of Structures Oberon Viaduct, France Blois Bridge, France

PIERS WITH TWlN

FLEXIBLE LEGS

5.6.1 Inttoduction

5.1 Introduction Probably the area most challenging to the civil engineer is that of foundation design and construction, presenting the largest potential dangers but

AND

Piers

with

THEIR

Flexible

Legs

STABILITY

DURING

Scope Description of Representative Structures with Tempo’;uy Supports Downstream Paris Belt Bridge, France

5.4.4 Main Piers for the Houston Ship Channel Bridge, U.S.A. PIERS WITH DOUBLE ELASTOMERIC BEARINGS 5.5.1

PIERS

of

CONSTRUCTION

Denmark 5.4.3 Concept of Precast Bell Pier Foundation for the I-205 Columbia River Bridge, U.S.A.

5.5

FLEXIBLE

Stability

5.8

Saint Jean Bridge In Bordeaux, France Review of the Various Methods of Providing Stability During Cantilever Construction

ABUTMENTS 5.8.1 Scope 5.8.2 Combined Abutment/Retaining Wall 5.8.3 Separate End Support and Retaining Wall 5.8.4 Through Fill Abutment 5.8.5 Hollow Box Abutment 5.8.6 Abutments Designed for Uplift 5.8.7 Mini-Abutment

5.9

EFFECT OF DIFFERENTIAL SETTLEMENTS ON CONTINUOUS DECKS 5.9.1

Effect of an Assumed Pier Settlement on the Stresses in the Superstructure

5.9.2 Practical Measures for Counteracting Differential Settlements REFERENCES

also yielding the most significant savings to proper design concepts or refined construction methods. The first industrial application of prestressed concrete was related to solving an insurmountable problem of foundation underpinning. 225

226


The transatlantic terminal built in Le Havre Harbor in France on the English Channel was opened for operation in 1934 to receive the new generation of fast passenger ships between Europe and America. Improper foundation of the rear bays of the new building caused immediate constant settlements at the rate of 1 in. (12.7 mm) per month with no foreseeable limit, except the total ruin of the facility, Figure 5.1. Eugene Freyssinet proposed a unique system of underpinning, which was immediately accepted and implemented, whereby prestressed concrete piles were manufactured in the basement of the existing building in successive increments and progressively driven by hydraulic jacks to reach the stable lower soil strata, found at a depth of more than 100 ft (30.5 m), Figure 5.2. This example should certainly make one cautious against excessive optimism in foundation design; at the same time it exemplifies the remarkable potential of prestressed concrete in solving unusual problems. In concrete bridges, often greater savings may be expected from optimization of foundation and pier design than from the superstructure itself. This chapter will deal with certain specific aspects of piers, abutments, and foundations for bridges built in balanced cantilever. Similar concepts may be extended to cover other construction methods (span-by-span, incremental launching, and so on).

Piers with many different shapes have been used in conjunction with cantilever construction. For example, single piers, double piers, and momentresistant piers have all been used. The cantilever segmental construction method has an important influence and bearing on the design concept of the structure. Resistance and elastic stability of piers during construction require careful investigation. Temporary piers or temporary strengthening of permanent piers or a combination of both have been used. However, the choice of piers that have adequate stability without temporary aids is highly desirable. Piers of a box section, or twin flexible legs, either vertical or inclined, are equally satisfactory. The use of full continuity in the superstructure implies that proper steps have been taken to allow for volume changes (shrinkage, creep and thermal expansion) at the supports. Bridges such as the Choisy-le-Roi (Section 3.2), Courbevoie (Section 3.2), and the Chillon Viaduct (Section 3.6) show how the use of piers with flexible legs makes it possible to achieve full deck continuity and to build frame action between deck and piers without impairing the free expansion of the structure. The converging pier legs used at Choisy-le-Roi reduce and even cancel the amount of bending transferred to the pier foundations. Vertical parallel legs such as those in the Courbevoie and Chillon

-__ iI

FIGURE 5.1. Le Havre transatlantic terminal, typical section.

227

Introduction

Jaws Ior str.efchmngE Steel Rods.

mj E n d P l a t e of Mould

Horizontal Section

b insu/at/nq

fnveiope,

Sheet Iron Internal Uould.

internal

Moo/U

FIGURE 5.2. Le Havre transatlantic terminal. (a) Vertical section and plan of composite foundation girder. (6) Details of pile mold. structures may be used on multispan structures because their additional flexibility accommodates larger horizontal displacements. For longer structures, bearings with a variable number of laminated elastomeric pads may be used to provide the desired horizontal flexibility.

If in the finished structure single slender piers are designed solely to transfer the deck loads to the foundations (including horizontal loads), the piers may be unable to resist the unsymmetrical moments due to the cantilever construction (i.e., with an unbalance of one segment and the equipment


228

load). Thus, temporary shoring is required, often at considerable cost. In some cases, the stability of the cantilever under construction has been provided by the launching gantry used for placing the segments. With double piers, two flexible legs (either inclined or vertical) make up the pier structure, which usually is supported on a single foundation. Stability during construction is excellent and requires little temporary equipment, except for some bracing between the slender walls to prevent elastic instability. Moment-resistant piers are designed to withstand the unbalanced moments during construction by providing a temporary vertical prestress between the deck and the pier cap, thus producing a rigid connection. Flat jacks are usually placed between the pier top and the pier segment soffit to permit the substitution of temporary bearings for the permanent neoprene pads. When the ratio between span lengths and pier height allows it, the rigid connection and corresponding frame action may be maintained permanently between the superstructure and piers.

I

+

42’

I

Piers do not necessarily have to be a massive solid cross section; a box section, Figure 5.3, mav be more effective and more economical. In the United States it was generally felt that a solid pier was more economical. However, for tall piers the economics of pier casting should be evaluated against the cost of the additional dead load supported by the pier shaft and transferred to the foundations. It may be desirable to precast the pier as tubular segments that are prestressed vertically to each other as well as to the foundation; this concept was used for the Linn Cove Viaduct in North Carolina and the Vail Pass structures in Colorado. In certain cases the tubular section may be replaced by an I section, Figure 5.4. However, the low resistance to torsion of this section imposes certain precautions to limit the deformation of the cantilevering superstructure during construction, in particular with respect to the effect of wind forces. For the case of a continuous structure on short stiff piers, the volumetric changes of the concrete (shrinkage, creep, and thermal expansion) compound the redundant effect of longitudinal prestressing to produce, by virtue of the rigidity of the

l

I _f/q IO’

p 13’

IO’

FIGURE 5.3. Code Bridge, box pier.

Introduction +

29’

+

u-

FIGURE 5.4. Pyle Bridge, I-section pier.

piers, bending forces that must be transmitted to the foundations, thus condemning the use of a rigid connection between the superstructure and its support. This disadvantage then requires the introduction of a continuous superstructure resting on a number of supports that permit the longitudinal movement of the superstructure (neoprene pads, teflon, and the like). However, it is necessary to insure the stability of the superstructure during cantilever construction. This may be accomplished as stated earlier by the use of temporary shoring in the proximity of the pier or by providing a temporary fixity at the pier. Another solution is the use of piers with twin slender flexible legs. The transmission of horizontal loads in the direction of the longitudinal axis of the bridge is accommodated by the legs’ flexibility. This type of pier offers three advantages: 1. Efficient fixity of the superstructure to the piers with regard to the vertical loads by the action of the separate supports,

229

2.

Large flexibility in the horizontal plane (relative to the displacements parallel to the longitudinal axis of the superstructure), permitting the resolution of the problem of expansion posed by the continuous structure, 3. Stability of the superstructure during construction by a simple temporary bracing. In the final structure, the leg flexibility is sufficient to accommodate the longitudinal braking forces. When the geometry of the structure permits, it is more economical to incline the walls in order to reduce the bending moment transmitted to the foundation. If the legs are hinged at the superstructure and if the axes of the two legs converge near the level of the foundation, the bending moment is either canceled or minimized and the distribution into the supporting soil is essentially uniform, as for a vertical reaction, Figure 5.5. This type of structure is similar to a frame or an arch. The thrust produced by the effect of a horizontal load parallel to the longitudinal axis of the bridge is translated into a tension force on one leg, which then acts as a tie beam, and a compressive force in the other leg, which then acts as a strut. For this reason it is often necessary to prestress the legs to accommodate the tension force. When the legs are vertical, they do not profit appreciably from the frame or arch action, and the stability is essentially contained in their bending resistance. For the case where the legs are hinged at both ends, no resistance is offered and it is neces-

FIGURE 5.5. Piers with flexible walls.

230


sary to stiffen a pier to provide a fixed point in the structure. Because of pier flexibility a careful analysis is required to assure the elastic stability of the structure. The legs supporting the superstructure are in effect very slender, and their resistance to buckling must be carefully examined. This type of pier structure will be examined in greater detail in the sections that follow. Another family of piers that lends itself to cantilever construction is that of moment-resisting piers with a double row of neoprene bearings between the pier top and the superstructure, such as to benefit from pier rigidity during construction or in the finished structure while allowing free expansion of the continuous deck, Figure 5.6. The proper choice of dimensions for the neoprene bearings will allow control of the amount of bending transferred to the foundation; in fact, rigid piers with double neoprene bearings behave in much the same way as piers with twin flexible legs. We see, then, that piers and foundations for cantilever concrete bridges will fall into one of the four following categories: 1. 2. 3. 4.

Moment-resisting piers either fixed or hinged to the superstructure. Moment-resisting piers with double neoprene bearings. Piers with twin flexible legs. Conventional flexible piers properly strengthened during construction to resist unbalanced loading conditions.

FIGURE 5.6. Piers with twin neoprene bearings.

After reviewing the loads applied to the piers and considering some suggestions pertaining to the aesthetics of piers and abutments for concrete segmental bridges, we shall deal separately with each of the four pier types. The chapter will conclude with a review of several types of abutments and the effect of unequal pier settlements on the stress in the superstructure. 5.2

Loads Applied to the Piers

All loads must be carefully considered in the design of the piers and their foundations, both in the finished structure and during its construction. 5.2.1

LOADS APPLIED TO THE FIAVISHED STRUCTURE

In addition to the various loading arrangements taken into account for conventional structures and used in combination as set forth in the AASHTO specifications, for example, it is necessary to include some design aspects particular to segmental cantilever construction as follows: 1. When a frame action is realized between superstructure and piers, proper transfer of moments to piers must be considered, particularly under unsymmetrical live loading. The piers are thus an integral part of the structural system and their flexibility must be first evaluated and then incorporated in the overall structural system. Figure 5.7 shows the usual parameters used to define the flexibility of a pier as the relationship between the applied loads (M, Q, and N) and the corresponding components of the deformation at the same point (0, u, and v). The four flexibility coefficients A, B, C, and K must include all components of the pier and its foundation: soil, piles (if used), footing, pier shaft (or walls), neoprene bearings (if used). Loads and deformations are taken at the level of the deck girder neutral axis. The deck construction scheme usually imposes special loads to the substructure. Piers adjacent to an expansion joint located at the point of contraflexure (see discussion of this aspect in Chapter 4) are subjected to appreciable bending moments due both to the relaxation of the hinge after cantilever construction and to live loading placed on either side of the hinge. Loads applied to the structure by the construction equipment result also in moment transfer in piers connected to the superstructure. Two typical cases often encountered are:

Loads Applied to the Piers

_

APPLIEO

231

LOADS

M,Q,N, cOIRESPOH313G

-0

i

OEFOeUATioN

,Mu,W,

AM + BQ

0=

AA r BM+CO IJ FIGURE 5.7.

=

KN Basic components of pier flexibility.

a. In precast segmental construction with segments placed with a launching gantry, the gantry leg reactions are applied to a temporary static scheme and released in another static scheme (after continuity between two adjacent cantilever arms is realized). b. In case-in-place cantilever construction, the weight of travelers is applied to the free cantilever arms during construction but it is removed from the structure after continuity is achieved. On long spans the effect on the deck is usually beneficial, but important moments may simultaneously be induced. 2. Volume changes (shrinkage and thermal variations) and long-term shortening of materials (concrete creep and steel relaxation) both induce moments and horizontal loads in the piers, which must be included in the design.

balance), the application of random loads (difference between actual and computed dead loads or wind gusts), or accidental conditions (such as the fall of a traveler). Normal Loua The most critical condition appears for one segment out of balance at the outboard end of the cantilever arm. Even in the case of cast-in-place construction with symmetrical travelers allowing simultaneous casting of both corresponding segments, the assumption of the total segment weight out of balance is a safe one, because no total guarantee can be given that concrete pouring will proceed simultaneously at either end of the cantilever. If construction equipment is designed to be installed on the deck, Figure 5.8, it

5.2.2 LOADS APPLIED DURING CONSTRUCTION

Balanced cantilever construction imposes on the piers a loading configuration that is globally symmetrical. Unbalanced conditions appear, however, as a result of intermediate construction stages (normal loads due to a traveler or a segment out of

FIGURE 5.8. Loading conditions during construc-

tion.

. Foundations, Piers, and Abutments

232

must be accounted for in the design of the pier. For example, a tower crane is often used on one side of a cantilever. Random

Loads

Random loads essentially are s u c h a s t o p r o d u c e systematic geometric difference, although within acceptable tolerances. With proper workmanship and supervision, it is reasonable to assume such difference in weight at 22%. It corresponds to a variation of top slab thickness of 2 in. (9.5 mm) for a 40 ft (12 m) wide box with a cross-sectional area of 60 ft2 (5.6 m’). However, it is very unlikely that the maximum weight decrease in one cantilever arm would appear simultaneously with the maximum weight increase in the other. It is therefore reasonable to limit the moment transferred to the pier to 2% of the maximum deck cantilever moment due to the girder weight. Other random loads related to the construction are produced by the small equipment, trucks, storage on the deck of materials such as post-tensioning tendons, and so on. An equivalent uniform load of 5 psf (24.4 kg/m*), together with a moving concentrated load of 20 k (9 mt), should be a safe allowance to cover these random loads. Taking as an example the Houston Ship Channel Bridge, which was considered in Section 4.17, the effect of these three random loads would be: difference in dead weight, 1,600,OOO ft-kips x 2% random uniform load, ( 5 x 60)/1000 x 365*/2 random concentrated load, 20 kips x 365 ft

32,000

ft-kips

20,000

ft-kips

7,000

ft-kips

59,000

ft-kips

This moment should be compared to the effect of one segment out of balance at the far end of a cantilever: 300 kips X 367 ft = 110,000 ft-kips One last source of random loading is provided by gusts of wind that apply an uplift pressure or suction to the box girder intrados during construction. For long spans and construction sites exposed to hurricanes, it is desirable to make special aerodynamic tests. For an incident angle of 10” above the horizon, the upward pressure would be 5 psf (0.2394 MPa) during construction. This value may be substantially increased in exposed sites. For construction of the Gennevilliers Bridge, a

maximum pressure of 9 psf (0.4309 MPa) corded in the wind-tunnel tests.

was re-

Accidental Loads These are the result of a construction incident or of human failure, causing either the fall of a traveler in cast-in-place construction or of the lifting equipment in the case of precast construction. Such loads should be multiplied by a factor of 2, representing the impact coefficient for the case of immediate loading. It is never envisaged to consider the fall of a cast-inplace segment and traveler after casting, nor the fall of a precast segment immediately after its placement in the structure. A very long record of safety in such construction methods justifies that approach. However, in the case where the consequences of such major accident would be exceptionally disastrous (where, for example, the work takes place over a highway or a railway under operation), special provisions should be incorporated in the design and in construction procedures to double all safety features at each step of erection.

5.3

Suggestions on Aesthetics of Piers and Abutments

The problem of aesthetics is subjective and controversial. There is, however, a consensus among engineers, owners, and users that certain bridge structures are more pleasing than others. At a time when so much emphasis is being placed on protection of our environment and of nature from aggressive man-made structures, it may be helpful to review some ground rules based on experience that contribute to aesthetics of concrete bridges with very little added cost. 53.1

STRUCTURE

LAYOUT

Generally speaking, an attempt should be made to match the structure to the environment and to preserve the existing landscape. Avoid long, high embankments at the ends of the bridge as well as long, high retaining walls that accentuate the intrusion of the new structure. Allow the number and shape of the piers to maintain a maximum of transparence. Cost optimization of superstructure span lengths will normally help to avoid serious aesthetical mistakes. It is equally disgraceful to see a heavy, long-span superstructure rampant over the ground as a multitude of closely spaced, high piers supporting a slender deck floating up in the air. The true appearance of a structure is usually not

233

Suggestions on Aesthetics of Piers and Abutments

conveyed by the drawings, where often a distorted scale is used for convenience. Finally, it is very important to keep the unity of appearance of a structure crossing different obstacles, in spite of the practical difficulties that may be entailed when project coordination involves different owners or agencies. When an overpass crosses, for example, a freeway and a parallel railroad track, nothing may be worse than to build two separate structures (probably of different height) connected by a short embankment contained at‘ both ends by wing walls of variable height, Figure 5.9.

FIGURE 5.10. Piers for the Broronne duct.

,~pp~o;~ch

via-

5.3.2 AESTHETICS OF PIERS

A significant advantage of segmental construction is to allow deck continuity, rather than simply supported structures. There is no longer a need for heavy bents protruding underneath the superstructure soffit. Piers can have simple graceful lines and be designed to receive directly the box girders of the superstructure. Box piers of prismatic section but with curvilinear shapes improve the appearance over the conventional rectangular section. The approach piers of the Brotonne Viaduct, Figure 5.10, utilized that concept and also the piers for the Linn Cove Viaduct in North Carolina. More refined shapes may be used, such as for the river piers of the Blois Bridge, Figure 5.11, where the sculpture of the faces was designed to recall the appearance of a pier with twin inclined walls similar to that of the Juvisy Bridge, Figure 3.25. Architectural studies may be pursued further and reach beyond the immediate structural needs of the designer. An interesting example is afforded by the river piers

FIGURE 5.9. An unacceptable example of’ an overpass built as two separate structures.

of the railroad bridge at Clichy near Paris, Figure 5.12. A difficulty arises often for skewed bridges when bents include multiple pier shafts. A satisfactory solution was developed for the Paris Downstream Belt Bridge, Figure 5.13. The four columns of a river pier are given the shape of a lozenge, with one axis of symmetry matching the alignment of the superstructure while two of the four faces exactly align the four columns in the direction of

the river flow.

FIGURE 5.11. Blois Bridge.

Piers with architectural shapes for

234

Foundations,

Piers, and

Abutments

FIGURE 5.12. Piers for Clicln Railroad Bridge.

FIGURE 5.13. Piers for a skew bridge (Paris Ring Road).

When the piers will be seen only from a great distance, it is usually not worthwhile to call for a special treatment of the concrete faces. The eye will judge only the general shape of the structure and its overall proportions. For urban bridges the situation is very different and often justifies some architectural treatment of the piers. The river piers of the Saint Cloud Bridge were cast with a system of closely spaced vertical grooves, which greatly enhance their appearance at very little added cost, Figure 5.14. 5.3.3

AESTHETICS

OF

ABUTMENTS

At both ends, the structure has to blend with the existing landscape with a minimum of disturbance. Between the two systems of wing walls shown in Figure 5.15, the preference should strongly be with type (a), which allows a much more gradual transition between the lines of the superstructure and those of the approach embankment. When tapered webs are used in the superstructure box girders, it has been found that the lateral wing walls in the abutments can be given the same

FIGURE 5.14. Saint Cloud Bridge. (CL) (b) General view.

River piers.

inclination to improve the transition between deck and abutments, Figure 5.16. 5.4

Moment-Resisting Piers and Their Foundations

We shall cover this topic by describing salient features of several characteristic structures. 5.4.1

MAIN PIERS FOR THE BROTONNE VIADUCT, FRANCE

The two main pylon piers for the Brotonne Viaduct rest on 41 ft (12.46 m) diameter cylindrical

235


Wing walls parallel to bridge Q

-.__.-. -.---.-.-.-.-

-

!I - _-.-.-

-

__.-.__ -.-.-.-.Wing walls perpendicular to bridge F.

(b)

FIGURE 5.15. Wing walls and abutments.

:i ” ::.*i .: columns with a maximum wall thickness of 9.3 ft (2.83 m) and are 115 ft (35 m) below ground level in a limestone stratum overlain by alluvium, silt, and gravel beds. The maximum reaction at footing level is 19,000 tons. Typical dimensions of a main foundation syst.em are shown in Figure 5.17. It was decided to select the theoretical foundation level at 115 ft (35 m) below the original ground level, where the limestone bed had the following minimum characteristics determined from laboratory soil tests and in situ tests: angle of internal friction 20”, cohesion 5 tons/ft2, and a pressure limit (on triaxial tests) of 45 tons/ft2. The foundation system had to resist very large loads (both vertical and horizontal) together with important overturning moments. The main foundation column embedded in the soil and resting on the lower limestone stratum was analyzed as a rigid body subjected to the applied loads (M, V, and H) shown in Figure 5.18 and receiving from the soil lateral reactions along the shaft and vertical reactions under the base. Values of lateral and vertical reactions were ascertained for the various soil strata and the equilibrium was determined by considering the total body to be subjected to an angle of rotation cy around the in-

.*\q*>

-.

.,I

\. ,

. >..:~ ..:.\ ‘,;,., ,.j \‘;:i\ ;\;cA. :& \ . .>;.y

FIGURE 5.16. Inclined wing walls in end abutment (Bordeaux St. Jean Bridge).

stantaneous center of rotation C. The coordinates of point C are the following: Vertically, it represents the level where lateral reactions from the soil change sign (change from direct passive pressure on the front face to counterreaction at the back face). Horizontally, it is the position of the neutral axis for the stress under the base. The maximum loading configuration is represented numerically in Figure 5.18 along with the diagrams for: Lateral reactions on the column Bending moments along the column Bearing stress under the base If there were no lateral support, the bending moment at the base would have been 370,000 ft-kips. In fact, the actual moment is only 130,000 ft-kips,

236


FIGURE 5.17. Brotonne Viaduct, pylon foundations.

which explains why the extreme fiber stress is no more than 24 tonsIft while the average bearing pressure is 14.25 tons/ft2. The actual safety factor for the foundation against soil failure is between 3 and 4, depending on the assumptions of soil characteristics. Insofar as the construction method is concerned, each main foundation column was built in the dry inside a cofferdam made up of a continuous slurry trenched concrete wall excavated down to the limestone stratum, Figure 5.19. Grouting of the base allowed dewatering of the site after excavation to inspect the foundation material and confirmation of the actual soil characteristics by in situ soil tests. Following this inspection, the cofferdam was flooded and a tremie seal was placed at the base to prevent any risk of washing out of the footing concrete due to water seepage; the water head was above 100 ft (30 m). The reinforced concrete footing was cast in the dry above the seal and the foundation shaft was then slip-formed inside the cofferdam. The pier shaft was given the shape of an octagon with curvilinear sides for aesthetic reasons. The general dimensions of the foundation shaft and of the pier shaft allowed a very natural

and direct transfer of loads at ground level with no need for a heavily reinforced footing. The construction of both foundations went very satisfactorily. The only incident was created by the fact that one panel of the cofferdam in the south pier was excavated out of plumb at its lower end. Consequently, the continuity of the horizontal ring to resist the hydrostatic pressure was not realized at the lower part of the cofferdam. Grouting of the surrounding soil was achieved in this area and an additional reinforced concrete ring was cast inside before the completion of excavation and final dewatering. Regular survey measurements at the site have shown that settlements of both pier foundations have been very minimal and are now stabilized. 5.4.2

PIERS AND FOUNDATIONS FOR THE SALLINGSUND BRIDGE, DENMARK

The substructure and piers of this structure present an interesting construction methodology and use of materials, Figures 3.89 and 5.20. The piles are steel tubes, which are concreted after driving. Their length is about 98 ft (30 m), the diameter is


n. 230.050

FT. k. v, 19.000

237

t

TSF

1

“““‘J

FIGURE

of

main

5.18.

Brotonne Viaduct, loads and soil reactions on column foundations.

274 in. (700 mm) and the wall thickness is about 0.4 in. (10 mm). Each pier has 24 piles. The first piles driven are tested in compression and tension be-

FIGURE 5.19. Brotonne Viaduct, view of pier exca-

vation.

,.--Jre

fore the remaining piles are driven. When the driving is accomplished, the template trough is filled with tremie concrete around the pile tops up to the upper edge of the template. The template is precast at a plant located in the harbor. It is shaped like a circular slab surrounded by an annular trough, in which there are holes for the piles. The template is transported to the pier locations by the floating crane and lowered down to rest on three temporary vertical piles. The bottom is about 52.5 ft (16 m) below the water level. For an exact positioning in its submerged position, it is provided with an alignment tower, the top of which is always above water, Figure 5.21. The pier box, shaped like a truncated cone approximately 39.3 ft (12 m) high, is precast in three lifts at the precasting plant in the harbor. First its lower part is cast on staging above water. During the following lifts it is progressively sunk. Since after the third stage it is too heavy to be lifted by the floating crane, it is provided with a lid, and com-


238 Assembled

ure 5.23. The finished bridge is shown in Figure 5.24.

pier

5.4.3 CO,VCEPT OF PRECAST BELL PIER FOC,VDATION FOR THE I-205 COLL’,WBI,4 RIVER BRIDGE, C’.S.A.

/-y-7

Gmcmting

rT e m p l a t e

Il.3

Concreting of piles

concrete

FIGURE 5.20. Sallingsund

A somewhat comparable system to that used for the Sallingsund Bridge was contemplated for approach spans 15 through 26 of the I-205 Columbia River Bridge in the State of Oregon, as shown schematically in Figures 5.25 and 5.26. Steel H piles of 200 ton capacity were to be driven through a template box, allowing tremie concrete to be placed inside the trough. The precast segments were designed to be stacked upon one another above the template to make up the pier shaft and transfer the superstructure load to the piles. This scheme was not actually used, as the contractor decided on a more conventional method of construction. However, the scheme of precast bell pier foundations was used on the Richmond-San Rafael Bridge and the San Mateo-Hayward Bridge, both in San Francisco Bay, and the Columbia River Bridge at Astoria, Oregon. A comprehensive discussion of these structures is presented by Gerwick in reference 3.

plug

Bridge. schematic of sub-

StrllCtlll-e.

pressed air is pumped into the cavity. The floating crane then transports the pier box to the pier location and lowers it down to rest on the template. A reinforced concrete ring structure is made by connecting the pile tops to the pier box by reinforcing and concreting the space between them, Figure 5.21. The icebreaker’s shell is a reinforced concrete box, precast at the harbor site, Figure 5.22, transported to the pier location by means of the floating crane and placed on top of the pier box. Its top is then 8.2 ft (2.5 m) above and its bottom 8.2 ft (2.5 m) below the water level. When the box is in place, the water in the cavity of the pier box and the icebreaker box is pumped out. Next, the piles are filled with concrete and the pile tops and the lower part of the pier box are cast together. Finally the cavity of the icebreaker is filled with concrete. A schematic sequence of operations in constructing the substructure is shown in Figure 5.21. Piers are cast in place in lifts 10 ft (3 m) high by means of climbing forms and are hexagonal, Fig-

5.4.4

MAI,: PIERS FOR THE HOUSTOS SHIP CHA,V,VEL BRIDGE, U.S.A.

Each main channel pier, Figure 5.27, is made up of the following: A rectangular shaft 161 ft (49 m) high with a cross section varying in dimensions from 20 X 38 ft (6.10 x 11.60 m) at the base to 20 x 28 ft (6.10 X 8.50 m) at the top. The section is a single-cell box with wall thicknesses of 2 ft (0.61 m). A reinforced concrete footing 75 X 81 X 15 ft (22.90 x 24.70 x 4.60 m). A group of two hundred and twenty-five 24 in. (0.61 m) diameter steel pipe piles having a wall thickness of 4 in. (12.7 mm). The superstructure is completely integral with the two main channel piers to form a rigid frame, both during construction and in the finished structure, Figures 1.67 and 2.80. Stresses in the concrete and reinforcing steel were analyzed in both stages with the service-load design approach, a n d u l t i m a t e s t r e n g t h w a s verified by the load-factor method. The analysis is


239

PLACING OF TEMPLATE

dteel pik

TEST LOADING

PILE DRIVING

PLACING OF PIER BOX

PLACING OF ICE BREAKER BOX

CONCRETING OF STEEL PILES AND FOUNDATION

FIGURE tions.

5.21.

Sallingsund Bridge, schematic of substructure opera-

rather strenuous, because in the completed struc-

ture only there were 19 unit loads combined into 37 load combinations for service-load design and into 42 loading combinations for load-factor design. The concrete cross-sectional area together with the corresponding reinforcing steel area is as follows: top:

A, = 176 ft2, A, = 200 no. 11 bars = 297 in.*, p = 1.17% bottom: A, = 216 ft2, A, = 264 no. 11 bars = 392 in.*, fi = 1.26%

Under service load the average concrete stress of

the cross section is as follows: top: bottom:

3 1,700 kips + 176 ft* = 180 kips/ft* 36,600 kips + 2 16 ft2 = 170 kips/ft*

In large structures, such as the Houston Ship Channel Bridge, the average concrete stress in the pier shafts usually varies between 160 and 200 kips/ft*. The use of a varying-width pier in the transverse direction allows the maximum stress and the required amount of reinforcing steel to increase at a slow rate with the pier height, while a prismatic pier shaft will be subjected to a very critical stress at the base.

FIGURE 5.22. Sallingsund Bridge, aerial view of precast yard and harbor for substructure construction. CLIMBING

FORM

Tower crane

Ice-breaker-

:’

.

Pierbox

CROSS SECTION

0 04Om

3 50m 55Om

FIGURE

5.23. Sallingsund Bridge, schematic of pier construc-

tion

FIGURE 5.24. Sallingsund Bridge, view of finished bridge.

241

Piers with Double Elastomeric Bearings

TYPICAL PIER 5810’

,

TYPICAL I_ S E G M E N T 5 _ I

I

I

MER

PROFILE

116’-2” CAST-IN-PLACE CONCRETE

DESIGN HIGH WATER ELEV. 28.0’ -

SEGMENT

4

SEGMENT

3

//

\\

I

r

SEGM SEGM

FIGURE 5.25. I-205 Columbia River Bridge, main piers and foundations.

P I E R ELEUTION

PLA.h S E C T I O N B.B PLAN SECTION 4.A

PRECAST BELL PIERS

SEGMENT

4

FIGURE 5.27. Houston Ship Channel Bridge, main river piers.

5.5 5.5.1

SEGMENT

SEGMENT

SEGMENT

FIGURE 5.26. I-205 Columbia River Bridge, schematic of construction of precast bell piers.

Piers

with

Double

Elastomeric

SCOPE AND GEAVERAL

Bearings

CONSIDERATIO,VS

Recognizing the inherent advantages of a rigid connection between piers and superstructure (stability during construction and increased superstructure stiffness reducing the effect of live load), the designer is rapidly limited in its use in long bridges because of unacceptable effects of volume changes. This situation allowed the birth of a new type of structure developed to maintain the two desirable features that were previously contradictory: flexural rigidity on one hand and horizontal flexibility on the other. The concept of the double row of elastomeric bearings was first developed for the Oleron Viaduct and used thereafter on a great many bridges.

242


With piers of this type, two observations are required concerning the transfer of forces between the superstructure and pier. The first observation concerns the transfer of service loads, Figure 5.28~. Under the effect of unsymmetrical loads, the upper and lower flanges of the superstructure are respectively subjected to unequal tension forces TL and T, and compressive forces CL and CR. If a vertical diaphragm is positioned over each of the two rows of bearings, the center portion of the top flange of the pier segment, to be in equilibrium, must accept the tension force TL - TR. This is not a satisfactory disposition, as the thickness of the flange and amount of reinforcing have to be increased between the two rows of bearings, and there is the risk of cracking. However, if the two diaphragm: are inclined and converge at the level of the top flange, the differential in tension, T, - T,, is divided into two components of force, C (compression) and T (tension), directed into the plane of the diaphragm, while the tension force may be accommodated by prestressing the diagonal bracings. Another important aspect of the pier segment loading condition design relates to the imbalanced resulting at the bottom flange from the unequal reactions R1 and R, of the bearings, which calls for careful analysis of the stress developing in the diagonal bracings in all loading stages of the structure. The second observation concerns the superstructure-pier connection during the temporary phase of constructing the superstructure in cantilever, Figure 5.286. To accommodate a moment unbalance resulting from the construction procedure, the pier segment is supported on four temporary bearings of steel or concrete, 0, and temporarily fixed by prestressing to the top of the pier, 0. After closure at midspan occurs, producing a continuous span, the joint is “unlocked” by releasing the prestressing. Flat jacks, 0, are then activated so as to substitute permanent bearings for the temporary bearings. 5.5.2 DESCRIPTION OF STRUCTURES

Many structures have been designed and built utilizing the system of piers incorporating a double row of neoprene bearings. This section will describe the salient features of three particular bridges as exemplifying the advantages of this system as used in connection with a variety of foundation schemes.

TL -

Neoprene bearings

CO++

i‘3

Flatjacks

Steel banded concrete block

fb) FIGURE 5.28. Connection of superstructure and

pier. (a) In service. (6) In temporary construction phase.

Piers with Double Elastomeric Bearings Oleron Viaduct, France

Of the 45 piers, only the 27 piers supporting the center portion of the viaduct with span lengths of 260 ft (79 m) are designed with a double row of bearings. In this portion of the viaduct there is an expansion joint every fourth span, and the elastomeric bearings had to accommodate the volume changes of the deck in a maximum distance of three spans (i.e., 780 ft or 237 m). Out of these 27 piers equipped with a double row of bearings, 12 are founded on spread footings constructed directly on limestone rock inside a temporary sheet pile cofferdam, Figure 5.29. The other 15 piers are supported by a system of pipe piles driven to the limestone, which in this area is at a depth of 75 ft (23 m) below mean water level, Figure 5.30. The 12 piles in each pier consist of four vertical piles, one at each corner, and eight battered piles, so inclined as to resist the horizontal loads (longitudinal and transverse) applied to the structure. For the most critical loading combination (comparable to the AASHTO requirements) the maximum load in a pile is 330 t (300 mt), which should be reduced to compare to American prac-

FIGURE 5.29.

243

tice by a factor of 1.33. The comparable design load would then be 250 t (230 mt) for a pipe pile 20 in. (500 mm) in diameter with a thickness of 3 in. (12.7 mm) driven to refusal in the rock and filled with concrete after driving. The corresponding steel stress of the pipe alone would be 16 ksi (110 MPa), a somewhat higher value than normally used in similar circumstances. When considering the global section of concrete and steel, the stress in the concrete is only 800 psi (5.5 MPa)-a very reasonable value, confirmed by the fact that none of the 15 piers showed any sign of settlement during the fifteen years of operation of this viaduct. The pipe piles were driven open-ended and excavated inside by a homemade airlift system conceived by the driving subcontractor. It took only a few minutes to perform this operation on each pile. For the piers on piles, a tremie seal was used inside the cofferdam to allow dewatering and construction of the reinforced concrete footing poured in the dry. All box pier shafts were slip-formed to a maximum height of 82 ft (25 m) at the rate of 15 to 20 ft (4.5 to 6 m) a day, and the construction of a shaft took approximately one week, Figure 5.31.

Oleron Viaduct, piers on spread footings.

Vertical

0.30

1.30 .

.-.

.

.

..’

5

.

6

0

-

.,

p-i-

+

Treme concrete

,‘.. I,-* I.I’ : 9’ c

r FIGURE 5.30. Oleron Viaduct, pier-s on piles. 244

7.30 l

prestresslng

The special feature of this project is that a very comprehensive optimization study of the substructure system with a double row of bearings allowed the use of only half as many piles as the basic scheme with single bearings, without increasing the unit bearing capacity of the piles. Upstream Paris Belt Bridge, France

FIGURE

5.31.

Oleron Viaduct, aerial view of founda-

tions. Blois Bridge, France The Blois Bridge crossing the Loire River is a fivespan, prestressed, precast concrete segmental superstructure consisting of twin box girders with the following span dimensions: 202, three at 300, 202 ft (61, three at 91, 61 m). It is supported by f o u r r i v e r p i e r s elastically restrained at the superstructure with a double row of bearings. Dimensions of a typical pier are given in Figure 5.32 and a view of a finished pier in Figure 5.11. 65' t

I

I

t

I

This important bridge was built over the Seine River to carry Europe’s most heavily traveled urban freeway, the Paris Beltway. As shown in a longitudinal section, Figure 3.22, it has two major river piers resting on a unique foundation system, while land piers and abutments are conventionally founded on piles. A typical transverse section of the bridge shows the orientation of the piers, Figure 5.33, and various cross sections through the piers is shown in Figure 5.34. Each of the twin bridges carries four lanes of traffic on two box girders, which are supported on two separate pier shafts connected below water by a single footing. Two lower foundation shafts extend under this footing to a maximum depth of 70 ft (21 m) to carry the bridge loads to the supporting soil strata through a series of heterogeneous seams of silt, fine sand, and clay. Each of these lower shafts (there are eight such shafts for the two river piers) w; s built inside a rectangular steel sheet pile cofferdam, driven as low as possible before excavation. The shafts were extended below the tip of the sheet piles to reach the load-bearing soil by incremental stages of excavation and continuous concrete lining, Figure 5.35. Cement grouting and temporary lowering of the aquifer by pumping allowed this work to be performed in the dry. Except for the minor blowout in one of the eight shafts, which called for special grouting work, the foundation project was performed safely and successfully. Figure 5.36 shows one of the river piers completed and receiving the precast pier segment of the superstructure. 5.5.3 PROPERTIES OF NEOPRENE BEARINGS Notation A neoprene bearing may be designated by the following physical parameters, Figure 5.37a: = plan dimensions of bearing (a < 6) = number of elastomer sheets = thickness of one elastomer sheet = thickness of the internal steel sheet (twice the external sheet) Ab = a * b = area of bearing a and b 12 t 2e

FIGURE 5.32.

Blois Bridge, dimensions of river piers.

246


EL

124

C c-

D t-

1 FIGURE 5.33.

34’

1 T

? 4’

a x 6 x n(t + 2e) 300 x 400 x 2(10 + 2) Where differing thickness of steel plates are used, the successive thicknesses of steel and elastomer are given: aXbXn( ) 300 x 400 x 2(5 + 8 + 2 + 8 + 1)

1

only with the material hardness, as indicated in Table 5.1, but also with the rate of loading. Tabulated values are for the case of slow loading; for an instantaneous loading the value of G is doubled. Vertical Deformation (Compression) Under a normal force V every lamination is subjected to a vertical shortening, v, Figure 5.376, such that: v=C

Deformation of Neoprene Bearings

The relationship between Young’s modulus (E) and the shear modulus (G) is presented in Table 5.1. The shear modulus, G, of neoprene varies not TABLE 5.1. Elastic Constants

45 50 55 60 65

3 4’

Upstream Paris Belt Bridge, typical elevation of river piers.

An example, with dimensions in millimeters, is as follows:

Hardness (IRHD ?4)

1

Young’s Modulus E (N/mm*)

Shear Modulus G (N/mm*)

1.80 2.20 3.25 4.45 5.85

0.54 0.64 0.81 1.06 1.37

t3 V z&-z

C is a coefficient that depends on the plan dimensions of the bearing and that expresses the restraint effect on the lamination by the steel plate; refer to Table 5.2. For a bearing consisting of n stacks or laminations, the value of the shortening is equal to: (5-l)

Rotational Deformation U n d e r a b e n d i n g m o ment M the upper face of each lamination undergoes a rotation 8 relative to the lower face: tI=C’&M b


247 5ECllDN

: D.-D

I

I

FIGURE

5.34.

Upstream Paris Belt Bridge, typical horizontal sections of river piers.

C’ is a coefficient that depends on the plan dimensions of the bearing and that expresses the restraint effect on the laminations by the steel plate; TABLE 5.2. bla C

0.5 5.83

0.6 4.44

0.7 3.59

0.75 3.28

0.8 3.03

0.9 2.65

TABLE 5.3. bla

0.5

C’

136.7

0.6 116.7

0.7 104.4

0.75

0.8

100.0

0.9

96.2

90.4

refer to Table 5.3. The value a is the dimension in plan of the bearing measured perpendicular to the axis of rotation, Figure 5.376.

Values of the Coefficient C 1.0 2.37

1.2 2.01

1.4 1.78

1.5 1.70

2 1.46

3 1.27

4 1.18

5 1.15

10 1.07

30 1

30

Values of the Coefficient C’ 1.0 86.2

1.2 80.4

1.4 76.7

1.5

2

3

4

5

10

75.3

70.8

66.8

64.9

63.9

61.9

60


FIGURE 5.35.

Upstream Paris Belt Bridge, detail of concrete lining of lower shafts.

For a bearing consisting of n stacks or laminations, the value of the rotation is equal to:

Horizontal Deformation (Distortion) Under a horizontal force, Q, the upper face of each lamination, relative to the lower face, undergoes a horizontal displacement u :

with a corresponding distortion u/t. For a bearing consisting of n stacks or laminations, the value of the horizontal displacement is equal to:

5.5.4

FIGURE 5.36. finished pier.

Upstream Paris Belt Bridge, view of a

DEFORMATION OF PIERS WITH A DOUBLE ROW OF NEOPRENE BEARINGS

In structures where deck and piers are rigidly fixed, it is necessary to analyze accurately the deformation of ihe various piers to incorporate their proper stiffness into the model of the total structure. This is particularly important for unsymnietrical live loading applied to one pier and for the effect of volume changes. There is a relationship between the loads applied at the top of one pier (usually at the level of the neutral axis of the deck over the pier) and the corresponding displacements at the same point that depends solely upon the mechanical properties of the pier and its foun-


249

s ~(arbj (a)

(b)

rdhep-cp

C” * hBp-‘p

(4

Cc)

‘n: -

cf)

FIGURE 5.37.

lcidely

Piers with double row neoprene bearings (Oleron Viaduct). The most used polychloroprene is Neoprene (trademark of Du Pont de Nemours).

dation, Figure 5.7. The elasticity coefficients A, B, C, and K may be computed from the material properties and dimensions of the pier. For example, a pier with constant section and the following properties: Height h, area of cross section A, Moment of inertia I Modulus of elasticity E

duces a partial fixity of the superstructure on the piers. The neoprene bearings intervene in the deformation of the pier by their normal force ( 2 M lpd) produced by the moment M, Figure 5.37~. The rotational stiffness of the neoprene bearings may be neglected. The moment M applied at the top of the pier may be divided into componentsf and m in the bearings, Figure 5.37d, such that: M =fd + 2m,

assumed to be fixed at the base onto a totally rigid foundation, has the following elasticity coefficients: A=&,

h* B = 2EI’

e

= 2vld

with:

h3 C=m,

B* h K=A-C=4EI from which: In structures where neoprene bearings are placed between piers and deck, the corresponding change in elasticity of the system must be taken into account. In fact, the presence of two rows of neoprene bearings, spaced at a distance d, pro-

In the majority of cases the quantity 2a*/C’ is small relative to d */2C.

250


Example Dimensions of the neoprene bearing: 600 X 600 mm. Spacing between the axes of the neoprene bearings: d = 2.4 m. -= b 1 a ’

C = 2.37,

C’ = 86.2

= +$ (1.215 + O.OOS)e In neglecting the second term in the parenthesis, in other words the rotational stiffness of the neoprene, it can be seen that the error is slight, of the order of 1%. Therefore: 2nCt3 ’ = pGAgzd2 M

M’=M

+Qh

where h represents the height of the pier. Because u = 0, one may write:

Accordingly, the flexibility coefficients of the neoprene bearings may be written as:

A,=(+?

increase of the moment in the bearings, Figure 5.37e. During construction of the superstructure by cantilevering, stability in the temporary construction phase may be provided by the substitution of concrete pads for the neoprene bearings and the use of a temporary vertical prestressing. By a judicious choice of neoprene thickness, it is possible to reduce the bending moments applied to the foundation. Consider a pier with a double row of neoprene bearings supporting a continuous superstructure. For a bending moment M at the top of the pier, under the effect of a loading in the superstructure with no horizontal displacement, the bending moment transmitted to the base of the pier is (Figure 5.37f):

t3

BM+CQ=O from which:

pd2 m

B, = 0

(5-4)

Cn=2LL 2P GAb

wherep represents the number of neoprene bearings per row. Therefore, if the flexibility coefficients of the pier shaft are denoted by A,, B,, C,, and K,, the total flexibility coefficient may be defined as: A =Ap+A, B =B, c = c, + c, K = K, + K, 5.5.5

PROPERTIES OF PIERS WITH A DOUBLE ROW OF NEOPRENE BEARINGS

Piers with a double row of neoprene bearings have properties similar to those of piers with flexible legs, by insuring an effective fixity for loads while allowing the free expansion of the superstructure. This fixity presents the advantage of reducing the bending moments in the spans without much

and M’ = (1 - +)M =

(1 - c,B$cn)M

=

4M

The value of the coefficient 4 varies with the thickness of neoprene pads. If it is desired to transfer no moment to the foundation at the level of the pier base, M’ = 0, the transfer coefficient 4 must be equal to 0, from which: C, = hB, - C,

(5-5)

On the other hand if the neoprene thickness becomes very large, the value of 4 tends to the limiting value of 1 and the bending moment remains constant in the pier; that is, M’ = M, Figure 5.37f. As an example, consider a pier with a constant moment of inertia, fixed at its base, with a double row of neoprene bearings and supporting a maximum reaction of 1000 tons. Pier characteristics: Assume a box section with external dimensions of 5.0 x 3.0 m and a wall thickness of 0.30 m, h = 33 m, I = 7 m4: E/&,=+=

4.71

251

Piers with Double Elastomeric Bearings = 2 x 860 x 2 x 160 x 0.24 = o 034 m 3.9 x 106

EB, = & = 77.7

nt = 34 mm

EC, = & = 1715 Four neoprene bearings are arranged in two rows at a spacing of 2.4 m in the longitudinal direction of the bridge. Dimensions of each bearing are 600 X 400 X 3(12 + 2) (see Section 5.5.3).

Flexibility of the neoprene bearing: a = 0.40 m, bla = 1.5, C = 1.7, Ah = 0.24 m*, n = 3,p = 2, t = 1.2 X 10m2 m, G = 160 t/m’, E, = 3.9 X lo6 t/m’:

5.5.6 INFLUE,VCE OF THICKNESS AAiD ARRA,VGEME,vT OF ,YEOPRE,\‘E BEARI,XIGS ON THE VARIATIOX OF FORCE IS A THREE-SPAN STRUCTURE

EB, = 0 nt EC,=E- = 915 2PGA h

Totaljlexibilit?

qf

the pier:

EA = 4.71 + 0.97 = 5.68 EB = 77.7 EC = 1717 + 915 = 2630 Elasticit?

of the pier in the structure:

EK = E (.4 - g) = 5 . 6 8 - ‘:;;;’ = 3 . 3 8

Elasticity qf’ the pier zuithout neoprene: Ek’ = E [;4 - $1 = 0.25+ = 1.18

Coe@cieut

A comparison of the constants A, B, C, and K with the number of neoprene laminations (for this example) is presented in Table 5.4. If the height of pier Were changed from 33 m to 20 m, the total neoprene thickness would correspondingly change from 34 mm to 8 mm.

of momerlt

transmission in the pier:

4 = 1 - B; - 1 - 77;76;033

= +0.03

The bending moment M’ transmitted to the base of the pier is very small (3% ofM). For the moment M’ to be theoretically equal to zero:

In order to better understand the influence of the thickness of neoprene pads, studies have been conducted to determine the variation of the bending moment in a three-span continuous structure when only the number of- neoprene laminations at the top of the intermediate piers is modified. The structure considered is a symmetric superstructure of three continuous spans supported on two identical piers; it consists of a box girder with a variable moment of inertia, whose spans are 44 m, 70 m, and 44 m. Bending moments in the superstructure and piers are calculated under the following assumptions: Superstructure fixed at the pier Superstructure partially fixed elastically at the piers with neoprene bearings with the varying lamina of 1, 2, 3, 6, or 9 (thickness 12 mm) Superstructure supported on the piers by simple supports Assumptions used in the conduct of the study are:

EC,, = 9 = 860 and the corresponding thickness of neoprene is then:

EC,=n t E 2p G A , or

nt = ‘L(EC.)PGA, E

Superimposed dead load represented by a uniform load, q = 1.9 t/m Expansion of the deck at a rate of 2 X 10e4, corresponding to an increase in temperature of 20°C. Shrinkage of the deck at a rate of 4 x 10P4, sponding to a decrease of temperature of combined with the effect of shortening and dependent deformations (creep) resulting prestressing (2 X 10m4).

corre20°C timefrom

252

Foundations, Piers, and Abutments TABLE 5.4. Number of Neoprene Lamina

Coefficient EA EB EC 1.18

5.03 77.7 2020

5.36 77.7 2325

5.68 77.7 2630

6.00 77.7 2935

6.33 77.7 3240

2.03

2.76

3.38

3.93

4.46

-0.27 Diagram of bending moment in the pier (h = 33 m)

Diagram of bending moment in the pier (h = 20 m)

+ 0.64

Applied load Sz = 4.5 t/m in the center span

Applied load S, = 6.8 t/m in the end spans Braking force F = 15 t on the superstructure, corresponding to approximately one-twentieth of the structure dead load The bending moments in the superstructure as a result of the above loads are tabulated in Tables

5.5a through 5.5~: Table 5.5a: bending moment at the top of the pier Table 5.56: bending moment at the base of the pier Table 5.5~: maximum bending moments in the superstructure This study leads us to the following observations: 1. Regarding the superstructure, the maximum moments vary little with the number of neo-

prene laminations. When the number of laminations increases from one to six, the maximum bending moment at the support decreases by 4% and the maximum positive moment in the center span increases by 10%. The extreme case of nine lamina is to be avoided because of risk of instability presented by the tall stack of neoprene (alnt < 5). Compared with a simple bearing support, the double row of bearings provides an important decrease in moment in the spans for a relatively smaller increase of moment at the pier support. 2. Regarding the pier, there exists an optimum thickness of neoprene allowing a minimal transfer of moment to the level of the foundations. In the example considered this thickness is equal to three lamina of 12 mm, which corresponds closely to the value determined in Section 5.5.4 for the case of a structure restrained horizontally.

Piers with Twin Flexible Legs TABLE

5.5~.

Bending Moment at the Top of the Pier as Function of the Bearing Thickness0 Number 0 (Fixed Pier)

Loading Superstructure D.L.,

q = 1.9

253

1

of’

Neoprene

2

Lamina

3

6

9

+ 124

+

106

+

93

+

84

+

68

+ 58

+

92

+

68

+

53

+

43

+

27

+

- 184

-

36

-

106

-

86

-

54

- 38

+ 216 60

+ -

174 30

+ -

146 13

+ 127 2

+ +

95 6

+ 77 + 20

+ 1700 - 1420

+ 1440 - 1240

+ 1270 - 1120

+ 1150 - 1030

+ 930 - 850

+ 790 - 740

? 101

+ 97 +1711 - 1367

2 93 + 1059 - 1226

k 90 + 1367 -1122

2 80 + 1105 - 924

k 74 +941 - 795

t/111

Deck

expansion, + 2 x 10-4 Deck shrinkage, - 4 x 10-4 I: moments I + ,\I ( n o L . L . ) -‘VI L.L in center span, si = 4.5 t/m L.L. in end spans, S, = 6.8 t/m Braking force, F = 15 t Maxi 1llu111 +‘M m o m e n t s I - ‘\/I “Values have heen

+2017 -1581

calculated at the intersection of the axis of the pier with the center of gravitv

T A B L E 5.5b.

19

of. the super-structure.

Bending Moment at the Base of the Pier as Function of the Bearing Thickness Number of Neoprene Lamina

Loading

0

Simple

(Fixed Pier)

Support, 2

3

6

9

t = 24 mm

13

+ 20

0

-111

- 77

- 6 0

-130

+258

+222

+154

+120

+260

4283 -188 -435

+243 -144 -198

+218 -115 - 47

+167 - 6 4 +176

+ 140 - 4 0 + 265

+260 -130 0

+ 197

- 7 4

-207

-265

-380

-400

0

+ 159 + 698 - 1243

2163 +609 -786

*167 +577 -518

-e170 +558 -550

+180 +527 -624

?I86 +591 -626

(+520) +780 -650

Superstructure D.L., q = 1.9 t/m Deck expansion +2 x 1o-4 Deck shrinkage - 4 x 10-4

-

C moments +M ( 1 1 0 L . L . )I -M L.L. in center span, S, = 4.5 t/m L.L. in end spans S, = 6.8 t/m Braking force, F = 15 t Maximum +M m o m e n t sI - M

5.4

1

62

- 31

-

- 202

-157

-129

+ 404

+314

+ 342 - 264 - 820

Piers with Twin Flexible Legs 5.6.1 INTRODUCTION

The concept of piers with twin flexible legs was first used with the first match-cast segmental bridge of Choisy-le-Roi. It was further used on several other

15

-

4

+

p r e c a s t s e g m e n t a l bridges either in France or Europe and more recently in the United States. Several examples of such structures will be described below with particular emphasis on the design and construction methods of the foundation system.


254 TABLE

5.5~.

Maximum Bending Moments in the Superstructure as Function of the Bearing Thickness

Number of Neoprene Lamina

Loading Moments 1 Center span at support Side span Moments in span

Center span, (0.5 &) Side span, (0.4 11)

0 (Fixed Pier)

1

2

3

6

9

Simple Supporl

-3125

-3060

-3020

-2985

-2925

-2895

-2660

-3105

-2960

-2845

-2770

-2635

-2545

-2055

+ 910

+ 960

+ 990

+1015

+ 1060

+ 1090

+1270

+ 890

+ 935

+ 965

+ 980

+ 1020

+ 1040

+ 1200

5.6.2 RIVER PIERS AND FOUNDATIONS FOR CHOISY-LE-ROI, COURBEVOIE, AND JUVISY BRIDGES, FRANCE These structures were described in Chapter 3. Choisy-le-Roi Bridge over the Seine This structure is composed of two parallel twin

bridges, Figure 3.3 and 5.38. Each structure has a continuous three-span superstructure in prestressed concrete with spans of 123 ft (37.50 m), 180.4 (55 m), and 123 ft (37.50 m), fixed at the center piers and forming a symmetric frame. Piers are supported on a system of steel pipe piles driven to refusal in rock. The superstructure is supported on two slender inclined legs having a thickness of 16 in. (0.40 m) and inclined to the vertical axis at 0.065. Dimensions of the substructure are shown in Figure 3.3. The precast legs with an approximate weight of 27.5 ft (25 mt) have their centerlines converging to a point approximately at the level of the foundations so as to reduce the bending moments to .the foundation. The legs are joined to the body of the pier at one end and to the superstructure at the other end by prestressing tendons. Before construction of the superstructure by the balanced cantilever .method, the legs are temporarily stiffened by a triangular steel framework in the space between them. The construction stages are described graphically in Figure 5.38.

cal spans of 131 ft (40 m), 197 ft (60 m), and 131 ft (40 m). Each river pier consists of two half-structures whose foundations are fixed in dense rock, Figure 3.9. The top portion of each half-pier consists of two vertical slender legs, oriented, in plan, perpendicular to the longitudinal axis of the bridge, and in a transverse section of the bridge, disposed in the shape of a V. These legs, which have a parallelogram form, are spaced in a longitudinal direction at 6 ft 9 in. (2.05 m) on center with a constant wall thickness of 18 in. (0.45 m). The legs were precast and joined to the superstructure and the lower portion of the pier by prestressing tendons. The Juvisy Bridge consists of six prestressed concrete continuous spans with a total length of 700 ft (213.5 m). Spans are successively from the left bank 62 ft (18.8 m), 62 ft (18.8 m), 137 ft (41.8 m), 218 ft (66.6 m), 137 ft (41.8 m), and 84 ft (25.7 ml. The two piers located in the Seine are split piers resting on a common foundation, Figure 3.26. The foundations were constructed inside a sheet pile cofferdam, which permitted the flexible legs to be fixed at the bottom and hinged at the top. The thickness of the legs varied from 24 in. (0.60 m) at their base to 16 in. (0.40 m) at the top. They were symmetrically inclined at 0.0805 to the vertical and were cast in place and prestressed. 5.6.3

PIERS AND FOUNDATIONS OF CHILLON VIADUCTS, SWITZERLAND

Courbevoie and Juvisy Bridges over the Seine This structure, 1.24 miles (2 km) in length, is a twin

The Courbevoie Bridge is very similar in concept to the Choisy-le-Roi Bridge. It consists of a continuous three-span superstructure with symmetri-

parallel viaduct overlooking Lake Leman and following a sinuous route corresponding to the contour of the hillside on which it is located, Figure 5.39. It

5

FIGURE

5.38.

Choisy-le-Roi Bridge, construction stages of foundations and piers.


256

1 (y 16-6


consists of 23 continuous spans of prestressed concrete, span lengths being 301.8 (92 m), 321.5 (98 m), or 341.2 ft (104 m). Four expansion joints divide each viaduct into sections with a maximum length of 1890 ft (576 m). The longitudinal stability of each section is provided either through a fixed bearing over the end abutment or by special fixed piers designed to withstand the horizontal reactions of the superstructure. The piers, Figure 5.40, consist of two slender vertical legs with a constant thickness of 2 ft 8 in. (0.80 m). Height of pier varies in increments of 26 ft (8 m) with a maximum height of 118 ft (36 m). Legs less than 72 ft (22 m) in height are hinged at the top and bottom. Legs over 72 ft (22 m) in height are fixed at the base and hinged to the superstructure. Because of the leg spacing there is no tension generated in the legs, so no vertical prestressing is required. During construction of the superstructure the stability of the pier is increased by temporary steel bracing anchored into the legs. 5.6.4

l-

Viaduct, general view.

MAIN PIERS AND FOUNDATIONS OF THE MAGNAN VIADUCT. FRANCE

The Magnan Viaduct consists of four continuous spans; span lengths are 413 ft (126 m), two at 433 ft (132 m), and 249 ft (76 m), Figure 2.98. The piers are constructed of twin H-shaped shafts 40 ft (12 m) on center and with a maximum height of 3 18 ft (95 m) above the valley floor, Figures 5.41~ and 5.41b. These piers are similar to slender vertical legs of variable cross section fixed at the base. Because this structure is located in an area of seismic activity, the superstructure is fixed at the west abutment and restrained transversly at the piers and the other abutment.

Ia56 -1t


5.6.5

Viaduct, pier section.

MAIN PIERS AND FOUNDATIONS FOR THE DAUPHIN ISLAND BRIDGE. U.S.A.

The Dauphin Island Bridge is an 18,000 ft (5.5 km) long structure over Mobile Bay connecting Dauphin Island to the mainland of Alabama. In order to permit ship traffic, the central portion of the structure was designed with a three-span continuous unit of 2 11, 400, and 211 ft (64, 122, and 64 m). This provided a clear shipping channel of 350 ft (107 m) horizontally and 85 ft (26 m) vertically. This project is currently (1980) under construction and is anticipated to be completed by late 1981. Each main pier of this three-span structure consists of twin, I-shaped walls spaced longitudinally at 21.5 ft (6.6 m) on center, Figure 5.42. An individual wall is 24 ft 7 in. (7.5 m) wide and is moment-connected to the single cell box girder superstructure as well as to the footing.

Piers with Twin Flexible Legs

The foundation is to be made with circular, standard sheet pile construction. Alternate pilings were detailed on the plans to be either 30 in. (0.76 m) square precast, pretensioned concrete or 54 in. (1.37 m) hollow, cylindrical, precast, posttensioned concrete. Piling will be driven to a capacity of 450 kips (204 mt) for the 30 in. (0.76 m) square pile or 550 kips (249 mt) for the 54 in. (1.37 m) cylindrical pile. A dewatering seal will be poured under water after the piles have been driven. This seal will be located 25 ft (7.6 m) below the water surface and have a thickness of 5 ft (1.5 m). After dewatering, a circular footing with a diameter of 44 ft (13.4 m) and a thickness of 10 ft (3.05 m) will be poured. The twin wall piers will be constructed from a point 10 ft (3.05 m) below the water level and reach a total height of approximately 93 ft (28 m). The design included checking of AASHTO loads and combinations, including a stream flow of 3.5 fps (1 mps). Additionally, the structure was checked at an ultimate condition for a storm wind of 200 mph (322 km/h). The load factor for this condition was taken as 1.0.

EL630

21:4*

2f.4"

i

, *

v

257

EL.315

EL.266 f

*

‘5’

*

*.

65.60'

)

i.1 J

5.66

DEFORMATION AND PROPERTIES OF PIERS WITH FLEXIBLE LEGS

The following notation is used (Figure 5.43): M, Q, W components of external load acting at point 0, m, t, n = components of load acting at the top of the leg of the pier, oriented to the axis of the leg, 8, U, z, = displacements corresponding to M, Q, N at point 0, W, (Y, /3 = displacements corresponding to m, t, n at the top of the leg, E = modulus of elasticity of the concrete leg, 1= length of the leg between points A and B, 2d = spacing of the legs at the top between points A and A’, a = cross sectional area of leg, i = moment of inertia of a leg, p0 = ad V2i dimensionless coefficient, 4 = angle of inclination of the legs with the vertical.

FIGURE 5.41. Magnan

Completed pier.

Viaduct. (n) Pier section. (h)

Identical and symmetrical legs, of length 1, are inclined to the vertical by the angle 4. The crosssectional area and moment of inertia of each leg at a distance x from the top, A or A ‘, are respectively a(x) and i(x).

258

Foundations, Piers, and Abutments 6

BRIDGE

tf HER

BRIDGE

SEGMENT

r-h-

24’7’

STEEL SHEET PILING

EL 0 0 0

L SEAL CONCRETE

SECTION

SECTION

The symbol u is designated as an equivalent area of the leg such that:

24’. 7 ” I

1 l’u!x -=1 s0 a(x) u and U, V, and W the characteristic integrals as: f PIER

UT

’dx 0 i(x)’

v=

s

w =

I--

E BRfDGt-

‘xdx s0 i(x) ’

‘x2d.x sO;(x)’

At the level of’the superstructure, AA ‘, the combined area and moment of inertia of the two legs, designated by A and I respectively, is represented bv: A = 2a and I = 2i + 2ad2

PLAN VIEW FIGURE 5.42. Dauphin Island Bridge, dimensions of main piers and foundations.

with 2d being the distance between the two legs at the top. Setting p. = ad2/2i, the combined moment of inertia of the two legs becomes I = 2i( 1 + 2p,).

Piers with Twin Flexible Legs *-A---+ mA*

FIGURE 5.43. Piers with flexible legs, notations.

The positive directions of forces and displacements are indicated by the arrows in Figure 5.43. The deformations of the pier are given by linear equations that relate the displacements of the top of the pier (0, U, v) to the applied forces (M, Q, N). Legs AB and A ‘B ’ are assumed to be connected at their ends by two rigid and indeformable sections AA’ and BB ‘. Section BB’ is assumed fixed (no translation), and the deformation equations are given by: 8=AM+BQ u=BM+CQ v=KN where A, B, C, a n d K represent deformation coefficients of the legs. Force components M, Q, N acting at point 0 (center of AA’) are the resultant of the external forces applied to the pier, and 8, U, u are the corresponding components of displacement of the section AA’ at point 0 (Figure 5.436). To determine the forces m, t, n and m’, t’, n’ in the legs atA and A’ requires the formulation of the equations of equilibrium, deformation, and compatibility. The equilibrium of the 1. Equilibrium equutions: system about point 0 is given by

M = m + m’ + d sin 4(t + t’) - d cos +(n - n’) Q = (t + t’) cos 4 + (n + n’) sin 4 N = - (t - t’) sin 4 + (n + n’) cos 4 2.

(5-6)

Deformation equations: Displacement o, (Y, P and w’, cr’, p’ at pointsA and A’ (with respect to the axis of the legs) are given by: w=w,+ a=o,,l+

s

‘m + t x mU oTdx=o,+E +tvE

mV tw ‘m + t x -xdx=cq,l+so Ei E + E

p=pY=lr EU

(5-7)

0 a

where w. is the rotation of the leg AB at B, and E is the modulus of elasticity of the concrete. Corresponding equations give the displacements o’, (Y’, /3’ at point A’. Displacements of points A and A’ with respect to the axis of the pier, 8, A, p and 0’, A’, p’ are determined as 8=6J A = a cos 4 + p sin 4 p = Q sin 4 + p cos

4


260

Legs hinged at both ends A’ = CY’ cos 4 - p’ sin p’ = CY’ sin

4

+ j3’ cos

4 4

3. Compatibility equations: The conditions of compatibility between the displacements of point A, A ‘, and 0 require that exoEw’

(if there are no hinges atA andA’) (5-9)

p) = 7l + de ‘The foregoing equations are sufficient to calculate 8, U, and 11 as fSunctions of the applied loads represented by ,M, Q, and ,V. Four practical cases need to be considered: Legs fixed at both ends Legs fixed at the superstructure and hinged at the base Legs hinged at the superstructure and fixed at the base TABLE

5.6

For any of these four cases the legs may be of constant or variable cross section, either inclined or vertical. A comprehensive study was made of this problem by J. Mathivat and reported in references 1 and 2, with several complete derivations of formulas applying to each particular case. An important practical application is that of twin vertical walls with constant cross section, for which equations become very simple. Table 5.6 summarizes the value of the global equivalent coefficients of elasticity of the pier. In this case p,, = nd2/2i, which becomes p0 = 6(dlh)2 with 2d the distance on centers of both legs and h the wall thickness. Usually p0 varies between 30 and 80. It is evident, in fact, that a pier made up of twin legs behaves much in the same way as a conventional pier with a cross-sectional area A and a moment of inertia I insofar as the effect of vertical loads and moments on vertical displacements and rotation is concerned. The behavior is completely different when considering the horizontal displacement due to the application of a horizontal load (braking force or thermal expansion). The conventional value of the

Flexibility Coefficients of a Pier with Twin Vertical Walls of Constant Cross Section” End

Flexibilit\ Coefficient E.wct Fonttui~i,s .4

5lultiplier Coefficient

Fixed stop and Botrom

Fixed .I‘op Hinged Bottom

Hinged .I‘op Fixed Bottom

Hinged ‘I‘op rind Bottom

1 El

1

I+1 2P,t

1+1

1+1

0

0

1 + 2Po

x

1

B

C

.4pproxitttntr Forttt~clnsh ‘4

Conditions for Legs

13 3EI

1+-p

2P”

2 PO

(1 + $)(3 + 2/J,,)

I El

“Notation: I = 2i(l + 2p,), equivalent global inertia of twin walls. p,, = nci2/2i thickness. *When l/p, is negligible with regard to 1.

= C(~/IZ)~,

with 2~f distance between walls, h wall

Piers 1:’

elasticity coefficient C = -

3EI

with

Twin

dimensionless factor 1 + po/2 in the case of vertical walls fixed top and bottom or by (1 + 2 pO) for walls hinged at one end. The elasticity coefficient becomes infinitely large for double-hinged vertical walls, which proves simply that stability toward horizontal loads must be obtained through some other restraint in the structure such as fixed connections or elastomeric bearings over the abutments. A detailed study of several typical cases was conducted for the Choisy-le-Roi Bridge, considering in particular:

The position of the point of contraflexure in the pier varies very little when the pier is subjected to a moment only; it is considerably more sensitive to the effect of a horizontal load. The horizontal rigidity of the pier varies appreciably with ttie degree of fixity of the legs. 5.6.7 ELASTIC STABILITY OF PIERS WITH FLEXIBLE LEGS It has been shown that the use of twin Hexible legs (whether vertical or inclined) provides an economic solution to the dilemma between rigidity for bending versus rotation and flexibility for horizontal load versus displacement. In this respect the elastic stability of the system is the limiting factor, because there must always be an ample margin against buckling. Assume the bridge superstructure to be displaced horizontally by 11 under a random horizontal load. The resistance against such displacement is offered by the pier rigidity, including the bending resistance of the legs if they are at least partially fixed at the top or bottom and possibly including the horizontal rigiditv of the bearings over the abutments. The minimum value of the vertical reaction in the pier (or the normal force in the legs), for which the imposed displacement does not have a tendency to spontaneously diminish until the cause provoking the displacement vanishes, represents the critical buckling load of the pier. This critical load is generally smaller than that where the legs are considered Isolated and subjected to the same load conditions.

Legs fixed at both ends Legs hinged at both ends Legs fixed on top and hinged at the base Table 5.7 presents the essential results of this studv, which also includes consideration of the flexibility of the body of the pier to the base of the foundation, where: MO = bending moment in the superstructure at the pier section (side of the center span), M, = bending moment in pier (top section), Q = horizontal reaction in the pier. The following conclusions may be drawn from the study: The superstructure is very efficiently fixed over the river piers by the twin inclined wall system. The end moment for the center span totally fixed at both ends would be 255. The actual end moment varies between 230 and 232 (i.e., 90% of the fixed end moment).

TABLE

5.7.

Choisy-le-Roi Bridge: behavior of River Piers under Horizontal and Vertical Load@

Unit Vertical Load in Center Span

Flexibilit) Coefficients Type of Legs Fixed Fixed/hinged Hinged

261

Legs

The elasticity of the pier depends very little upon the conditions of fixity of the walls at the top and bottom (0.92 to 1.03).

multiplied by the

IS

Flexible

Unit Horizontal Load Applied to Deck

Unit Volume Change

A

B

c

Elasticity El,

M,,

iv,

Mll

iM1

iv,,

4.06 12.7

54.6 234 -

973 4670 -

0.92 0.98 1.03

-232 -231 -230

-157 -154 -150

+3.4 +5.1 +6.3

+5.7 +a.7 + 10.7

+7.4 +6.4 +5.9

iv,

Q

+24.7 2.4 +21.5 1.3 + 19.7 0.9

“Notation: A, B, C = flexibility coefficients of pier. E L = global elasticity of pier. M, = end moment of center span (in tm). ,M, = bending moment at pier top (m tm). Q = horizontal reaction in pier. *Units: All coefficients in metric system. A uniform vertical load of 1 t/m is applied over the center span. A unit horizontal load of I t is applied at deck level. A unit shortening of the deck is applied such that EA = lo?


262

The deformations (8, U) produce internal forces (m, t, n and m’, t’, n’) in the top of the legs, which require the following conditions:

t, = t;,

m , = mi,

72, = -n;

If R. represents the rigidity of the superstructure against rotation and R, toward longitudinal displacements, and if M and Q represent the moment and horizontal force that the superstructure transmits to the pier, we have:

M = - RoO f h, b, n,) Q = - R,u gh nd

R,(a cos 4 - P sin 4) g’(n,, tJ

Normal force: no + nl B e n d i n g m o m e n t : m, t, Transverse force: The additional forces m, and t, may be expressed as a function of the displacement of the legs (w, (Y) and of the initial force rzo. By substituting these forces, as functions of (Y and o, into equations 5- 11, we obtain a system of linear equations in three unknowns, n, a, w. When we assume that the displacements (a, w) are different from zero when the cause inducing the displacement vanishes, the determinate form

Case

Number

5.8.

where r is a dimensionless coefficient which may be related to the usual Euler formula for buckling: ncr

x2

The example of the Choisy-le-Roi Bridge will again be considered. Seven typical cases were investigated with either vertical or inclined legs and different leg end restraints. Also the horizontal restraint of the bridge over the abutment was varied. Table 5.8 summarizes the results for the following numerical values: Wall length 1 = 8.50 m, on center spacing 2 d = 2.00 m Area a = 6.40 m2, moment of inertia i = 0.085 m4 Neoprene pads over the abutments: area A b = 1.28 m2, E/G = 20,000 The first six cases are hypothetical assumptions used for comparison. Case 7 is the actual case of the Choisy-le-Roi Bridge with the legs hinged at the base and fixed to the superstructure.

Choisy-le-Roi Bridge: Elastic Stability of Twin-Flexible-Legged Pier for Various Support Conditions

Conditions of Legs at River Piers

1

Hinged vertical legs Vertical legs hinged at the base and fixed at the top

6 7

T2Ei

*A& r

2 s I

4 5

=

with h equal to the effective buckling length. Thus the equivalent buckling length of one leg as part of the total pier system will be:

(5-l 1)

with aw = (Y sin 4 + /3 cos 4 and /3 = (IIEc)n,. The condition of initial load of the leg (expressed by no) is modified from the case of the displacement imposed to the structure and becomes:

TABLE

Ei ncr = r2l2

(5-10)

These equations may be transformed to substitute the deformations of the superstructure (0, U) for those of the legs:

Ref(m,, t,, n,)

of the three equations is nil, which allows us to obtain the value of critical load nIc. The critical buckling force of one pier leg may be expressed as:

Vertical legs fixed top and bottom Legs inclined 6.5%, hinged at base, fixed at top Legs inclined 6.5%, hinged at base, fixed at top

(actual case of Choisy-le-Roi)

Support Condition at Abutments Rigidity neglected Rigidity neglected Five neoprene pads

Three neoprene pads Rigidity neglected Rigidity neglected Three neoprene pads

A c

Factor of Safety 0

2.G 1.20 1.00 1.00 0.88 0.97

1.1 2.8 4.0 4.0 5.2 4.8

Flexible Piers and Their Stability During Construction

The designer should be aware that the following three factors play an essential role in the elastic stability of the structure: Inclination of the legs to the vertical Horizontal rigidity of the neoprene bearings at the abutments Fixity conditions of the ends of the legs in the piers The fundamental difference between cases 2 and 6 (Table 5.8) indicated by the considerable increase in the factor of safety (1.1 to 5.2) is due to the introduction in case 6 of the arch effect of the inclined legs. Horizontal displacements of the superstructure cannot occur without mobilizing the bending stiffness of the pier assembly. For case 2 the elastic stability relies solely on the bending stiffness of the legs, and the critical buckling force is the same as for a beam fixed at one end and free at the other. 5.7

Flexible Piers and Their Stability During Construction 5.7.1

SCOPE

In the preceding paragraphs we considered piers having a bending capacity allowing the deck cantilever construction to proceed with no further strengthening. Such moment-resisting piers are usually joined to the superstructure to benefit from the frame action, both to reduce the cost of foundations and minimize the effect of live loading in the superstructure. Another type of substructure remains to be considered here, one more conventional in design and where the piers receive the vertical reaction of the superstructure through a single row of bearings. Such piers are usually flexible, and the stability during cantilever construction requires that temporary supports be added to the self-bending strength of the pier shaft. 5.7.2 DESCRIPTION OF REPRESENTATIVE STRUCTURES WITH TEMPORARY SUPPORTS Downstream Paris Belt Bridge, France

The four river pier shafts previously described and illustrated in Section 5.3.2 rest on a reinforced concrete substructure built inside a cofferdam sealed with tremie concrete. Dimensions are shown in Figure 5.44.

263

Because of the limited dimensions of the pier shafts and their consequent marginal bending capacity, a temporary support was used during construction for stability of the superstructure before deck continuity was achieved. Only one support was used for each pier, Figure 5.45, on one side of the concrete shaft within the space available inside the temporary cofferdam. Consequently the lever arm between the pier and support centerlines was only 8.5 ft (2.40 m), so that a heavy reaction was imposed on the temporary support. The maximum reaction computed for the case of one precast segment out of balance, including the lifting equipment, was 1170 tons (1060 mt). Including provisions for random loads and the added reaction of the temporary prestressing tendons, the maximum design reaction in the support was 2030 tons (1840 mt). Each temporary support consisted of: A 40 in. (1 m) steel pipe filled with concrete, Figure 5.46, resting on the spread footing of the ‘permanent pier A V-shaped concrete frame placed upon the pipe and allowing the deck reaction to be transferred directly from the box section webs to the pipe Vertical prestressing tendons were also anchored in the pier footing and stressed from deck level to prevent accidental overturning of the cantilever, although limitations were imposed during construction to always start segment placement on the side of the temporary support. Temporary connection between the pier segment and the concrete pier shaft included one looped tendon and four high-strength bars. An immediate consequence of the high vertical reaction imposed upon the deck by the temporary support in case of unbalanced loading was a reversal of shear stresses between the temporary and the permanent supports. This situation was even more critical because of the permanent draped tendons, shown in the detail of Figure 5.47, located in that zone together with the Resal effect produced by the inclined bottom flange. The corresponding shear stress in the webs reached a maximum of 680 psi. Two special tendons (twelve 3 in. diameter strands) were placed on either side of each web of the box girder to reduce the shear stresses to allowable values. In fact, these four tendons worked as a tension tie between the top and bottom flanges of the box girder across the distance between the permanent and temporary support.

SECTION

TRANSVERSAL

l l ln

KOLW ----PLE UP -

cl+44

/' d-P-Y E -Tif i

f

PLAN VIEW

FIGURE

HOFUONTAL SECTION

5.44.

Downstream Paris Belt Bridge, dimensions of river piers.

-SEGMENT -MAX.

WEIGHTS: 60 to

STATICAL

VERTICAL

SUPPORT

IN

40

t

SUPPORT

U360 t

42.40 In-.-____c .-

PROVISIONAL __~ -.~

REACTION

PRESTRESSING .-

\PRESTREZfSING

RODS

FIGURE 5.45. Downstream Paris Belt Bridge, schematic of temporary support and stability of river pier during construction. 264

:

265

Flexible Piers and Their Stability During Construction STEEL

P/T

CAP N M

ANCHORS

FLANG

JOINT

Saint Jean Bridge In Bordeaux, France

FIGURE 5.46. Downstl c;m Paris Belt Bridge, details of temporary support. (a) Dimensions of support. (b) View of support.

This problem has been described at some length to show that a single temporary support subjected to high loads may call for a rather complex arrangement to satisfy all requirements of stability and resistance of all parts of the structure at each construction stage.

For aesthetic reasons the river piers were designed as rather slender shafts, which had to accommodate an important variation of the waterline due to tidal effects in the mouth of the Garonne River. The bridge was relatively low above the water, particularly at high tide. E a c h p i e r s h a f t w a s f o u n d e d o n a n opendredged concrete caisson anchored in a bed of sand and gravel of good quality, overlying a deep formation of marl and clay. Dimensions of the piers and foundations are shown in Figure 5.48. The caisson had a cuttingedge diameter of 18 ft 4 in. (5.60 m) and the maximum average bearing pressure on the sand and gravel bed was 8.1 t/ft2 at the time of first loading; the foundation settlement was a maximum of 1.1 in. (28 mm) and the long-term additional settlement was negligible, 0.16 in. (4 mm). Construction of the piers called for the use of an auxiliary floating platform that could be raised on eight temporary pipe piles, comparable in principle to the legged jacking platforms used on offshore work, Figures 5.49 and 5.50. The reinforced concrete caisson was floated into place, suspended from the platform resting on its legs, and incorporated into the permanent structure. As excavation proceeded inside the caisson to lower it to its final elevation, precast segments were added to increase the height of the caisson wall as required.

SHEAR AT SECTION OF TEMPORARY SUPPORT v, 1ooot

t

TRANSFORMED

/

CONCRE

TEMPORARY

TRANSFORMED

SECTION

:

BEARING PADS

L%m

FIGURE 5.47. Downstream Paris Belt Bridge, detail of loads on cantilever and temporary support.

FIGURE 5.49. copposite). St. Jean Bridge in Bordeaux, schematic of construction of river piers.

,

266

RAIN

CONCR~

TREMIE

CONCRETE

FIGURE 5.48. St. Jean Bridge, in Bordeaux, dimensions of river piers.

TING -

C - L A -M

FLEXI

c

R / C CAl%ON BORDEAUX

-

FLOAT5

-

PLACING

RIVER

CAlSON

_

ELEVATION low- 72.

FLOATING CLAM 5HELL C R A N E V_ERTICAL

F’PE

PIL

267


268 P

5.7.3 REVIEW OF THE VARIOUS METHODS OF PROVIDING STABILITY DURING CANTILEVER CONSTRUCTION

A situation is considered here where the permanent pier cannot provide adequate stability during cantilever construction. Several methods may be used, either separately or in combination, to provide the required stability under the loading combinations briefly reviewed in Section 5.2. In the general case Temporary Eccentric Prestress where the construction procedure allows the unbalanced segment in a typical cantilever to be placed always on the same side of the pier, the unbalanced moment varies between 0 and Wd (segment weight W at a distance d from the pier centerline as shown in Figure 5.52). Assume a temporary vertical tendon, anchored in the pier foundation or in a separate dead-man, to be stressed for this unbalanced loading configuration to a load P such that

FIGURE 5.50. St. Jean Bridge at Bordeaux, platform on legs used for river pier caissons. (a) Platform in floating stage. (6) Platform on legs and caisson during excavation.

Match casting was used for making the various segments, and it proved very efficient and very simple. The cofferdam required to build the pier shaft in the dry was made up of temporary additional caisson ring segments stacked upon the permanent caisson and bolted together. This cofferdam was used during construction of the deck to make a moment-resisting pier shaft as a substitute to the flexible permanent pier. The deck was therefore resting only upon the cofferdam and the lower caisson through two temporary caps, offering a stable base for unbalanced loading, Figure 5.51~. After cantilever construction was finished and continuity achieved in the deck, flat jacks were used to transfer the total reaction of the box girder from the temporary caps and cofferdam onto the permanent concrete piers. All the temporary ring segments above low water were further removed. This example shows how the foundations and even part of the substructure can be used to minimize the cost of temporary supports required for cantilever construction.

and the unbalanced moment in the pier now becomes +wd - 2 and the actual bending capacity of the pier is theoretically doubled. The true gain is somewhat lower, because it is not practical to change the tendon load at each stage of segment placing. A proper temporary connection with high-strength rods between pier and deck must always be provided. Unsymmetrical Distribution of Segments with Regard to the Pier If the pier segment is eccentrically

placed with regard to the pier shaft centerline, Figure 5.53, a permanent moment is applied to the pier when an even number of segments is incorporated in the deck. Dimensions may be such that the maximum unbalanced moment due to one segment’s being placed on the proper side of the pier will result in applying only half to the pier. This approach results in significant complications in the layout of the prestress tendons in the deck. Both methods described thus far have one disadvantage, in that the deck cantilever is never in balance over the pier and so it is more complicated to following up the geometry of the deck during construction.

FIGURE 5.51. St. Jean Bridge at Bordeaux, temporary arrangement of piers for deck cantilever construction. (a) Schematic of temporary cofferdam and deck support. (b) View of the pier segment and travelers. 269

270


///I/Py”Hf d FIGURE 5.52. Temporary stability of deck and pier during construction by prestressing tendon. Overhead truss in cast-in-place construction, Siegtal Bridge or Pine Valley Creek Bridge Launching gantry in precast construction, Rio Niteroi Bridge and the B-3 South Viaducts Overhead beam in precast construction, B-3 South Viaducts; a similar scheme is being contemplated for several contemporary projects in the United States. FIGURE 5.53. Unsymmetrical pier segment. Stability of the Concrete Cantilever Provided by the Deck Construction Equipment Figure 5.54 outlines a

Temporary Su@orts (Figure 5.55) If a single temporary support is used on one side of the pier at a distance a, the reactions are as follows:

pM

few typical schemes developed for either cast-inplace or precast construction where the stability during cantilever placing is achieved by the construction equipment itself, such as an overhead truss or launching gantry. Several such examples were previously described in Chapters 2 and 3: Ties for stability

Overhead truss

a

temporary support:

---_ __--

c

+M a

M=W.d

IV Q

----

M=Wd QV

Winch Overhead beam

FIGURE 5.54. Cantilever stability by deck construction equipment.

FIGURE 5.55. Cantilever stability by temporary support(s).

271

Abutments

If two symmetrical temporary supports are used, the system is statically indeterminate and the actual distribution of reactions depends upon the respective flexibilities of the pier and of the supports. The load distribution is as follows: Temporary Support, T,

Pier, P

Temporary Support, T,

PV

(1 - 2P)V

PV

0

M +2a

Effect of vertical load Y

M 2a

Effect of moment M

Total

Pv-g

If it is desired that the temporary supports never be subject to an uplift force, to resist which requires anchors and adequate foundations, the stiffness of the support must be such that a sufficient proportion of the vertical load compensates the effect of the moment. The minimum value of p must be such that: M 2a

pV--20 o

r

M 2a

p2---

Consequently the maximum reaction at support T2 becomes at least equal to M/a, which is precisely the value of the reaction for a single support with

(1 -

PV + $

2P)V

the same loading configuration. The double support system is therefore exactly twice as expensive as the single support system. The only advantage is to allow the construction of the deck to proceed indifferently from either side of the pier or to maintain an equal safety of the system should a mistake be made in the required sequence of operations for the case of a single support. Temporary Stays In a limited number of structures, stability during construction was provided by temporary vertical or inclined stays anchored in special foundation blocks or in the permanent footing of the pier, Figure 5.56. When feasible, this last system is particularly simple, because the temporary stays are usually made of simple prestressing tendons and are far less expensive than rigid temporary supports. Such a system must be used in conjunction with a strong temporary connection between pier and deck to reach an adequate level of safety.

5.8

Abutments

5.8.1

SCOPE

Although the abutments provided at both ends of the bridge are not necessarily of special design when associated with cantilever and segmental construction, it may be of interest to review briefly several types of structures actually used in completed projects. The abutments serve a twofold purpose:

FIGURE 5.56.

Cantilever stability by temporary stays.

They provide the first and last support to the bridge superstructure, allowing a smooth transition of the roadway surface from the deck to the

272


approaches while allowing free expansion with an adequate roadway and sidewalk joint, They make the retaining wall contain the fill of the approach embankment where geometric conditions require it. Design and construction methods of the abutments depend greatly upon the soil conditions and the level of the water table when present. Basically, the two functions outlined above ma! either be integrated into a single structure or filled by two separate structures. On the other hand, the function of a retaining wall may be greatly minimized by allowing the approach fill to take a slope of repose under the structure. By variously combining these characteristics, twelve different sketches were prepared in Figures 5.57 through 5.68 as an outline of typical structures encountered in practice. For convenience, these designs have been grouped into six different categories as described in the following paragraphs.

5.8.2 COMBINED ABUTlMESTIRETAI.VI.1’G

WALL

Type IA (Figure 5.57) A simple retaining wall perpendicular to the bridge centerline and anchored to a conventional spread footing both contains the approach fill and provides the deck end bearing. The back wall receives a transition slab to avoid the roadway profile discontinuity so frequent in earlier bridges between the rigid deck and the flexible pavement over the approach embankment. Two side walls of triangular shape contain the fill inside the abutment. Type ZB (Figure 5.58) The retaining wall is made of a vertical wall and a lower slab properly strengthened by longitudinal buttresses. The entire system is founded on piles.

Type IC (Figure 5.59) Where the poor quality of the soil makes it difficult to resist the horizontal loads due to earth pressure combined with braking and thermal reactions, the previous system may be founded on a system of vertical piles, while the

FIGURE 5.57. Abutment type

IA.

1

If/

1

FIGURE 5.58. Abutment rype

IB.

t-‘--t

FIGURE 5.59. Abutment type IC. 273

274


horizontal loads are resisted by embedded prestressed concrete ties anchored in the back into a continuous dead-man. 5.83

SEPARATE END SUPPORT AND RETAINI,VG WALL

Type ZZ (Figure 5.60) The two functions of deck support and retaining wall are entrusted to two separate structures. Shown in this figure is a front vertical column, resting on spread footings or piles, which provides the deck end bearing. Behind this column and separate thereto, a reinforced earth retaining wall contains the approach fill.

5.8.4

THROUGH

FILL

ABUTMENT

The fill extends under the bridge deck with a stable slope (3 : 2 to 2 : 1) to reduce as much as possible the amount of earth pressure applied to the abutment. Type ZZZA (Figure 5.61) Vertical longitudinal walls connect the lower spread footing to the abutment superstructure. It is important to avoid horizontal cross bracings at intermediate levels embedded in the fill, because settlements may cause significant overloads in such members such as to cause failure.

-

t-4.G -

-

& ~~~FOUR-4’@m

PILES-_

_

(b)

FIGURE 5.60. Abutment type II with reinforced earth. (a) Cross section. (b) Elevation and longitudinal section.

Abutments

275

FIGURE 5.61. Abutment type IIIA. Type ZZZB (Figure 5.62) The same system may be adapted to the case where a high water table and poor soil conditions call for pile foundation built in a cofferdam. 5.8.5 HOLLOW BOX ABUTMENT Trpe WA (Figure 5.63) Another way to avoid high earth-pressure loads on the abutment, where it is not possible or desired to extend the approach fill under the deck, is to build the abutment as a box with a front wall providing the deck end support and the cover slab carrying the roadway between the bridge deck and the approach fill. Such a structure may be founded on spread footing or on piles (as shown in the sketch). Type ZVB (Figure 5.64) The same structure may rest both at the front and at the rear on opendredged caissons excavated under water to the load-bearing soil.

5.8.6 ABUTMENTS DESIGNED FOR UPLIFT

The principle has been described previously in Chapter 4 (design) and for actual structures in Chapters 2 and 3 (cast-in-place or precast cantilever bridges). Type VA (Figure 5.65) A large caisson is opendredged and filled after completion of the excavation to the required foundation level with tremie concrete so as to obtain a sufficient weight to resist the uplift reaction from the deck. Type VB (Figure 5.66) Another variation of the same concept was developed for the Saint Jean Bridge at Bordeaux to combine into a single abutment a front downward bearing and a rear uplifting bearing to fix the last span of the bridge while retaining its free expansion.

276


FIGURE 5.62. Abutment type IIIB.

5.8.7

MINI-ABUTMENT

For decks of small height, when prevailing conditions allow the fill to be placed around the deck, the abutment reduces to a very simple inexpensive structure shown as types VIA and VIB in Figures 5.67 and 5.68.

5.9

Effects of Differential Settlements on Continuous Decks

The question has often been raised as to the adequacy of allowing continuous decks to rest on piers subjected to possible differential settlements. The authors are aware of a few cases where differential


277

6”

d-t

I i'

/

FIGURE 5.63. Abutment type IVA.

settlements were responsible for problems pertaining to the integrity of the superstructure (such as opening of joints between successive segments). Differential settlements, however, are very seldom critical in most soil conditions. In the isolated cases where they may be critical, precautions can be taken to counteract their eventual effects upon the structure. 5.9.1 EFFECTS OF AN ASSUMED PIER SETTLEMENT ON THE STRESSES IN THE SUPERSTRUCTURE

Starting with the simple case shown in Figure 5.69, where a continuous beam of constant depth with a large number of identical spans is subjected to the

settlement of one pier by a given amount, one may easily derive the effect in terms of moments and stresses in the superstructure. Taking the fixed end moment p = 6 EZu/12, the moments over the piers and at midspan are: Over the pier subjected to settlement Over the adjacent piers Midspan moment Quarter-span moment

+0.732p -0.464~ +0.134/L

+0.433p

The stress produced in the superstructure is f =

MC/I, where c is the distance between the centroid

278


FIGURE 5.64. Abutment type IVB.

and upper or lower flange. If the moment is expressed as A4 = Ap, the stress becomes: 6Ecu

f=+r which can be rewritten as follows:

The value of clh varies between 0.4 and 0.6 and that of hll between & aud &. Considering the quarter-span .point close to the pier where settlement occurred, the stress in the superstructure will be, with k = 0.433 and E = 300,000 kips/fP (for long-term loading):

f =

FIGURE 5.65. Abutment type VA.

23,400;

For a settlement u = r$6a the stress is equal to 23 kips/ft2 at the bottom fiber, a very nominal value. For a 100 ft span, the corresponding settlement is u = 0.1 ft = 1.2 inches. The amount of settlement to be considered is only that part taking place after continuity is achieved in the deck and so after most of the load has been applied to the structure.


awble

279

1

FIGURE 5.66. Abutment type VB.

5.9.2 PRACTICAL MEASURES FOR COUNTERACTING DIFFERENTIAL SETTLEMENTS

In most cases, the foreseeable differential settlements may be absorbed by the structure without any corrective measures and no special provisions need be taken in that respect.

For some structures the situation may call for special consideration. Such was the case, for example, with the Houston Ship Channel Bridge, where large long-term settlements could be anticipated at the time of design. In such instances, provisions for eventual realignment of the deck profile must be incorporated into the design.

280


FIGURE 5.67. Abutment type

VIA.

FIGURE 5.68. Abutment tvpe VIB Inertia : I

Modulus : z

FIGURE 5.69. Effect of differential settlement on a continuous beam with equal spans and constant depth.

References 1. J. Mathivat, “Reconstruction du pont de Choisy-leRoi,” Trauaux, Janvier 1966, No. 372. 2. J. Mathivat, “Structures de piles adaptees a la construction par encorbellement,” Problems speciaux d’etude et d’execution des overages, Journees A.F.P.C., Avril 22-23, 1974.

3. Gerwick, Ben C. Jr., “Bell-Pier Construction, Recent Developments and Trends,” Journal of the American Concrete Institute, Proc. V. 62, No. 10, October 1965.

6 Progressive and Span-by-Span Construction of Segmental Bridges J

6.1

INTRODUCTION

6.4.5 Guadiana Viaduct, Portugal 6.4.6 Loisach Bridge, Germany 6.4.7 Rheinbriicke Dusseldorf-Flehe,

6.1.1 Progressive Placement Method 6.2

6.1.2 Span-by-Span Method PROGRESSIVE CAST-IN-PLACE

BRIDGES

6.2.1 Approach Spans to the Bendorf Bridge, Germany 6.2.2 Ounasjoki Bridge, Finland 6.2.3 Vail Pass Bridges, U.S.A. 6.3 PROGRESSIVE PRECAST BRIDGES 6.3.1 Rombas Viaduct, France 6.3.2 Linn Cove Viaduct, U.S.A. 6.4 SPAN-BY-SPAN CAST-IN-PLACE

BRIDGES

6.4.1 Kettiger Hang, Germany 6.4.2 Krahnenberg Bridge, Germany 6.4.3 Pleichach Viaduct, Germany 6.4.4 Elztalbticke, Germany

6.1

Introduction

The concepts of the progressive placement and span-bv-span methods of segmental construction were introduced in Sections 1.9.4 and 1.9.3, respectivelv. .fhis chapter will explore these concepts in greater detail. These two methods have not made the conventional cast-in-place on falsework method obsolete; the conventional method is still applicable and economical where site, environmental, ecological, and economic considerations permit. What these two methods do is to open up a held where prestressed concrete structures were hitherto not practical and where they now can economically compete with structural steel. .The progressive placement and span-by-span methods are similar in that construction of the superstructure starts at one end and proceeds continuously to the other, as opposed to the balanced cantilever method where superstructure is constructed as counterbalancing half-span cantilevers

6.5

Germany

6.4.8 Denny Creek Bridge, U.S.A. SPAN-BY-SPAN PRECAST BRIDGES 6.5.1 Long Key Bridge, U.S.A. 6.5.2 Seven Mile Bridge, U.S.A.

6.6

DESIGN ASPECTS CONSTRUCTION 6.6.1 6.6.2 6.6.3

OF

SEGMENTAL

PROGRESSIVE

General Reactions on Piers During Construction Tensions in Stays and Deflection Control During

Construction 6.6.4 Iayout of Tendons for Progressive Construction REFERENCES

on each side of the various piers. Also, both methods are adaptable to either cast-in-place or precast construction. 6.1.1

PROGRESSIVE

PLACEMENT

METHOD

This method was developed to obviate the construction interruption manifested in the balanced cantilever method, where construction must proceed symmetrically on each side of the various piers. In progressive placement, the construction proceeds from one end of the project in continuous increments to the other end; segments are placed in successive cantilevers from the same side of the various piers. When the superstructure reaches a pier, permanent bearings are placed and the superstructure is continued in the direction of construction. The first implementation of this method, which used cast-in-place segments, was on the Ounasjoki Bridge near the Arctic Circle in Finland. It was 281

282

Progressive and Span-by-Span Construction of Segmental Bridges

later extended to the first use of precast segments in the Rombas Viaduct in eastern France. The essential advantages of this method are as follows: 1.

The operations are continuous and are carried out from that part of the structure alread, constructed. Access for personnel and materials is conveniently accomplished over the surface of the structure already completed (free of the existing terrain). This may be of importance with regard to urban viaducts cantilevering over numerous obstacles.

2.

Reactions to the piers are vertical and not subject to any unsymmetrical bending moments, thus avoiding the need for temporary bracing during construction. 3. The method is adaptable to curved structure geometry. The following are the disadvantages: 1.

It is difficult, if not impossible, to utilize this method in the construction of the first span. Usually the first span must be erected on falsework. In some rare instances it may be possible to cantilever the first span from the abutment. 2. Forces imposed upon the superstructure, depending on the method of construction, are completely different (in sign and order of magnitude) from those present in the structure under service load. Consequently, a temporary external support system is required during construction in order to maintain the stresses within reasonable limits and minimize the cost of unproductive temporary prestressing. Falsework bents may be used (as in the Linn Cove Viaduct), but the more usual solution is that of a mobile temporary mast and cable-stay system (Figure 1.57). For the progressive placement method the mast and cable-stay system is relocated progressively over the piers as construction advances. 3.

In this system the piers are subjected to a reaction from the self-weight of the superstructure approximately twice that in the final static arrangement of the structure. However, this is generally not critical to the design of the piers and foundations, as the effect of the dead load is rarely larger than half the total load including horizontal forces.

When cast-in-place segments are used in conjunction with the progressive placement method, the rate of construction is less than that t’or the balanced cantilever method, in that there is onlv one location of construction activitv. That is, onlv one segment can be cast (at the end of the completed portion of the structure) rather than two (one at each end of the balanced cantilevers). ~fhis slowness may be minimized by the use of longer segments, but this solution is limited bv the low resistance of the young concrete. On the other hand, the use of epoxy-joined precast segments ma! permit an average rapidity of construction comparable to that of balanced cantilever with a launching girder.

As indicated in Chapter 1, the span-by-span method was developed to meet the need for constructing long viaducts with relatively short spans such as to incorporate the advantages of balanced cantilever construction. From a competitive point of view, the capital investment in the equipment for this type of construction is considerable. It has been suggested’ that one-third of the cost of the equipment be depreciated for a given site and that at least four uses would be required to achieve full depreciation, including interest on the capital investment. However, costly modifications that may be required because of changes in bridge widths or span limitations are not considered in the above write-off policy. It would, therefore, be advisable for a contractor investing in this tvpe of equipment to consider some type of modular planning so that modification for future projects might be kept to a minimum. It might be possible to have a basic piece of equipment with interchangeable elements. There is, of course, the potential of leasing this equipment to others as a means of retiring the capital investment. Wittfoht1s2 has categorized stepping segmental construction intb four subgroups: 1. With-on-the ground nontraveling formwork. 2. With traveling formwork or on-the-ground stepping formwork. 3. 4.

With off-the-ground stepping formwork. In opposite directions starting from a pier.

The first category is generally used where there are a large number of approximately equal spans

283

Progressive Cast-in-Place Bridges of a low height above existing terrain. It is generally limited to structure lengths of approximately 1000 ft (300 m) and to nonuniform span lengths that prohibit a forming system of uniform size. Normally in span-by-span construction the superstructure is of constant cross section (at least insofar as external dimensions are concerned), and the work proceeds from one abutment to the other. If a large center span exists, it will be formed first, possibly to an inflection point in the adjacent spans. The formwork is allocated such that it is used to cast the spans in the approaches proceeding from the center, in both directions, toward the abutments. Forms and scaffolding are disassembled and reerected in an alternating sequence and in elements that can be conveniently handled by a crane. In the second category of span-by-span construction, for economical justification of equipment, the total length of structure must be at least 1000 ft (300 m), the overall cross section constant, the structure of low height, and the terrain along the longitudinal axis approximately level. Maximum span for this category is approximately 165 ft (50 m), and a large number of equal spans are required to achieve repetitiveness and thus econon1v.3 The falsework and forms are generally a span length (either the dimension from pier to pier or from inflection point to inflection point), Figure 6.1 .3 The formwork is fixed to the scaffolding and travels with it. The bottom of- the formwork is designed with a hinge or continuous trap-door device such that the scaffolding and forms can travel past and clear the piers. The scaffolding is moved forward on rails. If a foundation for the scaffolding, forms, and weight of superstructure is found to be too costlv or unsafe, a scheme may be used where the rails ‘carry only the load of the scaffolding and formwork. Once in position, the scaffolding is supported at the piers, or at the forward pier, and the completed structure at the rear by auxiliary brackets; thus construction loads are transmitted to the pier foundations. Where conditions exist as in the previous category, but the structure is high with reference to the terrain or crosses over difficult terrain or water, the third category may be used, whereby during the stepping and casting operations the equipment is supported by the piers or by a pier and the previously completed portion of the structure. Where consecutive spans in the range of 160 to 500 ft (50 to 150 m) are contemplated and the factors mentioned above prevail, the type of con-

i-2

rf

Scaffolding at concreting position Construction

direction

‘Under-carriage Advancement of Scaffolding

ffold m re _.,.._ Hinged bottom plate Section l-l

Section

2-2

FIGURE 6.1. Schematic of procedure for movable scaffolding, from reference 3 (courtesy of Zement und Beton). struction indicated by the fourth category may be considered. This system uses a gantry rig that has a length one and one-half times that of the span. In this method segments are cast in each direction from a pier, as in the balanced cantilever method, except that the form traveler and segment being cast are supported by the gantry. This method is actually a balanced cantilever method and not a span-by-span method of construction as defined here. The advantages of the span-by-span method of construction, besides those associated with segmental construction in general, pertain to the prestressing steel requirements. Since the segments are supported by the form travelers, there are no cantilever stresses during construction, and prestress requirements are akin to those of conventional construction on falsework or those for the final condition of the structure.

6.2 Progressive Cast-in-Place Bridges 6.2.X

APPROACH SPANS TO THE BENDORF BRIDGE, GERMANY

As discussed in Section 2.2, the Bendorf Bridge was constructed in two parts. The western portion


284 Main river

Flood

Construction in free cantilever

216.50 m

Construction on I-.___--.-__i

falsework

288.50

m

‘Phase 5 by progressive placing, segment length 4.00 m.

FIGURE 6.2. Bendorf Bridge, Part Two (East), construction procedure, from reference 1 (courtesy of Beton- and Stahlbetonbau). Phase 5 by progressive placing, segment length 4.00 m. (part one), Figure 2.9, consists of a symmetrical seven-span continuous girder constructed by the cast-in-place balanced cantilever method. The eastern portion (part two), Figure 2.10, consists of a nine-span continuous approach structure having an overall length of 1657 ft (505 m) with spans ranging from 134.5 ft (41 m) to 308 ft (94 m). In the construction of the approach spans, Figure 6.2, the five spans from the east abutment were built in a routine manner with the assistance of falsework bents. The four spans over water were constructed by the progressive placement method, using cast-in-place segments and a temporary cable-stay arrangement to reduce the cantilever stresses. The temporary stay system consisted of a structural steel pylon approximately 65 ft (20 m) high and stays composed of Dywidag bars. 6.2.2 OUNASJOKI BRIDGE, FINLAND

This structure is near the city of Rovaniemi, Finland, and crosses the Ounas River just above its junction with the River Kemi near the Arctic Circle. The structural arrangement consists of two 230 ft (70 m) interior spans and end spans of 164 ft (50 m), prestressed longitudinally and transversely. The first end span and 75 ft (22.75 m) of the second span were cast-in-place in a conventional

manner on falsework inside a temporary windshielded protective cover, Figure 6.3. Outside temperature during this operation ranged from - 2 0 t o -30°C. Subsequent progressive cantilever construction was performed-with the aid of a temporary pylon and stays, Figure 6.4. The same stages were repeated in the remaining spans. The superstructure was cast-in-place with the assistance of one form traveler, Figure 6.5. During these stages of construction, for protection against low temperatures, form traveler and form were fully enclosed, Figure 6.5. This enclosure was insulated with 4 in. (100 mm) of fiberglass. Hardening of the concrete took an average of 76 hours. Temperature of the concrete was maintained between 35 and 45°C at mixing and between 20 and 25°C during casting. Curing inside the form traveler enclosure was assisted by warm-air blowers. Concrete strength was 5000 psi (34.5 MPa). Segment length was 11.5 ft (3.5 m), and it was possible to reach a casting rate of two segments a week. Construction started in 1966 and was completed in 1967. Table 6.1 lists the temperatures recorded during seven months of the construction period. The progressive placement method proved effective and work progressed throughout the year even during arctic conditions.

FIGURE 6.3. Ounasjoki Bridge, temporary protective structure (courtesy of Dyckerhoff & Widmann).

285

Progressive Cast-in-Place Bridges TABLE 6.1. Ounasjoki Bridge, Temperature Variations Month Temperature

March

April

May

June

July

August

September

Average “C Maximum “ C Minimum “C:

-2.5 +5.8 -26.4

-0.4 +9.9 - 16.8

+5.6 +24.6 - 12.2

+11.7 +24.9 +0.1

+ 14.3 +25.7 +3.0

+ 14.8 +28.5 +5.8

+8.7 + 19.3 -4.7

62.3 VAIL PASS BRIDGES, U.S.A.

FIGURE 6.4. Oulla+ki Bridge, winterproof ing form (courtesy of Dyckerhof’f & Widmann).

FIGURE 6.5.

travel-

The Vail Pass structures are part of Interstate I-70 near Vail, Colorado, in an environmentally sensitive area. Of the 21 bridge structures in this project, seventeen were designed and bid on the basis of alternate designs (Chapter 12). In the segmental alternative the contractor was allowed the option to construct as cast-in-place segmental. A group of four bridges approximately 7 miles (11.3 km) southeast of Vail were successfully bid as cast-inplace segmental and used the concept of progressive placement. Two of these structures are contained in a fourspan dual structure over Black Gore Creek, Figure 6.6. The other two structures are a three-span

Ounasjoki Bridge, progressive placing scheme.

286


Existing grcamdlim

TYPICAL ELEVATION

MID-SPAN

NEAR Q PIER TYPICAL

SECTION

FIGURE 6.6. Vail P ass B I -‘cl I ges, Black Gore Creek Bridge, typical elevation and section

eastbound bridge and a four-span westbound bridge, both crossing Miller Creek, Figure 6.7. Because the structures are relatively short and the spans small, they were constructed by the progressive placement method with temporary falsework bents. ‘The work and time required to transport and reassemble the form travelers (as in the balanced cantilever method) was therebv minimized. Construction started from both abu;ments and proceeded progressively toward the center of each bridge.” For each of the two structures in the Miller Creek Bridge, form travelers were assembled atop 30 ft (9.1 m) long segments at the abutments. As segment casting began, the side spans were supported at every second segment by a temporary bent. After reaching the first pier, segment con-

struction proceeded in normal fashion to midspan of the eastbound structure. In the westbound structure, when midspan of both interior spans was reached, temporarv bents were again used to conlplete the remaining half-spans to the center pier. After reaching the center of the bridge, one form traveler of each bridge was dismantled, and the remaining form traveler was used to cast the closure pour. In this manner the form travelers for each bridge were assembled and dismantled only once, as opposed to the method of assembling t\vo forms at each pier and dismantling upon completion of two half-span cantilevers about each pier. For the Black Gore creek structures, to save critical construction time, both end spans of one structure and one end span of the other structure were built on falsework, while the form travelers

Progressive Cast-in-Place

Bridges

455’- 3 ”

E. 8 ELEVATION

518’-3”

Bridge abut. 2 E Dridge

abut 1

W. 8. ELEVATION

42'-0"

e Elri ge

,_

:

r2"

1 Asphalt

lo’-o-

9’-

33-e.

I'-8f'

TYPICAL SECTION

FIGURE

6.7.

Vail Pass Bridges, Miller Creek Bridge, typical elevation and section.

were occupied at the Miller Creek Bridges. Upon completion of their work at Miller Creek, the form travelers were transported over the completed end spans of the Black Gore Creek Bridges and con-

struction continued in the progressive placement manner, Figure 6.8. Because of the limited construction time a three-day cycle was required for segment casting.

288


Construction specifications required a concrete strength of 3500 psi (24 MPa) at the time of posttensioning and 5500 psi (38 MPa) at 28 days. Since the time required for f-orming and placing of rebar and tendons is somewhat fixed, the only operation that could be adjusted was the concrete curing time. This was accomplished by using a special water-reducing agent that allowed the development of 3500 psi (24 MPa) concrete in 18 hours. Because of lack of experience with the specific water reducer, honeycombing was experienced in the early stages of construction. Eventually a 24 da) cvcle was achieved. FIGURE 6.8. Vail Pass Bridges, Black Gore Creek Bridge, under construction (courtesy of Dr. Man-Chung Tang, DRC Consultants, Inc.).

ROM BAS - P L A N VlEL/(a) FIGURE 6.9. Rombas Viaduct, plan and sections. (a) Plan. (6) Typical bridge sections. (c) Typical segment section.

289

Progressive Precast Bridges coupe c

Coupe A

Coupe D

Coupe El

/ Variable I

Var 680 760 I-

-l

F i g u r e 6 . 9 . (C,‘o~rtitr~rd)

6.3 Progressive Precast Bridges

The Rombas Viaduct is a constant-depth superstructure, supported on neoprene bearings, with nine continuous spans ranging from 75 ft (23 m) to 14X ft (45 m). This structure is curved in plan with a minimum radius of 900 ft (275 m) and of a variable width, owing to the presence of an exit ramp, Figure 6.9. Total length is 1073 ft (327 m),

and the viaduct has two parallel single-cell boxes. In cross section each single-cell box is 8.2 ft (2.5 m) deep and has a width of 36 ft (11 .O m). A construction view of the end of a segment is presented in Figure 6.10. Construction of this structure employed the progressive placing of the precast segments. Temporary stability was provided by a cable-stay system, Figures 1.57 and 6.11, which advanced from pier to pier as the construction progressed. Segments were progressively placed, starting from one


290

FIGURE

6.10.

Rombas Viaduct, end view of segment.

FIGURE

6.12.

Rombas Viaduct, view ot swivel crane.

FIGURE 6.13. Linn Cove Viaduct, photomontage.

FIGURE 6.11. Rombas Viaduct, view of cable stays

and mast.

abutment, by means of a swiveling hoist, Figure 6.12, advancing along the deck. 6.32

LJNN COVE VIADUCT, U.S.A.

A progressive placement scheme is being used for the Linn Cove Viaduct on the Blue Ridge Parkway in North Carolina, Figures 6.13 and 6.14. It is a

FIGURE

6.14.

Linn Cove l’iaduct,

artist’s

rendering.

291

Progressive Precast Bridges

1243 ft (378.84 m) eight-span continuous structure with spans of 98.5, 163, 4 at 180, 163, and 98.5 ft (30.02, 49.68, four at 54.86, 49.68, and 30.02 m) and sharp-radius curves, Figure 6.15. In cross section it is a single-cell box girder with the dimensions indicated in Figure 6.16. Because of the environmental sensitivity of the area, access to some of the piers is not available. Therefore, the piers will be constructed from the tip of a cantilever span, with men and equipment

being lowered down to construct the foundation and piers. The piers are precast segments stacked vertically and post-tensioned to the foundation, Figure 6.17. The extreme curvature of the alignment makes the use of temporary cable stays impractical. Temporary bents at midspan will be used to reduce cantilever and torsional stresses during construction to acceptable levels. The temporary bents are erected in the same manner as the permanent

+ Pier 3

i i

FIGURE 6.15. Linn Cove Viaduct, plan.

$ Pier 4

H A L F S E C T I O~~~~ N AT POST- TENSIONING BLOCK _---

FIGURE 6.16.

TYPICAL HALF SECTION

THRU

I,inn Cove Viaduct, typical segment cross section.

SEGMENT

Span-by-Span Cast-in-Place Bridges

293

4.4 Span-by-Span Cast-in-Place Bridges 6.4.1 KETTIGER H,4,VG,

Construction bar tendons through segments not shown

The first application of’ the ot‘t-ground tnethodology (category 3), Section 6.1.2. was in 1955 on the Kettiger Hang structure neat- Andernach (Federal Highway 9), Figure 6.19.3 This system consists of f’our scaffblding trusses of’ slightly more than a span length and two cantilever girders of’ about a two-span length. The scat‘folding trusses support the entire concrete weight during casting. The cantilever girders serve to transfer or advance the scaf’folding trusses to the next span to be cast. The concrete fortn or mold rides with the scat‘folding trusses and is thus repeatedly reused. h.-l.2

FIGURE 6.17. Lint1 Cole Viatiuct, segtnental

pier.

piers, using a stiff-leg derrick at the end of’ the completed cantilevered portions of’ the structure, Figure 6.18. When the temporary bents are no longer required, they are dismantled and removed bv equipment located on the completed portion of the bridge deck.

GER.LC4.Yk

KRz4H.\‘E.\‘BERG

BRIDGE, GERA11.4.Y).

A variation of the of‘f-the-ground system was used on the Krahnenbergbrticke near Andernach constructed from 1961 to 1964, Figure 6.20.‘*3 This structure has a length of’ 3609 f’t (1100 m), a constant depth of 6.56 fi (2.0 m), a width of’ 60.i f’t (18.5), and spans of’ 105 f’t (32 m). The site is on a slide-susceptible hillside, requiring difficult foundations, and its curved alignment follo\vs the topography, all of which economically favored the span-by-span technique.

STIFF LEG DERRICK LACING PRECAST SEGMENTS

FIGURE 6.18.

Linn Cove Viaduct, erection scheme for progressive placement.

39,26-39, M Section l-l

Scaffolding truss at concreting position Cantilever

~39,20~3920~3Q,20 39,20-t----3Q,M-39,20

3%

%i%f

Advancement of the Scaffolding truss including forms Forward Rear

slope

Section

2-2

0 n;

I 39.20-39.20

39,20--c-------39>20

39.20

,

39.20

Advancement of the cantilever beams

FIGURE 6.19. Kettiger Hang, schematic of’ the construction procedure, from reterence 3 (courtesy of’ Zement und Beton).

Exterior scaffold girder I’I

‘\

Interior scaffold girder

m

t t

(b)

294

Span-by-Span Cast-in-Place Bridges In this project four fbrmwork supporting girders were used. Two interior girders were rigidly connected together by transverse horizontal bracing. The formwork was arranged so that the forms hinged at the bottom and folded down to allow passage, during advancement, past the piers, Figure 6.200. ‘The four girders were supported on the hexagonal piers by transverse support beams attached to the pier. In this manner the four longitudinal formwork support girders were supported on two piers, while an additional set of transverse support beams were attached to the forward pier. Figure 6.206. Latticework cantilever extensions at both ends of the longitudinal formwork support girders extended their length to twice the span length, so that a stable support was provided by the transverse support girders during advancement. The outside girders had joints or links at the connection with the cantilever latticework so that the curvature of the structure could be accommodated during their advancement. The elevation of the outside girders was adjusted by hydraulic jacks to accommodate superelevation. During the advancement operation the outside girders were advanced first and then the center two girders, Figure 6.20~. When the forward end of the interior girders reached the transverse supporting beams, the rear transverse beams of the previously cast span were no longer required. They were dismantled from the pier. These transverse beams were erected on the next forward pier by a crane, Figure 6.20b. The exterior formwork of’ the two-cell box girder was attached to the longitudinal support girders and only required adjustment for curvature. The interior forms of- the cells were dismantled and reassembled on the next span after reinforcement was placed in the bottom flange and webs. FIGURE 6.20. (Opposite). Krahnenberg Bridge, schematic of construction, from reference 1 (courtesy of the American Concrete Institute). (a) Cross section. (6) Formwork equipment in working position. (c) I: Working position: reinforcing, and concreting on formwork equipment; installing the supporting construction on the next following pier by means of derrick and straight-line trolley. II: After concreting and prestressing: lowering of equipment; opening of formwork flaps; shifting forward of outer girders; dismantling of the first rear supporting girder by straight-line trolley; intermediate storage at center pier. III: Partial pony-roughing of center girder; dismantling and placing in intermediate for storage of second rear girder. IV: Final shifting forward of center girders; jacking up of equipment; closing of formwork flaps; new working position.

295

Average casting rate was 706 ft3 per hr (20 m3). Fourteen days was required for construction of a span. 6.43 PLEICHACH VMDUCT, GERMANY In 1963 construction started on the 1148 ft (350 m) long Pleichach Viaduct1a3 carrying a federal highway between Wurzburg and Fulda; it was the first use of the span-by-span technique for a dual structure, Figure 6.21. Span length is 119 ft (36.25 Rear crane truck Forward crane truck

rk

! ‘“__ -.-_-----~~-c-~~--~R Fiv 1’; ” ,I ,A +; I //

__---

/

I I

Scaffolding girder at concreting position

36.25+-x,25-

--i -36.25

Advancement of the scaffolding girder including forms

Construction joint

Advancement of the scaffolding and cantilever girders

R-Scaffolding girder and forms W-Scaffolding and cantilever girder

I, i I

I

I !

i

Cross section

FIGURE 6.21. Pleichach Viaduct, schematic of the construction procedure, from reference 3 (courtesy of Zement und Beton).

296


m), with each two-cell box girder having a width of 47.2 ft (14.4 m) and a depth of 7.2 ft (2.2 m). The was equipment superstructure construction erected behind an abutment in a position to construct one superstructure. Upon reaching the opposite abutment, the equipment was shifted laterally for the return trip to construct the other superstructure. Because of the narrowness, only one longitudinal support girder was required, as opposed to the two girders required for the Krahnenberg Bridge. This girder is slightly longer than twice the span length. The two outside girders are approximately one span length. The outside girders were advanced simultaneously by a carrier traveling at the front of the central girder and at the rear by carriers running on the deck of the previously completed section. During concreting, the two outside girders are supported on brackets at the forward pier and suspended from the completed portion of the superstructure. The center girder, relieved of the load of the two outside girders, is then advanced one span and again connected to the outside girders by the hinged bottom formwork, thus functioning as an auxiliary support girder. This sequence of operations is commonly referred to as the “slide-rule principle.” The piers have a width of 16.4 ft (5 m) and have an opening at the top to allow passage of the central support girder, Figure 6.21. The width of the pier is determined by the need for sufficient bearing area for the bearings and clearance for the central support girder. Whether the central opening at the top of the pier should be concreted in is one of aesthetics.

64.4 ELZTALBRUCKE,GER~~A~ Figure 6.22, was constructed The Elztalbrticke,5,6 in 1965 at Eifel, West Germany, approximately 18.6 miles (30 km) west of Koblenz. It crosses the deep valley of the Elz River with a total structure length of 1244 ft (379.3 m), Figure 6.23. The superstructure has a width of 98.4 ft (30 m) and is supported on a single row of octagonal piers up to 328 ft (100 m) in height, Figure 6.24. Owing to the height of the valley, conventional construction on falsework would have been economically prohibitive. Therefore, a span-by-span system of selfsupporting traveling scaffolding was used, Figure 1.53. The Autobahn between Montabauer and Trier, which had been in planning before World War II,

FIGURE 6.22. El,t,dtmde,

\ ie\\ of c0111plrteti structure (courtesy of Dipl. Ing. Manfred Bockel).

had to cross two large natural obstacles, the Rhine River north of Koblenz (see the Bendorf Bridge, Section 2.2) and the Elz Valley. In 1962 tenders were called for on the Elz Valley structure. Bidders were provided with the grade requirements, dimensions for a single or a dual structure, the location of the abutments, and the foundation conditions. A consortium of Dyckerhoff SC Widmann AG, W a y s s SC Freytag K G , and Siemens-Bauunion GmbH investigated four possible prestressed concrete construction possibilities? 1. A three-span variable depth structure similar to the Bendorf Bridge 2. A six-span constant-depth structure 3. A frame bridge 4. A nine-span “mushroom” construction with a center row of piers These four schemes were proposed, as were a large number of different ones in both steel and concrete by other firms. The successful low bid was for scheme 4 above. The nine-span “mushroom” construction was approximately 4% less costly than an orthotropic-deck, three-span continuous steel girder and 7% less costly than a prestressed concrete girder bridge of six spans.6 The Elztalbrticke, extending the methodology used earlier for primarily low-level urban viaducts, was the first application of the “mushroom” cross section for a high-level structure crossing a deep valley. Previously, this type of construction, because of its short, stiff piers, required a number of expansion joints in the deck to accommodate thermal forces, elastic shortening, creep, and shrinkage. In this structure, owing to the flexibility of the tall piers, only one expansion joint was used,

Koblenz east abutment 3m5-4 A

31aYlo

c 319,941

D

xu,514

E

F

321,128

6

321,779

322 4m

H

D&W rock anchors ; ; z

Longitudinal cross section (a)

Total length = 379.30 m

~~~

~

Plan (b)

FIGURE 6.23. Elztalbticke, longitudinal cross section and plan, from reference 5 (courtesy of Der Bauingenieur). (a) Longitudinal cross section. (h) Plan.

I

Trier west

abutment

298


.

28

435

cross section at pier E, FIGURE 6.24. Elztalbticke, from reference 5 (courtesy of Der Bauingenieur).

in the center span. This joint is located 38 ft (11.6 m) from pier E. The superstructure is monolithically connected at all piers and the abutments. At the center of each span is a 43 ft (13.1 m) long, massive flat plate, which in cross section has a thickness varying from the centerline (crown ot roadway) of 2% in. (650 mm) to 17% in. (450 mm) at the outside edges. The “mushroom” portion ot the span varies in thickness, transversely and longitudinally, to 8 ft (2.45 m) at the pier. The superstructure is prestressed longitudinally and transversely. The octagonal piers have, in cross section, external dimensions 01 15.75 by 19 ft (4.8 by 5.8 m) with a wall thickness of 11% to 1% in. (300 to 350 mm). Any given pier has a constant cross section for its entire height. The percentage of vertical reinforcement, with a concrete cover on the outer and interior faces of 1.5 in. (40 mm), varies from 0.8 to 1.2% of the gross concrete area. Piers were constructed by slip-forming. The eight pier shafts were constructed in seven months. The tallest pier, 3 11.6 ft (95 m) in height, was slip-formed and cast at a rate of about 26 ft (8 m) per day and thus required 12 days to construct. The top 4 ft (1.2 m) portion of the pier was cast with the superstructure by the traveling scaffolding. On the top of the slipformed pier four 7.2 ft (2.2 m) high pedestals were cast to provide the support for the cantilever girder from the traveling scaffolding, Figure 6.25.’ The traveling scaffolding was assembled at abutment A after completion of the abutment and the half-mushroom projecting therefrom. This form traveler, Figure 6.26, accommodates a fullwidth span-length segment of 123 ft (37.5 m). After the first span, two weeks were required to complete a superstructure span. The first opera-

FIGURE 6.25. Elztalbticke, construction view (courtesy of Dipl. Ing. Manfred Bockel).

Side longitudinal girder &Q*Llo

i-9000--+/ I

L+P-.--

dp

Center support bearing i

Center girder

Catwalk

Upper

longitudinal

catwalk

Travel direction 37500

Longitudinal cross section

Center girder

longitudinal

Side longitudinal girder

Concreting

, + Hydraulic

sequence

II)

a Formwork at concreting position

I Scaffolding after advancement

concreting

traveling position

position Section A-B

jack

III)

Forms in stripped position

Section CD Cd)

Cc)

FIGURE 6.26. Elztalbticke, form traveler, from reference 5 (courtesy of Der Bauingenieur). (a) Longitudinal cross section. (b) Plan. (c) Section A-B. (d) Section C-D. 299

300


tion was to cast a 42.65 ft (13 m) wide center portion of the bridge. After hardening and initial stressing, the two outside edges, each 27 ft (8.25 m) wide, were cast. Subsequently the form traveler was advanced to cast the next span.5 As mentioned previously, an expansion joint is located in the center span. During construction this joint was “locked” until construction reached pier G; then the joint was released.5 During concreting the forms are suspended by steel bars, and during advancement the forms are carried by the bottom arm of the transverse cantilevered steel members. The form traveler, Figure 6.26, essentially consists of two approximately 141 ft (43 m) long longitudinal girders and eight transverse frames in a “C” configuration which surrounds the deck construction. The transverse frames may be provided with a covering to protect the workmen and the construction from the weather. At the forward end an approximately 72 ft (22 m) long cantilever beam, located on the centerline, is projected to the next pier for support.

This structure is located o n national route 260 crossing the Guadiana River between Beja a n d Serpa, Portugal. The viaduct has a total length of’ 11 15 ft (340 m) and consists of 197 ft (60 m) spans except for the river spans, which are 164 ft (50 nl). Transversely, the superstructure is 53.8 ft (16.4 m) in width composed of two single-cell box girders. Each box girder is 19.35 ft (5.9 111) wide, with the depth varying from 6.5 ft (2.0 nl) at midspan to 9.8 ft (3.0 m) at the piers. After construction of the box girders, a longitudinal centerline closure is poured and cantilevered sidewalks are constructed. The superstructure is constructed by the spanby-span method, from inflection point to inflection point, by an overhead self-launching f’orm carrier, Figure 6.27. The form carrier consists of 279 ft (X5 m) long trusses of a depth varying f‘ronl 9.8 ft (3.0 m) lo 16.4 ft (5.0 m). Forms fol- concreting the superstructure are supported bv two series of suspenders. One set pierces the concrete flanges and

Forward support ,

I

End traveler support

Elevation (a)

Typical cross section

Section at forward support-forms open

(b)

FIGURE 6.27. Guadiana Viaduct, elevation and sections of form carrier. (a) Elevation. (b) Section at forward support-forms open. (c) Typical cross section.

Span-by-Span Cast-in-Place Bridges is located inside the box cell. The other set is arranged outside the box and supports the forms when stripped and traveling past the piers in an open position, Figure 6.27. During concreting of the superstructure the f’orm carrier is supported on the forward pier b\ a n a r r a n g e m e n t o f a telescoping tubular cross frame, at the rear: it is supported on the superstructure at a location 26 ft (8.0 m) forward of’ the rear pier. When the form carrier is being launched forward, it moves over a support at the tip of rhe completed superstructure cantilever (near the inflection point), and its rear support rides on the surface of the completed superstructure. ‘The form carrier (including all equipment) weighs 209 tons (190 mt). 6.4.6 LOISACH

BRIDGE, C;ER,\lA,\‘k

‘l-he federal autobahn between Munich and Lindau has an alignment that transverses the Murnauer swamp area near Ohlstadt and thus crosses the Loisach River and the old federal highway B-2 (Olympiastrasse), Figure 6.28. Because of flooding and poor soil conditions an embankment was not possible, and a decision was made requiring a dual viaduct bridge structure with a total length of 43 14 f’t (1315 Ill).’ .I‘he 232.8 ft (70.96 m) main span crossing the Loisach River is a variable-depth single-cell box girder constructed by the free cantilever method. Depth of’the box girder varies from 9.84 ft (3.0 m) to 5.58 ft (1.7 m), Figure 6.29. The approach spans are of a T-beam cross section, Figure 6.29, constructed by the span-by-span method with the form carriers running below the superstructure. Figure 6.30 is a longitudinal section of the bridge within the area of the approach spans, showing the form carrier running below the level of the top slab. Figure 6.31 shows the form traveler in action.

Box girder

301 T-beam

FIGURE 6.29. Loisachbriicke, cross sections, from reference 8 (courtesy of Dyckerhoff & Widmann). The dual structure has a total width of 100 ft (30.5 m), Figure 6.29, and each half is supported on two circular piers, excepting the Loisach span which is supported on wall piers. In the total length, the dual structures are subdivided into three sections by two transverse joints, Figure 6.28. In plan the structure has a radius of 4265 ft (1300 m) at the Munich end, and the curvature reverses at the Loisach with a radius of 6562 ft (2000 m).” The completed structure is shown in Figure 6.32. The circular piers are 4 ft (1.2 m) in diameter and are supported on 20 in. (500 mm) driven piles with an allowable load capacity of 176 tons (160 mt). Pile depths vary from 42 to 72 ft (13 to 22 m). A total of 1182 piles were driven for a total length of piling of 63,650 ft (19,400 m), with an average length of pile of 53.8 ft (16.4 m). Load capacity of the piles was determined from eleven load tests taken to 265 tons (240 mt). Because of the poor soil conditions and ground-water pressure, the substructure was con-

FIGURE 6.28. Loisachbriicke, layout and underside view of bridge, from reference 8 (courtesy of Dyckerhoff & Widmann).

302


FIGURE 6.30. Loisachbriicke, longitudinal and cross section showing form traveler (courtesy of Dipl. Ing. Manfred Bockel).

FIGURE 6.31. Loisachbriicke, view of form traveler

in action (courtesy of Dipl. Ing. Manfred Bockel).

quired at midspan. The radius and superelevation in a support length were held constant. Superelevation varies from +5.5 to -4%. For a normal span 8830 ft3 (250 m3) of concrete were placed in nine hours.s Because of the tight time schedule, work was continued through the winter months in defiance of the extreme harsh weather conditions in the Loisach Valley. A weather enclosure was mounted on the form traveler and heated by warm-air blowers. In this enclosure the reinforcement and preheated concrete was placed. In addition, the fresh concrete was protected by heat mats. In this manner the work could proceed up to an outside temperature of 5°F (- 15°C). Construction cycle per span was gradually reduced, after familiarization, from an original 14 days to seven days. Following completion of the western roadway up to the Loisach the form traveler was transferred to the eastern roadway for the return trip to the Munich abutment. All 38 spans on the Munich side were completed by the end of February 1972, saving nine weeks in the construction schedule. On the Garmisch side of the Loisach the movable scaffold system consisted of four principal girders 292 ft (89 m) in length and 9.8 ft (3.0 m) deep, Figure 6.33. Superelevation varies from +4 t o -5.5%.

FIGURE 6.32. Loisachbticke,

view of’ completed structure (courtesy of Dipl. Ing. Manfred Bockel). strutted in pits enclosed by sheet piling. The round piers vary in height from 9.8 to 23 ft (3 to 7 m). Because of the delay in pile driving, resulting from the soil conditions, the foundation completion was delayed from October 1970 to April 1971. The 73 T-beam spans were constructed with two span-by-span form travelers whose operations were synchronized. On the Munich side of the Loisach four 223 ft (68 m) long and 4.26 ft (1.30 m) high principal form support girders are supported in the 100 ft (3 1 m) spans on cross beams at each pier, which in turn are supported off the pile caps. For the longer spans an auxiliary support was re-

Because of the delay in the pile driving, the first span was started in December 1970 with a 12-week delay. The last approach span on the left of the Garmisch side was completed in August of 1971. The traveler was then transferred to the other roadway for the return trip and all 35 bridge spans were completed by March 1972. By a gradual reduction of thk work cycle from 14 days to seven days, nine weeks were saved in the construction schedule. Not only was the loss of time resulting from the foundation work made up, but a time advantage was attained. The four box girder spans (two in each dual structure) on either side of the principal span over the Loisach were cast on stationary falsework. Auxiliary cross beams to support the falsework girder were supported on driven piles. The two main


303

ers were transferred to the opposite pier for the remaining seven segments.” After a construction time of approximately 30 months the bridge was completed in 1972, shortly before beginning of the Olympic Games.

6.4.7 RHEINBRikKE DUSSELDORF-FLEHE, GERMANY

FIGURE 6.33. Loisachbrucke, cross section of movable scaffold system, from reference 8 (courtesy of Dyckerhoff & Widmann). spans of 232.8 ft (70.96 m) were constructed by the free cantilever method. Thirteen segments of 16.4 ft (5 m) were required; six segments were cast from one pier and then the cantilever form travel-

FIGURE 6.34. Rheinbticke erhoff & Widmann).

This is an asymmetric cable-stayed bridge with an inverted concrete Y-pylon, Figures 6.34 through 6.37. The overall length from abutment to abutment is 3764 ft (1147.25 m). The Rhine River span is 1205 ft (367.25 m) long and is a rectangular three-cell steel box girder with outriggers to support a 135 ft (41 m) wide orthotropic deck, Figures 6.36 and 6.37. At the pylon there is a transition from the steel box girder to prestressed concrete box girders, which are used for the thirteen 197 ft (60 m) spans in the approach viaduct. The structure is continuous throughout its entire length, having expansion joints only at the abutments. The approach viaduct has from pier 9 up to pier 13, Figure 6.37, a five-cell box girder cross section with a width of 96.8 ft (29.5 m) and a depth of 12.5 ft (3.8 m). This heavy cross section, Figure 6.36, resists the anchorage forces from the cable stays. For the balance of the viaduct length from abutment to pier 9 the cross section consists of two single-cell boxes, a continuation of the exterior cells of the five-cell box girder cross section. However, the interior webs of each box are of less

Dusseldorf-Flehe, artist’s rendering (courtesy of Dyck-

304


: FIGURE 6.35. Rheinbticke Dusseldorf-Flehe, view from construction end of approach viaduct looking toward the pylon under construction. thickness than that of the five-cell cross section. The width of each box then becomes a constant 23 ft (7.0 m) outside-to-outside of webs. A diaphragm occurs at each pier. The approach spans were constructed segmentally by the span-by-span method with construction joints at approximately the one-fifth point of the span. As described in Section 6.1.2, the method used here employed movable falsework, Figures 1.54 and 6.38, supported from the ground. The 197 ft (60 m) spans were poured in place in one unit from construction joint to construction joint. This required continuous placement of as much as 3200 cubic yards (2500 m3) of concrete. After each section was cast in place and reached sufficient strength, the prestress tendons were stressed and the falsework was moved forward to repeat the cycle. 6.4.8 DENNY CREEK BRIDGE, U.S.A.

The Denny Creek Bridge is the first implementation of the span-by-span method of construction in

the United States. It is located a few miles west of Snoqualmie Pass in the state of Washington and will carry the I-90 westbound traffic down off the pass. It is a three-lane, 20-span, prestressed concrete box girder design with a total length of 3620 ft (1103 m) on a 6% grade, Figure 6.39. The contractor, Hensel Phelps Construction Company, elected a construction method similar to those used in many German and Swiss designs where the area is environmentally sensitive. Because of the ecological and environmental sensitivity of the project site, construction of the piers was carried out under extreme space restrictions. The contractor was allowed a narrow access road for the full length of the project and additional work and storage area at each pier.” The 19 pier shafts have a hollow rectangular cross section with exterior dimensions of 16 by 10 ft (4.88 by 3.05 m), a wall thickness of 2 ft (0.61 m), and heights ranging from 35 to 160 ft (10.7 to 48.8 m), Figure 6.40. Twelve piers are supported on rectangular footings. The other seven piers are supported on pier shafts sunk through talus and till and keyed into solid bed rock, Figure 6.41. Piershaft diameter is 12 ft (3.66 m) with a maximum depth of shaft below the terrain of 80 ft (24.38 m). The superstructure was constructed in three stages, Figure 6.42. In the first stage, bottom flange and webs were constructed from a 330 ft (100 m) long movable launching truss, Figure 6.43. The two trusses used for constructing the “U” portion of the box section rested on landing wings at the piers, Figures 6.44 and 6.45, as the launching truss moved up the valley, sliding from pier top to pier top. The construction schedule called for one span every two weeks. The entire scaffold system was supported on six jacks to adjust for proper alignment, two jacks at the rear of the span or initial pier and four jacks at the advance section or next pier. The launching truss was designed to support the outside steel forms of the box section, Figure 6.46, and to facilitate removal of the inside forms,9 Figure 6.42. Track-mounted cranes installed at the top of the truss frame lifted and moved the inside forms from the web, hanging them on the truss so that they were moved forward with the advancement of the launching truss. Figure 6.47 is an interior view of the working area between trusses. Visible are the overhead track for the 15 ton ( 13.6 mt) cranes located near each web. Also visible are the cable hangars from the roof frame for the bottom slab support during casting. The steel trapezoidal box form used for con-

Steel Superstructure

Reinforced Concrete Superstructure Heavy Section .*I t.*LL-

Reinforced Concrete Superstructure Normal Section

FIGURE 6.36.

Rheinbticke

Dusseldorf-Flehe, elevation of pylon and cross sections.

13. 60.0

* 760.0

367.25

SD

11‘7.26

Bearingcordnion6

+

+

ffl +

+

+

+

+

+

+

+

+

+

+

G4

FIGURE 6.37. Rheinbticke Dusseldorf-Flehe, plan and elevation.

+ unreai3aimd =windbeE&lg + u-nad


FIGURE 6.38. Rheinbticke Dusseldorf-Flehe, end view of girder. struction was insulated with Styrofoam, Figure 6.48, and had heat cables installed (actuated if need be) to help maintain the temperature and rate of cure. Also, heat blankets were available to go over the section to reduce heat loss and maintain a constant temperature in cold weather. Concrete was batched from a plant erected neal the west abutment using the highway right-of-way. ‘[‘he contractor used three 8 cu vd (6.1 m3) ready-

307

mix trucks for mixing the concrete, which was then pumped to the proper location. Superstructure pours were about 300 cu yd (229.4 m3) and took about nine hours, using two concrete pumps and the track-mounted cranes installed in the truss frame. Concrete strength required was 5000 psi (34.47 MPa). The contractor obtained 3500 psi (24.13 MPa) in three days using $ in. (19 mm) aggregate. The 28-day strength ranged fi-om 6100 to 6600 psi (42.06 to 45.51 MPa). In stage two the top flange between the webs was placed. Metal f’orms, Figure 6.49, were supported from the bottom flange and webs, Figure 6.42.’ In stage three the two top flange cantilevers were placed, Figure 6.42, by a movable carriage that rode on top of’ the box cast in stage two, Figure 6.42. Upon completion of’ stage three, the transverse prestressing of’ the top flange was accomplished. The completed section is 52 f’t (15.08 m) wide, providing three traflic lanes. The Washington DOT sponsored the design. Three alternatives were prepared f’or bidding purposes. One was an in-house state design; the other two were prepared by outside consultants. ‘l‘he Dyckerhof‘t‘ & Widmann design proved to be

FIGURE 6.39. Denny Creek Bridge, perspective sketch.


FIGURE 6.40. Denny Creek Bridge, view of piers under construction (courtesy of J. L. Vatshell, Wash-

ington DOT).

the most economical. VSL Corporation was the subcontractor providing the prestressing expertise.

6.5 Span-by-Span Precast Bridges 6.5.1

LONG KEY BRIDGE, U.S.A.

Long Key Bridge in the Florida Keys carries U.S. Highway 1 across Long Key south to Conch Key. The existing bridge consists of 2 15 reinforced concrete arch spans, ranging in length from 43 to 59 ft (13.1 to 18 m) for a total bridge length of 11,960 ft (3645 m). The new bridge, presently under construction, is 50 ft (15.2 m) between centerlines and just north and parallel to the existing structure. It is a precast segmental box girder constructed by the span-byspan method and consisting of 101 spans of 118 ft (36 m) and end spans of 113 ft (34.4 m) for a total length of 12,144 ft (3701 m). The roadway width between barrier curbs is 36 ft (11 m), Figure 6.50, to accommodate a 12 ft (3.66 m) roadway and a 6 ft (1.83 m) shoulder in each direction. Figure 6.51 is

FIGURE 6.4 . Denny Creek Bridge, substructure types.

an artist’s rendering showing the precast V-piers with the 7 ft (2m) deep box girder segments. In the preliminary design stage three methods of segmental construction were considered: balanced cantilever, span-by-span, and progressive placement. The progressive placement method was discarded because it was felt (at the time) to be too new for acceptance in U.S. practice. It was later introduced on the Linn Cove Viaduct in North Carolina (see Section 6.3.2). This is the first use of a precast span-by-span method in the United States. The segments are transported from the casting yard to their location in the structure by barge. The segments are then placed with a barge crane on an erection truss, which is supported by a steel grillage at the V-piers. Each span has a 6 in. closure pour after all the segments have been placed on the erection truss and properly aligned. The essential operations are indicated in Figure 6.52. Segment weight is approximately 65 tons (59 mt). Each segment is placed on the erection truss on a three-point support and brought into its final position. It takes approximately four to six hours

Schematic of movable scaffolding

Stripped position

Staga 2

7

7

50

I

Overhead dollies

Stage

two

T

Jacks for grade, superelevation and camber

Stage one

LJacks

Rollers and jacks --

Stage three

FIGURE

6.42. Denny Creek Bridge, schematic of construction stages, from reference 9 (courtesy of the Portland Cement Association).

FIGURE 6.44. Denny Creek Bridge, view of landing FIGURE 6.43. Denny Creek Bridge, view of launching truss.

wings at piers (courtesy of J. L. Vatshell, Washington DO-I-). 309

FIGURE 6.46. Denny Creek Bridge, view of outside

steel forms (courtesy of J. L. Vatshell, Washington DOT).

Denny Creek Bridge, close-up view of landing wing (courtesy of J. L. Vatshell, Washington D O T ) . FIGURE 6.45.

to place the segments required for one span. The contractor has placed as many as three spans per week for a total of 354 ft (108 m) of completed superstructure per week and has averaged 2.25 spans per week. Another major deviation from United States practice in this project was the use of external prestressing tendons (located inside the box girder cell). This requires that the tendons be considered as unbonded for ultimate-strength analysis. Placing the tendons inside the box girder void allows the web thickness to be minimized. Tendon geometry is controlled by deviation blocks cast monolithically with the segments at the proper location in the span, Figure 6.53. These blocks perform the same function as hold-down devices in a pretensioning bed. The tendon ducts between deviation blocks or anchorage locations or both are composed of polyethylene pipe, which is then grout-injected upon completion of stressing operations- a corrosion protection system similar to that used for the cable stays on some cable-stay bridges. l”,ll

FIGURE 6.47. Denny Creek Bridge, view of interior

working area between trusses (courtesy of Herb &hell, FHWA Region 10).

FIGURE 6.48. Denny Creek Bridge, insulation on

exterior steel forms with installed heat cables (courtesy of Herb Schell, FHWA Region 10).

Span-by-Span Precast Bridges

Creek Bridge, vie\v 01‘ tnc.tal form used for stage-two construction (courtesy of J. L. Vatshell, Washington DOT). FIGURE 6.49. l)c~lny

Section

FIGURE

6.50.

at

pier

311

The external tendons overlap at the pier segment to develop continuity. The bridge is continuous between expansion joints for eight spans, 944 ft (288 m). After the closure pour reaches the required strength, the post-tensioning is accomplished and the span is complete. A 30 in. (‘760 mm) diameter waterline is installed inside the void of the box girder. The erection truss is then lowered and moved away from the completed span. The erection truss is handled at a one-point pick-up location by a C-shaped lifting hook, Figure 6.52. The truss is supported against the barge crane and moved parallel to the new bridge until it

Section at midspan

Long Key Bridge, typical cross section of superstructure.


312

The span by span erection concept utilizes a temporary steel assembly truss In conjunction with a barge mounted crane as shown. The steel truss 3 between the piers is equipped with post-tensiontng tendons along )m chord to facilitate adjustments for deflections and kwenng the LIUw ,,on completion of the span. PREVIOUSLY A55EMBLfP

5PAN 3

,

\ i A55E’40~Y

#.

1 TRUSS

FIGURE 6.52. Long Key Bridge, span-by-span erection scheme.

reaches the position for a new span, and the cycle is repeated. 6.5.2 SEVEN MILE BRIDGE, U.S.A.

The Seven Mile Bridge, Figure 6.54, in the Florida Keys carries U.S. Highway 1 across Seven Mile Channel and Moser Channel from Knights Key west and southwest across Pigeon Key to Little Duck Key.

The existing structure consists of 209 masonry arch spans, 300 spans of steel girders resting on _ _ masonry piers, a n d a s w i n g s p a n o v e r M o s e r Channel. The spans range in length from 42 ft 7t in. (13 m) to 47 ft 4$ in. (14.4 m) for the masonry arches and from 59 ft 9 in. (18.2 m) to 80 ft (24.4 m) for the steel girders resting on masonry piers, which along with the 256 ft 10 in. (78.3 m) swing span, produce a total bridge length of 35,716 ft 3 in. (10,SSS m).

313

Span-by-Span Precast Bridges

PERSPECTIVE VIEW

DETAIL 2

ELEVATION

FIGURE 6.53. Long Key Bridge, typical tendon lay-

FIGURE 6.54. Seven Mile Bridge, artist’s rendering. The new bridge, presently under construction, is located to the south of the existing bridge. It is a precast segmental box girder constructed by the span-by-span method with 264 spans at 135 ft (41.15 m), a west-end span of 81 ft 7$ in. (24.88 m), and an east-end span of 141 ft 9 in. (43.2 m) for a total length of 35,863 ft 44 in. (10,931 m). The roadway requirements are the same as for the Long Key Bridge and the cross section is almost identical, Figure 6.50. Seven Mile Bridge crosses the Intracoastal Waterway with 65 ft (19.8 m) vertical clearance, and its alignment has both vertical and horizontal curvature.

The consultants, Figg and Muller Engineers, Inc., used the same concepts as had been used for the Long Key Bridge, except they omitted the V-pier alternative in favor of a rectangular hollow box-pier scheme that is precast in segments and post-tensioned vertically to the foundation system. As mentioned in Section 1.9.3, the contractor elected to alter the construction scheme in this bridge from that of the Long Key Bridge by suspending the segments from an overhead truss rather than placing them on an underslung truss. The essential operations for construction of a typical span are as follows: 1. 2.

3.

4.

Transportation of all segments by barge to the erection site. Assembly of all segments in a span (with the exception of the pier segment) on a structural steel frame supported by a barge. Placing the pier segment on the pier adjacent to the previously completed portion of the deck with the overhead truss working in cantilever. Launching the overhead truss onto this newly placed pier segment.

314

Progressive

and

Span-by-Span

Construction

5.

Lifting in place the entire assembly of typical segments with four winches supported by the truss. 6. Post-tensioning the entire span after the closure joint has been poured between the finished span and the new span. 7. Launching the overhead truss to repeat a new cycle of operations. After a period of adjustment, the method has allowed a speed of construction equal to that for the assembly truss scheme used for the Long Key Bridge. One complete span may be constructed in one day, and as many as six 135 ft spans have been placed in a single week. Figure 6.55 shows the assembly of segments being erected in a typical span.

6.6

Design Aspects of Segmental Progressive Construction 6.6.1 GENERAL

.,

The use of temporary stays to carry the weight of segments during construction induces only a normal compression load in the deck and a very limited amount of bending. Consequently, the static scheme of the structure during construction is very close to that of the finished structure. This is a significant advantage over the conventional cantilever construction scheme, where continuity of the successive cantilever arms results in two static schemes significantly different between construction and service. Because of this similarity of static scheme throughout erection and service, it is expected that the layout of prestress tendons found in cast-inplace structures or in span-by-span construction

FIGURE

span.

6.55.

Seven Mile Bridge, erection of a typical

of

Segmental

Bridges

should be applicable to progressive construction, with the added advantage that the tendons can be regularly stressed and anchored at the successive joints between segments in a simple manner. On the other hand, progressive construction differs in several aspects such as pier design and deflection control during construction, calling for a more detailed examination.

6.62

REACTIONS ON PIERS DURING CONSTRUCTION

Construction of a typical span proceeds in two stages, as shown in Figure 6.56: (1) pure cantilever erection, of a length a from the pier, and (2) construction with temporary stays on the remaining length (L - a). Length a should be selected (within the nearest number of segments being placed) such as to keep the girder load moments over the pier within allowable limits. Assuming that this moment is of exactly the same magnitude as the fixed end moment of a typical span under the same unit load W, one may write: Wa2 WL’ -=-

2

FIGURE

12

6.56. Progressive construction, deck reactions on piers.

Design Aspects of Segmental Progressive Construction

for a constant-depth girder, which is the general case for- progressive construction. Thus:

over the support (15%), 2.6 x 1.15 =

315 3.0 12.5 ksf‘

0 = 0.408L = 0.4OL For (1 = 0.4OL the moment over the pier is equal to ,M = 0.08WL’. l‘he moment over the preceding pier, for a structure with a large number of’ identical spans, is equal to 0.26&\1. Therefbre. the reaction over the pier at the end of’ this first stage of construction can be easilv computed as:

-I‘he dif‘ference i s s m a l l a n d usuallv m o r e t h a n offset hy the fact that horizontal loads during construction are smaller than during service.

R = 0.4OWL + 1.268 x 0.08WL = O..5OW’L During the second construction stage the lveight of the remaining part of’ the span is supported b:, t h e temporarv stays, which are anchored in the rear span as close as possible to the previous pier so as not to induce undesirable variations of. moments i n t h e last c o m p l e t e d s p a n . ConsequentI!,, t h e lveight of’t hat part of’t he span induces in the pier a react ion equal to: 1 .io 0.6W’L + - = 1.02WL 1 . o o ___ The total reaction during construction applied to the pier is t bus: R = 0.5OWL + 1.02WL = 1.62WL as opposed t o R = N’L for cast-in-place or spanh\--span construction. ‘l‘his temporary increase of’ girder load reaction of’ 62% \vill eventuall!- \,anish Ivhen construction proceeds. It is important to \.erifv how critical this pier temporary overload ma! be f’or the design of’ the substructure. Taking the example of’ a 150 to 200 f’t span, the average loads are as follows fi)r a 40 f’t wide bridge designed f’or three lanes of’ traf’fic: Girder load Superimposed load Equivalent live load including impact

8.0 ksf 1.5 ksf 2.6 ksf

The maximum reaction during construction compares jvit h that after completion as follows (values given are the ratio between reaction and span length): 1. During construction, 1.62 X 8.0 = 2. Completed structure: a . Girder load 8.0 b. Superimposed load 1.5 C. Live load, including provision tot- continuity

13 ksf

As shown previously, progressive construction of’ a typical span entails two successive stages: Cantilever construction on a length (I ~l‘emporary suspension by stays on the remaining part of’ the span (L - n) .I‘his second stage induces small deflections and rotation, provided that the vertical component of’ the sta!- loads balances the total deck weight. On the other hand, the first-stage construction not only creates substantial deflections but also changes the geometric position of’the entire span, as mav be seen in Figure 6.5f. The xveight (Wa) of’ the deck section produces: A rotation of’ the previous span, w,, which will project at the f’ollowing pier and create a vertical deflection, J, a deflection of’ the cantilever proper, yr a rotation at the end of the cantilever, wL, which Lvill p r o j e c t a g a i n a t t h e f’ollowing p i e r i n t o a deflection mt (L - n) Altogether the total deflection is: Wa’ (2Ll v5 + 4nL - n’) I’= 2 4 E I If’we let 14 = N/L, the deflection can then be written as: WL1 u’(2 fi + 4U - u’) )‘= 2 4 E I With u = 0.4 as assumed betore, the total deflection is: WL4 y = 0.0327 EI where W = unit deck load, L = span length,

316

Progressive and Span-by-Span Construction of Segmental Bridges IZ’ = concrete modulus, I = section inertia.

A simple parametric analysis will reveal the importance of this problem. If W is the specific gravity of concrete and A the cross-sectional area, then W = GA. It was shown in Chapter 4 that the efficiency factor of a box section is:

=

I - = 0.60 to 0.63

Ac,c,

If the section is symmetrical, c, = cz = 0.5 h (h = section depth), and I = 0.157 Ah” max. If c, = 0.33 h and cg = 0.67h, which is the practical dissymmetry of a box section, I = 0.133Ah2 min. For all practical purposes, assume I = O.l4Ah*. The deflection then becomes:

Ey = 0.23ZL2 + * t 1

L-o

Because the construction proceeds rapidly, E should be taken for short-duration loading; that is, E = 800,000 ksf; the specific gravity of concrete is W = 0.15 kcf. The slenderness ratio L/h varies between 18 and 22. Results are shown in Figure 6.58. Construction of a 200 ft span, for example, with a slenderness ratio of 20 will be accompanied by a deflection under girder load (without prestress) at the next pier of 8.3 inches. The construction method is therefore very sensitive to concrete deflections, which are magnified by the great lever arm of the first-stage construction of the span projecting its intrinsic deformation to the following pier. Fortunately, prestress will give a helping hand and contribute to substantially decreasing the girder load deflection. The minimum prestress required at this stage is to balance the tensile stresses induced by the girder load moments. With the same notations as above, one may compute the prestress force and the corresponding moment for three positions of the neutral axis:

Efficiency factor Distance fi-om centroid of‘ prestress to top fiber Eccentricity of’ prestress Lower central core Lever arm of prestress Prestress moment (ratio of’ girder load moment)

1

FIGURE 6.57. Progressive construction. def’ot-mations. For an efficiency factor p = 0.65 the corresponding values would be: 0.58

0.47

0.39

The prestress will therefore reduce the deflections by the same amount-that is, approximately half the total girder load deflections. The resultant deflection (girder load + prestress) still remains very significant as soon as the span length is above 150 ft. These deflections must be taken into full account to compute the camber diagram (for segment precasting). The next important point to consider here is the second-stage construction of a typical span when the remaining part of the girder is suspended from the temporary stays. The concrete girder and the group of stays form an elastic system that supports the applied loads: girder load for the segments already in place, swivel crane and new segment

c,lh = 0.5 c,lh = 0.5

c,lh = 0.4 c,lh = 0.6

c,lh = 0.33 c,lh = 0.67

p = 0.60

d = 0.05h

p = 0.60 d = 0.05h

p = 0.60 d = 0.05h

e = 0.45h 72/c, = 0.30h

flc, = 0.36h

0.75h

0.71h

0.45 = 0.60 0.75 E

c = 0.35h

c = 0.28h 0.40h

r21c, =

0.68h

317

Design Aspects of Segmental Progressive Construction

I

I

IC

FIGURE 6.58. Progressive construction, deflections.

traveling over the bridge with the trailer and tractor. -Two examples have been considered to show the relative response of the various components of this elastic svstem toward the application of a load.

1. 108) (JJm) s p a n This was one typical span of’ the Rombas Viaduct. The span has been assumed to be completed except for the pier segment over the next pier. For this construction stage, the swivel crane and the new segment apply to the staved cantilever a load of 88 tons (80 mt). In view of the great stiffness of the concrete girder compared to the group of stays, the total moment induced bv the load remains ahnost entirely in the concrete girder and there is only a small spontaneous increase of the stay loads, as shown in Figure 6.59. The magnitude of temporary prestress in the deck must be designed accordingly to keep all joints under compression for all intermediate loading cases. This example is taken from a recent design for a large project in Europe

2. 260 ft (80 m) span

where progressive construction was contemplated for a viaduct with a large number of identical 260 ft (80 m) spans all made up of 26 segments 10 ft (3 m) long. Figures 6.60 and 6.61 show the distribution of moments between concrete girder and temporary stays at three successive stages of segment placing: segments 15, 20, and 25, respectively. The first nine segments are placed in cantilever; the following 15 segments are suspended from tern porary stays, while the last typical segment and the adjacent pier segment are placed without stays. The proportion of the load (and corresponding moment) taken by the stays increases as the cantilever length increases and, when the last segment is placed, more than half the load is supported by the stays. For verv-long-span stayed bridges, this distribution of load between stays and concrete girder reaches the situation where the load is almost entirelv supported by the stays and the concrete girder’is subjected only to an axial force, except in the area of the longest stays. The consideration of distribution of loads and moments between stays and concrete girder has an important aspect during construction-that is, the accuracy of the tension in the stays and consequences of an accidental deviation between computed values of stressing loads in the stays and their actual values in the field. For example, take the simple case of a span L with 40% built in pure cantilever and the remaining 60% suspended by stays (see Figure 6.56). The moment over the pier due to the second-stage construction load is M = 0.42WL’. Assume that an accidental deviation took place of 5% between the design loads for the stays and the actual values obtained in the field (owing to friction in the jacks, inaccuracy in the pressure gauges, and so on). As a result, an additional moment will appear over the pier of AM = & 0.42 WL’ = 0.021 WL’. The corresponding tensile stress at the top fiber (assuming the error in stay loads was to reduce the theoretical values by 5%) can be easily computed by:

Af=AM’-cI

0.02 IZAL2c, APC,C, = 0.0217X~ PC2

With W = 0.15 kcf, p = 0.60, and cp = 0.60/z:

Af=

0.0088:

318

Progressive

and

FIGURE 6.59.

Span-by-Span

Pt-ogwssi\t’

Construction

co11structiotl.

‘l‘hc stress in kst’ t’or L//r = 20 (slcndert~ess ratio) i3 the l’ollo~vit~g 1’01. sewx~l spmi l e n g t h s : L (11) .I/ = O.li.il.

(tdl)

100

130

1X

“6

“00

250

3 .i

-l-l

I‘his stress is not critkil 1’01. short spans but I~;II hecome sigtiificatIt t’or lotig ones. .I‘he simple dervation given above sho~3 that control of the stabtetisiottitig operations at the site shoitld albx~~s be ott the salt side Ivith due allo\vatice fol. iti;icciIr;ic~~. .-I dcviatioti in the tension of‘ the stavs bill aiso al‘f’ect the deflections during constructio;l. Without the presence of the stavs the total deflection over the next pier due to the load 011 the length (L - n) w o u l d be:

which gives t’or u = 0.4 as befixe:

of

Segmental

incwase

Bridges

of’ hta\ Ioatlittg

A-\ssutiie that the itiaccutx~~ of the sta\ lo;itts lea\,es itI the concrete girder 5% of its O\VII lveigtit to be carried lx bending: the resulting deflection m.et the pier ~vould be:

.I‘his value should be compared to ttte ef‘tect of‘ the first-stage cotistructiotl, lvhich bxs pre\iortsl\ gilen :1s:

Ill sLltllm;II‘~ ( il 5% de\iatioti of the st;t\. tctisiotl loads will increase the cantilever defectiotl due to girder load by 36%. Considering the twneficial etfeet of prestressitlg fi)r the latter, \ve see that approximately 7% deviation ot’the sta!. l0;1d produces the same defection 2s the first-stage cotistructiot~ loads including prestressing. .I‘tiis 41~0~3 that the d e f l e c t i o n s a r e itnportant, particulart\ fi)r lotlg spa~is built in progressive construction, hut that proper deflection control is an excellent tool to

319

References

BY CONCRETL

3 S.GML”T, *~~~~~~ I” CANmEYER FIGURE o

F I G U R E 6 . 6 0 . I’rogrc\si\c o

f

lttolttclt1

tx~t\\wll

sl;l\s

~ott~tt~tt~~tiott. clistt-ibrttiott ;Illcl

gitxlcT.

Let-it\ t ttal st xwcs in t tie c-ottc‘rele girder ,tt-e ;11\\-;t\ 5 Lcpl \vittiitt alfo\v;lt)fe littiila.

f

6.61.

OIMDCR

I5 ?2WR”DED SCfMLNTS----i w’ 9*h

I’ro~t~c’~~i\ c consli~uc lion. (li\tt~itttttiott (

IlIoItlcI1t twt\\cul

\t,t\s

‘Incl

~itxlcl

Possit)l\. ;I third fatnil\ 01’ tendotta mtde of intet~tt;il st;i\ s b,‘itli ;I dl-aped profile and attchot-ed over the piet-s iii t tie di~tptirqp~. the put.posc of’ lvttictt is to sttpplettietit both ottiet- f;itiiilies Ichile t~ecittcittg the ttet stie;tt- sttwses in the \vet)s txcattse of’ the \,ctTiCd cotiipotietit o f prestt‘ess.

References Because the silatic wttente at the cticl of’ cacti conslrttctiott 5tep ia idctttical to that of‘ ;I ca\t-itt-place slntct tire, (tie pet~tttatietit tendotis can Ix ittstalletl i i i t t t r sttxtctttw itiitiiediatel~, vAthut the tratisit i o t i s i t u a t i o n s I-eqttil-ed t,y o t h e r cotistrttctioti ttt~tlioclologies such 2s itict-emetital lautictiitig. ~1 t\ pical pi-esttws la\ out for progressive COIIstruction b,iff thus include:

1

H .

\Vitttoht.

sti-tictiott

.A wcottcl f’atttil\ of‘ tendons located aloti~ the spati iii the tmtlom flatige and Am ;itictiored in blisters inside the tms section. L’suall\. the top atid bottom hlisterh ~tre,joitieci to ;t \veb rib, allowing tetiipotxt-1 pi-estt-css hi-s t o Ix atictiored during s e g m e n t placittg.

~:oIIcl~cIc

Steppittg

Form\\-0t.L

F i r s t Iittet-~tatiortal sign,

t’;~pcr

SP

S\ tttlx~sit~itt 2%2X. .-\C:l

.\tttt~Cc;ttt C:ottct-ete

”

H .

\Vitttottt.

Itlstitute.

“Die

lkti-oil.

Siitttt H.

-I

Heft

I‘mg,

~I‘cchniclties

pot-tatioit

Keswrch

ing. \‘ot.

Zawrtl rrtcd

196X.

“ R e c e n t lkvelol~itiertt o f C o n -

structiott

in

(Ionct-etc

Kewi-tl

B r i d g e s . ” I’rans-

66.5. Kridgr

2. Procwd~nq~ of the 7‘,n,~.rportcctrorl

Rof~rd Co,,Jrrf~rrcf~,

.\fo.,

Itttc’i~n;ttiott;tI

im BriicLelltxttt.”

4”. Ik/ctttt,el~

M;rl~-<:tlullg

D.C.

Constrttctioii),

\~ol-scttt)l-its-

of I‘i-:t\eliitg

Bridge

I‘hul. “Spatttttxton

Hem.

lWi9.

VOII

a~td Sttxctui-al kkgittceriitg, Congress. ;\tttatet-da~tl, .\l;r\ H- 13, lYi2.

.Association

3

c:or1-

E:qttipilit’itt.”

13ritlgc lb l’t~t~lic~ttiott SI’-23.

\‘ei-~eiidrtttg

iii Bridge

Rlxlge

(Zorict-ctc

tuiigeil bictn Bt~iickeithu” ( I‘ttc L‘w Foi~rn~~o~L

.-I fit-t f;ttttil\ of. tettclotts located in t h e t o p flange o\‘et‘ 1tie \~t~ious piers, h,itti atictiol-s s~tiitiietl-icall~~ loc;tlecl in hfistet-s, the purpose of Ivtiicli is to resist ticgtti\ c tttotttetits 01 er t tic suppot-ts.

"l'lY?sllw\ed

bitti

Satioital

Ettgtneel--

RPWIW/I

S e p t e m b e r 2.5’Li . 19i8. S t . Imttis;,

Ac;tdeiii\~

ot S c i e n c e s .

~l’astiittgtott.


320

5 . U. F i n s t e r w a l d e r a n d H . Schambeck, “ D i e Elztalbriicke,” Der Buuzngenieur, H e f i 6 , J u n e 1 9 6 6 , a n d Heti 1, January 1967. 6 . H . .I‘hul, “Bt-iickenbau,” Heft 5, May 1966.

7.

Anon.,

Beton- und Stnhlhetorrbnu,

“Bau del- Loisachbriicke

Dyu&g-Berichte

bei O h l s t a d t , ”

19713, D y c k e r h o f t 8s N’idmann,

AC, M u n i c h . 8. .Anon., “Bauausf‘iihrLlng d e r XutobahnbrLicke ilbel die Loisach bei Ohlstadt,” Dydq-Betichte 1972-5, Dvckerhoff & Widmann, AG. 1lunich.

9.

10.

AIlon., “ D e n n y Creek-FrallkliII E‘,~lls \‘i;ttluct. Washington,” Bt-idge Report SK 202.01 E, l’ortl;t11(1 Cement Association, Skokie, 111.. 1978.

Anon., “Florida’s L ong E(e) Bridge to Ltilile l’rcca\t Segmental Box Girder Span-b\-Spa11 <:onstI.uctiot~,” Bridge Report, Post ‘I‘ensioning IIIstitute. l’hoellix. Arizona, January 1979.

1 1 . W a l t e r Podolny, J r . , “;\n O\ et-k iew of t’reca\t Prcstressed Seynental Bridges.“Jo~c,-,/NI of /AC, ~w\l,r\wtl Concrete Ztt,ditute. 1.01. 24, So. 1, jaI~II;in -Fe111 udn 1979.

7 Incrementallv Launched Bridges J

7.1 INTRODUCTION 7.2 RIO CARONI, VENEZUELA 7.3 VAL RESTEL VIADUCl-, ITALY

7.9

7.9.2 7.9.3

7.4 RAVENSBOSCH VALLEY BRIDGE, HOLLAND 7 . 5 OLIFANT’S RIVER BRIDGE, SOUTH AFRICA 7.6 VARIOUS BRIDGES IN FRANCE

7.8.1 Mtihlbachtalbriicke, Germany 7.8.2 Shepherds House Bridge, England

7.1

Introduction

-The concept of’ incrementally launched segmental pres’ressed concrete bridges was described in Section 1.9.5. .Fhis chapter will describe the implementation of this innovative concept in several representative projects. Since the in~plementation of the incremental launching technique on the Rio Caroni Bridge, some eight\ bridge superstructures have been consrructed 1;~ t h i s m e t h o d t h r o u g h 1 9 7 6 , w i t h gradual refinements and improvements in the method.’ Bv concentrating the casting of segments behind an ;Ibutment with a temporary shelter, if required, this method can provide the same quality control procedures and quality of concrete that can be achieved in a concrete ‘precasting plant. It minimizes temporarv falsework, extensive forming, and other teniporary expedients required during construction bv the conventional cast-inplace on falsework meihod. Basically the method entails incremental fabrication of the superstructure at a stationarv location, longitudinal movement of the fabridated segment an incremental

OF

INCREMENTALLY

LAUNCHED

BRIDGES

Type, Shape, and Dimensions of Superstructure Span Arrangement and Related Principle of Con-

struction 7.9.4 Design of Longitudinal Members for Flexure

7.6.1 Luc Viaduct 7.6.2 Creil Viaduct 7.6.3 Oli Viaduct 7.7 WABASH RIVER BRIDGE, U.S.A. 7.8 OTHER NOTABLE STRUCTURES

DESIGN

7.9.1 Bridge Alignment Requirements

7.9.5

and

Tendon Profile Casting Area and Launching Methods

7.9.6 Launching Nose and Temporary Stays 7.9.7 Piers and Foundations 7.10

DEMOLITION OF A TAL LAUNCHING REFERENCES

STRUCTURE

BY

INCREMEN-

length, and casting of a new segment onto the one previously cast. In other words, the procedure can be considered as a horizontal slip-form technique, except that the fabrication and casting occur at a stationary location. Stringent dimensional control, however, is an absolute necessity at the stationary casting site, since errors are very difficult to correct and result in additional costs in launching.’ Straight superstructures are the easiest to accommodate; however, curvature (either vertical or horizontal) can be accomplished if a constant rate of curvature is maintained. If the grade of the structure is on an incline, it is preferred to launch the structure, wherever possible, downward. Where the fall is 2’$%, the superstructure has to be pushed or held back, depending upon the coefficient of friction. Where the fall is in excess of 4%, special provisions are required to prevent a “runawav” superstructure during launching.’ To the authors’ knowledge, this situation has never occurred. Piers, either temporary or permanent, should be designed to resist the lateral force produced by the launching operation. A friction force varying from 4 to 7% has been considered for de321

Incrementally Launched Bridges

322

sign purposes, although values of’ only 2 to 34% have been observed in the field. At present, it is felt that this system cm be used for superstructures up to 2000 t‘t (610 m) in length; fbr longer structures incremental launching is accomplished f’rom both abutments toward the center of’ the structure. .l‘he technique has been applied f’or s p a n s up to 200 f’t (60 m) lvithout the use of’ temporal-v supporting bents and for spans u p t o 3 3 0 ft ( 1 0 0 m ) with such bents. Girders IISUally have a depth-to-span ratio ranging f‘rom one-tbvelfth to one-sixteenth of’ the longest span and are of’ a constant depth. ‘l-he launching nose has a length of’ approximately 60% of’ the longest

either straight or curved: holvever, cur\‘ature, either vertical or horizontal, must be of’ a COW stant radius. 2.

As mentioned above, strict dimensional control during casting is required. .4n\ mistakes in casting are difficult and expensive to correct, especially if the\. are not discovered until af.ter some length of’ bridge has been launched.

3.

l‘he superstructure must be of’ a constant section and depth. .l‘his is a disad\.antage in long s p a n s , lvhere a variable-depth section \vould provide a better econom\ of’ materials.

4.

Considerable area is required behind the abutment(s) for casting the segments. In some project sites this may not he feasible.

spa11.

.I‘he principal advantages of the launching method are the following’: 1.

2.

3.

4.

incremental

No f’alseworh is required f’or the construction of’ the superstructure o t h e r t h a n possibl! f’alsework bents to reduce span length during construction. In this manner cantilever stresses during launching can be maintained lvithin allo~vable limits. If‘ fhlsework bents should prove to be impractical, then a system of‘temporare stays can be used as indicated in Figure I .63. Obviously, depending on site conditions, ;Inv or all combinations of’ temporary bents, launching nose, and temporar!. stays may be used, the point being that conventional use of’ f’alsework is cqreatlv minimized. -l-his is partitularly interesting f’or projects in urban areas or spanning over water, highways, or railroads. Depending on the size of’the prqject there can be a substantial reduction in form investment. Because casting of’ the segments is centralized at a location behind the abutment, the economic advantages of mass production and a precasting plant operation can be duplicated. ‘l-he method eliminates transportation costs of segments cast at a fixed plant and transported to the site. It eliminates heavy cranes or launching trusses and associated erection costs.

5.

It eliminates epoxy joints. Since epoxy is not involved, construction can continue at lower temperatures.

6.

Camber control and other geometry controls are easily obtained.

Disadvantages are as follows: 1. As mentioned in Section 1.9.5, bridge alignment fbr this type of’ construction must be

In the present state 01’ the art of i~icrementall\ launched bridges there appear to be basicall\ tlvo methods of’ construction, \\,tiicli we shall call co?/tirluou.c ctstrng and trnluncd c a s t i n g . .l‘hey are diff‘erent in mode of’ execution and in their areas of‘ utilization. The continuous casting method is some\\.hat analogous to the span-by-span method. and halanced casting is similar to the cantilever method. .l‘lie continuous casting method is generall! used for long viaduct-type structures with numerous equal (or nearI>, equal) spans. Its principal characteristics are the f,llo\ving: 1.

Entire spans, or portions of’ spans, are concreted in fixed forHIS. The f’orms are reused, as in the span-bv-span method, except that the f o r m a a r e fised instead of mobile and are moved from s p a n t o s p a n . Subsequent spans (or portions of a span) are cast and joined to the one previously cast, and the superstructure is progressively launched.

2.

Usually the casting area behind the abutmeIlt is long enough to accommodate either a span length plus launching-nose length or some multiple of span segment length plus launching-nose length. Operations involve successive concreting and launching. The principal phases aI-e: forming; placing of’reinf’orcing and tendons; concreting and curing; tensioning and launching.

3.

4. 5.

The two types of’ superstructure cross section used ha1.e been box girder and double I‘. Longitudinal prestressing consists of’ two fhmilies of tendons: tendons concentricall\ placed and tensioned before launching, and tendons placed and tensioned af’ter launch-

Rio

Caroni,

Venezuela

323

ing-that is, negative-moment tendons over the supports and positive-moment tendons in the bottom of the section in the central portion of the span. The balanced casting method is used for smaller projects up to a total length of 650 ft (200 m). It is used for symmetric three-span structures where the central span is twice the end span. Its principal characteristics are: 1.

Concreting of segments is accomplished symmetrically with respect to a temporary support located in the embankment behind the abutments. This method is similar to the balanced cantilever except that the forms are supported on the embankment fill. 2. Two areas of casting are required, one behind each abutment. The half-superstructures are constructed at opposite ends of the project. The distance between the abutment and the axis of the temporary massive support is generally slightly less than one-fourth the length of the project. 3. After the two half-superstructures have been concreted on the access fill, the two halves are launched over the piers and joined at midspan of the central span by a closure pour, which usually has a length of 3 ft (1 m). 4. Longitudinal prestressing consists of three families: cantilever tendons for each segment, located in the upper portion of the cross section and stressed before launching; continuity tendons, tensioned after closure and situated in the lower flange; and provisional tendons, located in the lower flange, tensioned before launching, and opposing the cantilever tendons. There are two methods of launching. The method used on the Rio Caroni Bridge, Figure 1.67, has the jacks bearing on an abutment face and pulling on a steel rod, which is attached by launching shoes to the last segment cast. The second, and more current, method is essentially a lift-and-push operation using a combination of horizontal and vertical jacks, Figure 7.1. The vertical jacks slide on teflon and stainless steel plates. Friction elements at the top of the jacks engage the superstructure. The vertical j a c k s lift the superstructure approximately & in. (5 mm) for launching. The horizontal jacks then move the superstructure longitudinally. After the superstructure has been pushed the length of the hor-

FIGURE 7.1. Incremental launching-jacking mechanism (courtesy of Prof. Fritz Leonhardt). izontal jack stroke, the vertical jacks are lowered and the horizontal jacks retracted to restart the cycle.’ Figure 7.2 is a schematic depiction of this cycle. To allow the superstructure to move forward, special temporary sliding bearings of reinforced rubber pads coated with teflon, which slide on chrome-nickel steel plates, are provided at the permanent piers and temporary bents, Figures 7.3 and 7.4. A sequence of operations showing the bearing-pad movement on the temporary bearing is depicted in Figure 7.5. A temporary bearing with a lateral guide bearing is shown in Figure 7.6. 7.2 Rio Caroni, Venezuela The design for this structure was proposed by consulting engineers Dr. Fritz Leonhardt and Willi Baur of the firm Leonhardt and Andra, Stuttgart, West Germany, in an international competition. Design and planning occurred in 1961 and construction in 1962 and 1963. This structure, Figure 7.7, consists of a two-lane bridge with end spans of 157.5 ft (48 m) and four interior spans of 3 15 ft (96 m), for a total length of 1575 ft (480 m).’ The site provided some formidable construction problems. The Rio Caroni River during flood stage reaches a depth of 40 ft (12 m) with velocities of 13 to 16 ft/sec (4 to 5 m/set), thus eliminating the consideration of a cast-in-place concrete superstructure on falsework. Balanced cantilever segmental construction was considered; however, the interruptions during high-water periods would require an extensive construction period with attendant high costs.3 The proposed method consisted of assembling and prestressing the entire length of bridge on

324


(a)

FIGURE 7.3, Incremental launching-longituclinal

section of launching bearing, from reference 3 (courtes, of the American Concrete Institute).

fb)

FIGURE 7.4. Launching

BI idge, Indiana.

Cd)

FIGURE 7.2. Schematic of’ launching jack operation.

(cr)‘Lit‘t. (h) Push. (c) Lower. (cl) Retract.

land adjacent to the bridge site, using precast segments, and launching in a longitudinal direction, over the piers, into final position. Temporary piers were used at midspan of’ each interior span to produce ten equal spans of 157.5 ft (48 m) during the launching of the superstructure. Accommodation of’ on-site assembly of’ the total superstructure required a 1600 f’t (500 m) long f’abrication bed to the rear of one abutment, which was partly excavated in rock and had to be backfilled and compacted upon completion of’the project. At the f’ar end of

bearing, I+‘abash

River

this fabrication bed stationary steel forms were installed to cast the precast box segments, which were 18 ft 4 in. (5.6 m) high and cast in 30 ft (9.2 m) lengths. After the precast segments attained sufhcient strength thev were stripped from the f’orm and positioned in the fabrication bed to correspond with their location in the final structure. The segments were moved f’rom the form on wooden rails accurately positioned in the assembly bed, employing formica sheets and a petroleutn-base lubricant between the bottom of the segment and the top of the wood rails, Figure 7.8. A space of 1 ft 4 in. (40 cm) wx lefi between the precast segments f’or an in situ joint. Accurate positioning of’ the segments in the assembly bed was required before casting of the joints. To avoid shrinkage damage, the joints were cast during the second half’ of’ the night so that the temperature expansion of the precast segments during the heat of the day wc~ulcl compensate for the shrinkage in the cast-in-place joint.” After the joints were cast, concentric prestressing located inside the box and passing through openings in the web stif’f’ening ribs, Figure 7.9, was prestressed with a force of’ 5000 tons in one opera-

Rio Caroni, Venezuela

325

iI v-5’ +

FIGURE 7.7. Completed Rio Caroni Bridge, from

reference 3 (courtesy of the American Concrete Institute).

FIGURE 7.8. Precast

segments in assembly I)ed (courtesy of Arvicl Grant).

FIGURE 7.5.

operations.

l‘emporarv

FIGURE 7.6. I~~c~xmrc~r~al

sliding bearing, sequence of

I;tut~chitlg-tetll~)ot~~l~~ bearing and lateral guide bearing (courtesy of Prof. Fritz Leonhardt).

tion. The prestress tendons were continuous around a large half-round concrete block at one end of the structure, Figure 7.10. This block reacted against a number ofjacks and a 10 ft (3 m) thick concrete bulkhead wall. Bv activating the jacks between the block and the bulkhead and causing a movement of 9 ft (2.X m) in the stress block, the initial prestress force was induced into the tendons. The prestressing tendons were not attached to the webs. To reduce the hazard of an\ accidental elastic instability condition, temporarv steel bracing frames were installed at 60 ft (20 m) intervals.” The 33 ft 10 in. (10.3 m) top flange of the box girder section was transversely prestressed, Figure 7.9. Upon completion of the prestressing operations the superstructure was ready for launching over the temporary and permanent piers to its final POsition. To maintain acceptable levels of concrete stresses, as the girder was launched over the 157.5 ft (48 m) spans, a 56 ft (17 m) tapered structural steel launching nose was attached to the leading

326


-’

/r9’-l0’16’-51-y.---uI-9~-10’~

FIGURE 7.9. Rio Caroni, girder cross section, fl-om reference 3 (courtesy of the American Concrete Institute).

FIGURE 7.10. Kio Ckror~i, patressing tesy of Awiti Grant).

Mock (wur-

end of the superstructure, Figure 7.11. Two double jacks with a total capacity of 600 tons, mounted against the bridge abutment and pulling on steel rods fastened to the girder, provided the horizontal force required for the longitudinal launching movement. To accommodate movement over the piers, two sliding bearings were provided at each temporary and permanent pier top. These bearof chrome, polished steel plates ings conststed which supported teflon covered bridge bearings which were placed in an inverted position such that they bore against the underside of the girder and slid on the steel plates. After a launching movement of 3 ft (96 cm) in the longitudinal direction

FIGURE 7.11. Rio Carom. Ltu1~111ng nose, 11 OIII refe~ence 3 ((ourtesy of the American Concrete Institute).

the operation was halted to allow the entire superstructure to be jacked vertically, simultaneously at all piers. The teflon plates were then moved back to their original position (the one they occupied when the launching operation started) and rotated 180 degrees, with respect to a vertical axis, to compensate for any one-directional movement of the teflon coating. Longitudinal launching movement occurred at a rate of 24 in./min (6 cm/ min); thus, one 3 ft (6 cm) increment of movement took 16 minutes. A total cycle of operation, after subsequent synchronization, which included the simultaneous jacking at 22 locations and repositioning of 22 teflon bearings, required 30 minutes

Val Restel Viaduct, Ztaly for each 3 ft (96 cm) of movement. In this manner, a daily movement of 63 ft (19.2 m) could be accompiished. The required initial jacking force for launching was 220 tons; this gradually increased to 400 tons f’or the total girder weight of 10,000 tons, which indicates a friction of 2 to 47c.3 After the launching operation was completed, the initial concentric prestressing tendon profile was changed to accommodate the loading condition in the superstructure after temporary piers were removed. To accomplish the change in tendon profile, special L-shaped rods were installed so that the! projected upward through the top flange or downward through the bottom flange, the tendons being cradled in the U rods. The rods ivere then jacked simultaneously at 24 points up~\a~-ct or downward, depending on their location. During this operation the half-round stress block, Figure i. 10, ~vas gradually released such that upon final positioning of the tendons it had retracted 8 ft 6 in. (2.6 III). After the tendons had been relocated, they lvere attached to the \veb and concreted for corrosion protection.” The procedure used for the construction of the Rio Caroni Bridge, although technically adequate, is prohibitively expensive. The methodology has since been refined such that segments are cast directly behind the abutment in lengths of 33 to 100 ft (10 to 30 m) and incrementally launched after curing of the last segment cast.’

7.3

Val Rested Viaduct, Italy

Because of rugged mountain terrain the alignment of’ a 1050 ft (320 III) portion of this viaduct re-

327

quired a sharp horizontal curvature of 492 ft (150 m) radius, and a vertical curvature of approximately 8860 ft (2700 m) radius, Figure 7.12. Maximum pier height is 212 ft (64.61 m). Site conditions and alignment precluded construction by the balanced cantilever method or conventional cast-in-place on falsework, leading to the decision to construct by the incremental launching method. The curved 1050 ft (320 m) length of this viaduct consists of 52.5 ft (16 m) long segments, which were fabricated in an enclosed shed behind an abutment. The bottom flange and bottom stubs of the webs of the first segments were cast first, Figure 7.13~1, 6, in a 52.5 ft (16 m) length, and approximately 118 ft (36 m) behind the first abutment. After curing and stressing of the partial segment it was jacked forward an increment of 52.5 ft (16 m) toward the abutment, where the balance of the section was cast, Figure 7.13~2, c. At the same time the formwork vacated by the first-segment bottom flange was reused for the casting of the bottom flange of the second segment, monolithically with the previous segment. After launching another 52.5 ft (16 m) increment the cycle was repeated until the superstructure was completed.4 Placement of the bottom flange mild steel reinforcement is shown in Figure 7.14, with the web forms in the background. The side forms for the webs and underside of the top flange cantilever, and the hydraulic jacking arrangement for stripping, are illustrated in Figure 7.15. Reinforcement in the top flange is shown in Figure 7.16 and the completed top flange with the following segment in the background in Figure 7.17. The completed segment with rails in place as it emerges from the casting shed is shown in Figure 1.6 1.

e=t

t

320.00m

Elevation

FIGURE 7.12. Plan (n) and longitudinal profile (6) of the Val Restel Viaduct, showing: .-\, shed for the construction of the deck segments; B, hydraulic equipment used for Iannching. From reference 4.

t

(b)

Cc)

FIGURE

7.13.

Construe

Cd)

tion stages Val Rested \&duct, from 1 eference 4.

FIGURE 7.14. Val Restel, placement of bottom flange reinforcement, from reference 4.

FIGURE 7.16. Val Restel, top flange reinforcement. from reference 4.

FIGURE 7.15. Val Restel, side form stripping mechanism, from reference 4.

FIGURE 7.17. Val Restel, completed top flange. with reinforcement for next segment in background, from reference 4.

328

Ravensbosch Valley Bridge, Holland

The superstructure cross section is shown in Figure 7.18~. Width of the segment is 29.5 ft (9.0 m). Total depth ofsegment is 8.13 ft (2.48 m), for a depth-to-span ratio of l/13. The top flange has a thickness of 9.8 in. (250 mm) and the bottom flange a thickness of 5.9 in. (150 mm). Figure 7.186 is a longitudinal section of the superstructure showing a layout of the second-stage prestressing tendons required after launching to accommodate loads on the final structure. Figures 7.19 and 7.20 show the interior anchorage blocks for the second-stage prestressing before and after concreting, respectively. A complete cycle of fabricating and launching a 52.5 ft (16 m) segment was accomplished in four nine-hour working days. Actual launching time for one segment was 60 to 65 minutes.4 Figures 7.21 and 7.22 show the launching nose approaching and landing on a pier. Views of the completed structure are shown in Figures 7.23 and T.24. Construction of this bridge was accomplished in ten months, from Januarv 1972 through October 1972.

329

of 6 ft (1.8 m) by 19 ft (5.8 m) with wall thickness of 1.3 ft (0.4 m), Figure 7.26. The superstructure consists of two siigle-cell trapezoidal box girders connected at the interior upper flange tips by a 8.3 ft (2.5 m) slab and prestressed transversely, Figures 7.26 and 7.27. Each box has a width of 56.8 ft (17.32 m) and a constant depth of 10.8 ft (3.3 m) for a depth-to-span ratio of l/17. The top flange has a thickness of 9.8 in. (250 mm) and the bottom flange a thickness of 7.9 in. (200 mm). Top flange cantilever is 13 ft (4.01 m). Each dual structure consists of 22 segments approximately 62 ft 4 in. (19 m) in length. The con-

7.4 Ravensbosch Valley Bridge, Holland

The 1378 ft (420 m) long Ravensbosch Valley Bridge near Valkenburg represents the first bridge in Holland built by the incremental launching method of’ segmental construction, Figure 7.25. ‘This dual structure has end spans of 137.8 ft (42 m) and six interior spans of 183.73 ft (56 m). Hollow rectangular piers vary in height from 21 ft (6.5 m) to 77 ft (23.5 m) and have exterior dimensions

FIGURE 7.19.

Val

Restel, second-stage prestressing

anchorage block before concreting, from reference 4.

FIGURE

7.20. Val Restel, second-stage prestressing

anchorage block after concreting, from reference 4.

:I tt40

32.00111

ICavi 4207mm-Cab/es

4207mm

t

32.00m

Ccrvo 16 17mm - Cob/e /617mm (b)

FIGURE 7.18.

Val Restel. (a) Cross section of deck. (b) Longitudinal section of deck.

From reference 4.

t


FIGURE 7.21. VA Kc\td, launching IIOW ing pier, from reference 4.

C~l~l~~

o,~h-

FIGURE 7.24. Val Kesrel, ~on~plctctl reference 4.

\‘iaclur.c, 1’1 o r n

FIGURE 7.22. VA Kestel, launching nose landing on pier, from reference 4.

Valley IS1itlge, general FIGURE 7.25. Ka\;ensbosch view (courtesy of Brice Bender, BVNiS’TS).

F I G U R E 7 . 2 3 . Vitl Kestcl, \,ic\c o f incrcnlcntally launched curved viaduct after launching, from reference 4.

struction of’ the superstructure was based upon a cycle of one segment per week. To accommodate bending moments during launching operations a 52.5 fi (16 m) long launching nose was used, Figure 7.28, in conjunction with a concentric first-stage prestressing consisting of’26 12 in. (32 mm) diameter Dywidag bars per box girder. In addition, temporary piers were used at midspan, Figure 7.28. During launching, friction amounted to 2 to 4%, equivalent to a maximum pushing force of 430 tons for a completed box girder.

Olifant’s

River Bridge, South Africa

331

FIGURE 7.26. Ravensbosch Valley Bridge, dual structure cross section (courtesy of

Brice Bender, BVNISTS).

After completion of the launching, second-stage prestressing following a parabolic profile and consisting of 12-0.62 in. (16 mm) diameter strands was installed and stressed. This structure was completed in 1975.

7.5 FIGURE 7.27. Ravensbosch Valley Bridge, girder

cross section (courtesy of Brice Bender, BVNISTS).

FIGURE 7.28. Ravensbosch Valley Bridge, view of

launching nose (courtesy of Brice Bender, BVNLSTS).

Olifant’s River Bridge, South Africa

This railroad structure, upon completion, held the world’s record for the longest bridge accomplished by incremental launching. It has a total length of 3395 ft (1035 m), consisting of 23 equal spans of 147.6 ft (45 m). The final structural arrangement consists of 11 continuous spans on each side fixed at the abutment and one simply supported center span-that is, an expansion joint on either side of the center span. With this structural arrangement the braking force of the trains (transporting iron ore) is transmitted to the abutments (10% of live load). In this manner the flexible piers can be used, resulting in an economy in the foundations by comparison with the classical solution, where the longitudinal force is transmitted through the piers to the foundations. All 23 spans were incrementally launched as 23 continuous spans from one abutment, Figure 7.29. During launching the two expansion joints were made temporarily continuous by temporary prestressing. The joints were released after the structure was in place and before it was rested on its permanent bearings. A launching nose, 59 ft (18 m) long, was prestressed to the first segment to maintain the cantilever stresses, during launching, in the concrete within allowable limits. The tip of


332 END BENT

PO

/

Pl

Construction of the superstructure \vas accomplished in nine months. Segments lvere span length, with the theoretical cycle per span of ten hours attained in the tenth operation and grx!ually reduced to seven hours at the conclusion of’ casting operations. Reinforcing cages Ivet-e pref’Ahricated in templates at the side of. the tornis. A qcle of operations consisted of the follo~ving: Clealling

and adjustment of forms

Placement of reinforcing loiver flange and \+.ebs

and

t e n d o n s for t h e

Concreting of this first phase

4

Placement of‘ upper flange

reinforcing

and

t e n d o n s t’or t h e

Concreting of’ this phase FIGURE 7.29. OMant’s launching awangenwnt.

Rive]- Bridge. incremental

Tensioning of’tendons in second phase of’ pre\

ious

span cast

the launching nose had a -jacking arrangement to accbmmodate deflection of’ the nose as it approached the pier. In cross section, Figure 7.30, the superstructure is a constant-depth rectangular single-cell box girder. Depth is 12.5 f‘t (3.80 tn); the top flange is 18 f‘t (5.50 m) wide and the bottom flange 10 f’t (3.10 m) wide. The webs and flanges are of a constant thickness throughout the structure. Web thickness is 13.75 in. (0.35 m) and contains vertical bar prestressing tendons to carry shear. Longitudinal prestressing is straight and contained in the flanges. Anchorage blocks for the longitudinal tendons at-e continuous across the width of both flanges (interior buttresses) to assure a more favorable distribution throughout the section. There are no diaphragms at the piers; the interior corner fillets are such as to permit the ef’f’ect of’ torsion to be accommodated by a transverse box frame.

r $ / ......... ‘:. “., -JL ____41 5.50

-I

Safety platform

.:

s l-4

Lm I”” L 3.10

FIGURE 7.30. Olifant’s River

I 4

Bridge, cross section

-1‘ensioning of’ tendons in first phase of span in forms Stripping of forttis Launching Af‘ter launching, and before placing the structure its final bearings, it \vas necessary to adjust the joints lvithin 2 itt. (10 mm). l‘he principal &f.ticuties in accomplishing this operation lvere: on

‘l‘emperature differential between night and da!., which produced a \ariation in length of’ the superstructure of 9.X in. (250 mm) Age of‘ concrete at time of‘ adjustment, lvhich ied t‘rom nine months to ten hours

V;II‘-

J a c k i n g op e1‘; 1t ’IOIIS, which could not retract the structure in case of an error in pushing forward .I‘he solution of the temperature problem ~~1s to quickly accomplish the adjustment early in the morning. Because of’ the constant temperature during the night the temperature of the superstructure was known, and its length was determinable in- spite of the thermal inertia of the concrete. The superstructure w a s then j a c k e d into its theoretical position on the abutment and firnil\, maintained by a system of blockage. The temporary tendons that had fixed the first joint were released and jacks were placed into the joint to push the remaining 12 spans and place the central simple span in its exact position. The second joint was then opened, and jacks at the other abutment po-

333

Various Bridges in France

sitioned the last 1 l-span portion of the superstructure. M’hcn rhc superstructure had thus been placed in position, it was -jacked up off the piers, and the temporar\ sliding bearings were replaced bv rhe petmatiet~l bearings.

7.6 Various Bridges in France 7.6.1 I.1.C

I~I.-lDl’C7

‘I‘his is a dual structure 912 f’t (278 in) long on a curve of a 3280 ft (1000 m) radius. The superstructure \vas constructed by incretnental launching of’ complete spans on sliding bearings. Resistance of rite structure to its dead load during b!- a temporal-! l a u n c h i n g \vas ~iccotntiiod~tretl cable-stay s! h,tetn in which the tension \vas adjusted proceeded, Figure 7.3 1. No a s c.otistt~uctioti supplementary prestressing \vas provided during the taunching phases. A 26 ft (8 m) launching nose leas pro\Gied at the leading end in order to reduce the Jveight of’ the cantilevered structure. It is a continuous structure supported on neoprene bearings and has a double-T cross section, as indicated in Figure 7.32. Roadway width is 46 ft (14.0 m), and depth of superstructure is a constant 10.3 f’t (3.15 in). Spans at-e 133.5 ft (40.i in).

‘l‘his structure consists of eight continuous s p a n s having a total length of 1102 ft (336 m), crossing a railroad and the Oise Ri\-er. The project is of interest in that it \vas launched from both abuttnents \vit bout the use of a launching nose or a tetnporark cable-sta\svstem. However, tetnporarv bents were used to control the cantilever stresses. In cross section the superstructure is a single-cell bos, Figure 7 .3 3 S e g m e n t s f’or e a c h o f t h e t w o h a l f - s u p e r structures were from 65.6 to 98.4 ft (20 to 30 m) in length. .A launching \vas effected upon completion of’ each segment. After the two halfsuperstructures had been launched to their final position, a closure pour of 3.3 ft (1 m) in length was c.onsutiitiiated to provide continuit\..

Concentric tendons frotn one end to the other of each half-superstructure, coupled together at each phase of concreting of segments Straight, short tendons in the top flange over the piers and in the bottom flange, centered in the span and tensioned after launching Continuity tendons, tensioned af’ter launching, situated in webs and anchoring at the upper flange Short parabolic tendons, located in the webs and a n c h o r i n g i n t h e t o p flange, t e n s i o n e d a f t e r launching ‘retnporary tendons in the upper flange, having the satne effect as the cantilever rendons

i.h.3

0I.I

L’I‘-tDI’CT

I‘his viaduct spans the valley of Oli in 15 spans of 134.5 ft (41 m) for a rotal length of 2017 ft (615 m) at a height of 197 ft (60 tn). The structure has a grade of 5.355% and a horizontal curve lvith a radius of 6700 ft (2046 tn). Total weight of the superstructure is 16,500 tons (15,000 mt). Incremental launching in this structure, rather than pushing the superstructure out over the piers, was accomplished bv a restrained lowering down the grade. The fo;-ce required in braking the structure was approximately 660 tons (600 tnt) as compared to the estimated force of 1540 tons (1400 tnt) to push the structure uphill. In its final configuration, because it was difficult to accommodate horizontal forces due to braking and seismic effects in the tall flexible piers, the superstructure is anchored in the terrain in the area of the abuttnents by a tie of a large stiffness. All of this longitudinal global force is accomtnodared in the large stiff tie, the abutments, and the relativelv short stiff piers in each bank. A central joint diiides the structure into two independent srructures. Upon cotnpletion of launching and before placing the superstructure on its pertnanent bearings, it was necessary to “unlock” the joint that held the two half-superstructures together during construction and to adjust its position within approximatelv i in. (10 mm). This operation was conducted as follows:

Longitudinal prestress consists of six sets: Cantilever tendons, tensioned bef-ore launching, located in the top flange and anchored in fillets at the intersection with the web

The superstructure was restrained at the upper abuttnent until the distance between its theoretical position and the end of the lower abutment was approximately 8 in. (200 mm).

334


.6 .

r.

F+,,.

1n

c;:--..

placing of the launching nose concreting and prestressing of the first span launching of the first span

concreting and prestressing of the second s?an erecting the cable-stay system launching of the first two spans

concreting and prestressing of the third span launching of the first three spans

concreting and prestressing of the fourth span launching of the first four spans

thing operations disassembling of the launching nose and cab1 e-s Lay systm

placing on permenant bearings placing and tensioning of phase 2 prestressing FIGURE 7.31. Luc \‘iaduct. incremental launching phases. ((I) Placing of the launching IIOSC. concreting atd prestressing of the fit-st span. launching of the first span. (b) (:OIIcreting and prestressing of the second span. erecting of the cable-sta! s\ stem, launching of the first t\vo spans. (c) Concreting and PI estressing of the third span. launching of‘ the first three spans. (rl) (:oncreting and Prestressing of the fourth span, launching of‘thc firat four spans. (P) Completion of launching operations. disassembling of’ the launching 11osc and cable-stay system, placing on permanent bearings, placing md tensioning of’ phase-two prestressing. .The t e m p o r a r y t e n d o n s c o n n e c t i n g t h e t w o h a l f s u p e r s t r u c t u r e s w e r e successivelv detensioned. However, two temporary tendons’ restrained the l o w e r h a l f - s u p e r s t r u c t u r e . T h e u p p e r half-superstructure was fixed to the upper abutment by a

s y s t e m o f ‘ p r e s t r e s s b a r s a n d c.otn~,lementar\ reinforcement installed in the upper abutment. .l‘he t w o temporat-!- t e n d o n s r e s t r a i n i n g t h e lowet half-superstt-uctttre lvere d e t e n s i o n e d i n increments, allowing the lolver half-super-strttcrure to

Wabash River Bridge, U.S.A.

t-

14.65 I

1

FIGURE 7.32. Luc Viaduct, cross section.

FIGURE 7.33. Creil \‘iaduct.

cross section.

descend to a blocking system in the lower abutment. Fixing of the lower half-superstructure to the lower abutment was then accomplished. The superstructure was positioned on its final bearings.

7.7 Wabash River Bridge, U.S.A. ‘I-his structure, the first incrementally launched segmental bridge constructed in the United States, carries two lanes of U.S. 136 over the Wabash River near Covington, Indiana. It is a six-span structure with end spans of 93 ft 6 in. (28.5 m) and four interior spans of 18i ft (57 m), Figure 7.34. Roadway width is 44 ft (13.4 m). Pier heights are approximately 40 ft (12 m); average river depth is 11 ft (3.35 m) with low water at 8 ft (2.4 m) and high water at 24 ft (7.3 m). The superstructure is a t\\‘o-cell box girder with a constant depth of 8 ft (2.4 m). .I‘he prqject was awarded in September of 1976 lvith a completion date of October 1978. The entire superstructure was completed in November of 1977. Original design plans prepared by American Consulting Engineers, Inc., of Indianapolis for the State Highway Commission called for a precast segmental balanced cantilever design; however, the bid documents permitted alternative methods of constructing the superstructure. The successful contractor, a .joint venture of Weddle Bros. Con-

335

struction Co., Inc., and the Ralph Rodgers Construction Co., both of Bloomington, Indiana, investigated three alternatives for the superstructure construction. These alternates included cast-inplace segments supported on falsework, incremental launching, and the cast-in-place segmental balanced cantilever method. Incremental launching was the successful method and reportedl) saved $100,000 over the other precast segmental method:j The V.S.L. Corporation of Las Gatos, California, was the subcontractor for prestressing and launching. A 140 ft (42.7 m) casting bed was located behind the west abutment of the bridge and could acconmodate three 46 ft 9 in. (14.25 m) segments. .I‘he forms for casting were supported on I beams, which were supported on steel piling to provide a solid foundation and prevent any settlement of the casting bed, Figure 7.35. The bottom third ot. the two-cell box superstructure was cast at the most westerly end of the casting bed, Figure 7.35. It was then advanced 46 ft 9 in. (14.25 In), where forms for the balance of the section were positioned, mild tendons steel reinforcement and prestressing placed, and the balance of the segment cast, Figure 7.36. After the segment had been poured and cured, the 20-ton jacks that held the forms in position, Figure 7.37, were released to break the bond and remove the forms. ‘I-he large metal forms stayed in place and were simplv swung in and out as needed. The segment was then advanced to the forward third of the casting bed for surface finishing by a conventional Bidwell screed, Figure 7.38, before launching over the abutment. In this methodolog! W;lS m a n n e r a production-line maintained. Three segments were always in various stages of fabrication, with reinforc&,lent and prestressing tendons continuous between segments. The first-stage pour required approsimately 53 vd3 (40.5 m”) and the second pour required from iO1 to 130 yd3 (77.2 to 99.4 111”). It took approximately four hours for each pour. ‘I‘jventy-eight-da) design strength \\‘as 4800 psi (3.37 kg/mm”), and 6000 to 7000 psi concrete strengths were actualI> attained (4.2 to 4.9 kg/mm’). A 3500 psi (2.46 kg/ mm2) strength was required before stressing, and this was normally achieved in 24 to 30 hours. As segments were completed, each was stressed to its predecessor by first-stage prestressing consisting of eight tendons of twelve f in. (12.7 mm) diameter 27 ksi (190 kg/mm2) strands, Figure 7.39. Initially the contractor was able to complete one cycle of segment fabrication and launching in two weeks:

A T

MIDSPAN

AT PIERS

PLAN D I R E C T I O N O F MOVEMW Fabrication area

CON>mJON -.

w

ELEVA.

FIGURE 7.34. Wabash River Bridge: cross section of girder, from reference 6; construction details, from reference 2.

however, as experience was gained, two cycles per week were attained. To accommodate the launching stresses a 56 ft (17 m) launching nose was attached to the lead segment, Figures 7.34 and 7.40. In addition, the four interior spans had temporary steel bents at midspan, Figures 7.34 and 7.41. In this manner the total structure length was divided into ten

equal spans of 93 ft 6 in. (28.5 m) during the launching procedure. Because of the longitudinal force on the piers during launching, the permanent piers were tied back to the abutment with four prestressing strands each. These strands were stressed to 96 kips (43,545 kg) before launching commenced. Each temporary pier was tied back to the preced-

FIGURE support.

7.35.

Wabash

River

Bridge,

casting-bed

FIGURE 7.39. Wabash River Bridge, first-stage prestressing.

FIGURE 7.37.

Wabash River Bridge, side form jacks.

FIGURE 7.38. top flange.

Wabash River Bridge, surface finishing

incrementally

FIGURE

7.40.

Wabash River Bridge, launching nose.

Launched

Bridges

horizontal jacks an 18 in. (457 mm) stroke. The vertical jacks lifted the superstructure about 4 in. (13 mm) and the horizontal jack pushed it forward 17 in. (432 mm). Each jacking cycle required about five minutes, and the entire launching of a 46 ft 9 in. (14.25 m) segment required about three hours. Temporary bearings, Figure 7.4, were located at each temporary bent and permanent pier. During the launching operation workmen were stationed at each bearing location to insert the teflon pads as the superstructure slid over the bearings. To maintain lateral alignment of the superstr&ture, lateral guide bearings, Figure 7.43, were also located at each temporary bearing and also used teflon pads. Workmen would tighten bolts on one side of the superstructure and loosen them on the opposite side to push the superstructure laterally. Final positioning of the superstructure on the east abutment was within & in. (0.8 mm) of its prescribed location.

7.8 Other Notable Structures 7.8.1 MiiHLBACHTALBRiiCKE,

GERMANY

ing permanent pier by two stays of 10 in. by 10 in. (254 mm by 254 mm) structural steel tubing, Figures 7.34 and 7.42. The jacking procedure during launching used the two-jack system (one vertical and one horizontal) and teflon pads, as described in Figure 7.2. The vertical jacks had a 2 in. (50 mm) stroke and the

Another example of this type of construction is the Miihlbachtalbriicke about 30 miles (50 km) southwest of Stuttgart, West Germany, Figure 7.44. This structure has an overall length of 1903 ft (580 m) with 141 ft (43 m) spans. The far-side trapezoidal box girder is shown in Figure 7.44 completed from abutment to abutment; the near-side trapezoidal box girder has been launched from the left abutment and the launching nose has reached the first pier. A general view of the structure is presented in Figure 7.45.

FIGURE 7.42. Wabash River Bridge, structural steel tubing tie.

F I G U R E 7 . 4 3 . Walmsh bearing.

FIGURE 7.41. W,lb,1sh Rile1

BI Age, temporary steel

bent.

Ki\;er Br-idge,

lareral g u i d e

FIGURE 7.44. Miihlbachtalbrticke. aerial view

7.47. Miihlbachtalbriicke, stressing tendon anchorage.

FIGURE

first-stage

pre-

FIGURE 7.45. Miihlbachtalbriicke, general view. Some idea of the size of the box girder may be obtained from Figure 7.46, showing the interior of the formwork at the rear of the abutment. Firststage prestressing tendon anchorage at the top of the web may be seen in Figure 7.47. The anchorage block for the second-stage prestressing is located inside the completed box, Figure 7.48.

FIGURE 7.48. Miihlbachtalbriicke, second-stage prestressing anchorage block.

7.8.2 SHEPHERDS HOUSE BRIDGE. ENGLAND

FIGURE 7.46. Miihlbachtalbriicke, segment in stationary forms.

The Shepherds House Bridge is the first incrementally launched bridge constructed in England. This highway structure crosses four railroad tracks at Sonning Cutting, near Reading, about 30 miles (48 km) west of London. The new structure contrasts sharply with an existing brick arch structure built in 1838 by Brunel, a famous English engineer. The existing structure consists of three circular brick arches supported on tall brick piers with the abutments founded in the sides of the cutting.’ A general plan showing the existing bridge, railroad tracks, and alignment of the new structure is presented in Figure 7.49.s

340


FIGURE 7.49. Shepherds House Bridge, general plan, from reference 8 (courtesy of Institution of Civil Engineers).

In 1971 the north abutment settled and the existing bridge was temporarily closed for repairs. In March of 1972, because the life expectancy of the existing structure was in question and because it did not comply with current highway standards, the Ministry of Transport instructed consulting engineers, Bullen and Partners, to prepare a study to determine the type and method of construction for a new structure. The new bridge provides a dualing of the existing road, and in the future the existing bridge will be replaced by a parallel structure. Because British Rail was engaged in extensive maintenance and upgrading of the tracks prior to introduction of high-speed trains, there would be severe limitations on track possession. Further, it was dictated that piers between tracks were to be avoided and that f-oundations on the north slope of the cutting were not to disturb the foundations of the existing bridge abutment. Construction working area was restricted because traffic was to be maintained on a residential street at one end and a trunk road at the other end. Soil conditions required that any temporary conditions that would load or disturb the slopes was to be avoided, thus requiring pile foundations with the pile caps at the surface to avoid extensive excavation in the slopes.s The consultants initially studied five possible schemes for construction of a bridge. Schemes using cast-in-place construction on falsework had earlier been rejected. An incremental launching scheme was recommended, even though there were no accurate cost

data for construction in the U.K. The consultants concluded that this scheme, although of shorter length than customary for this type of construction, would solve the problems of restricted working space and interference with residential streets and would require the least track downtime. The west elevation of the bridge is shown in Figure 7.50. Span lengths, determined by track location, are 75.5 ft (23 In), 121.4 ft (37 m), and X2 ft (25 III). The bridge is fixed at the south abutment with an expansion joint at the north abutment. ‘The casting bed for the production of 31.5 ft (9.6 rn) segments was located to the rear of the south abutment. The south abutment was located to provide maximum work space for the casting bed and to clear a large number of Post Office communication cables. Interior piers b and c were designed to withstand the friction forces exerted during launching operations. In addition, pier c, located close to the railroad tracks, was subject to damage or complete demolishment in the event of a derailment. Therefore, the superstructure was designed to withstand the removal of pier c by an accident. Six untensioned but anchored Macallov tendons in certain segments were added so as to preclude ultimate collapse with no live load on the bridge and pier c removed.7*H Normally, in this type of construction, the casting bed is of sufficient length to accommodate at least two and sometimes three segment lengths, such that the bottom flange may be cast separately in advance of the webs and top flange. In this project, with restricted space for the casting bed, it was decided to cast one complete segment in one pour.

341


- a ohllk rmll

‘we,. h n,ns 00. Dbnwula h mWumb*.

FIGURE 7.50.

Shepherds House Bridge, west elevation, from reference 7 (courtesy of The Concrete Society, London).

A maximum of three weeks was allowed for construction and launching of a segment. This time was later reduced to two weeks except for those segments with a diaphragm.’ A typical cross section of the box girder segment is shown in Figure 7.51. The launching sequence is shown in Figure 7.52. The steel launching truss nose was first erected using a temporary intermediate support. The first segment was cast against the launching nose and post-tensioned by Macalloy bars, some of which were used to connect the launching nose to the first segment. The launching nose, in position, before the launching of the first segment is shown in Figure 7.53. After the first segment had been launched forward, the next segment was cast and post-tensioned to the previous one. This procedure was repeated until the completed bridge was

FIGURE

launched to the north abutment. The launching nose passing over pier c is shown in Figure 7.54. Arrival of the launching nose at pier b is shown in Figure 7.55. The launching nose was removed after the concrete superstructure arrived at pier b, Figure 7.56. The superstructure was launched over temporary bearings, which consisted of high-grade concrete pads with a +Z in. (1 mm) thick stainless steel plate clamped and tensioned across the top surface. Lateral guide bearings were also provided to keep the superstructure on line. Upon completion of launching the superstructure was jacked in a predetermined s e q u e n c e a n d t h e t e m p o r a r y bearings were replaced with permanent bearings8 The jacking force for launching was provided by two jacks pulling on a set of nine 0.6 in. (15 mm)

7.51. Shepherds House Bridge, girder cross section, from refer. ence 8 (courtesy of The Institution of Civil Engineers).


342

Stage 1: Cast first unit and connect to launching

nose

Stage 2: Launch to pier C Stages 3-5: Launch over tracks

FIGURE 7.54. Shepherds House Bridge, launching nose passing over pier c, from reference 7 (courtesy of The Concrete Societv. London). Stage 6: Launch to per B Stage 7: Conttnue launch

Stage 8: Reach pw 9 and remove “018 Stages 9 and 10: Complete launch

FIGURE 7.52. Shepherds House Bridge, sequence of incremental launching, from reference 8 (courtesy of The Institution of Civil Engineers).

FIGURE 7.55. Shepherds House Bridge, launching nose at pier b, from reference 7 (courtesy of The Concrete Society, London).

FIGURE 7.53. Shepherds House Bridge, launching nose in position before launching, from reference 7 (courtesy of The Concrete Society, London). diameter cables passing under the casting bed and

anchored to the front of the abutment. The load was applied to a fabricated bracket secured to the rear of the segment by bolts coupling with the projecting ends of the Macalloy bar tendons in the top and bottom flanges of the segment, Figure 7.57. The two jacks were operated in tandem by a single pump. This system required 30 seconds for jacking and 30 seconds for retracting for each 10 in. (254 mm) str0ke.s

FIGURE 7.56. Shepherds House Bridge, superstructure launched to pier b and launching nose removed, from reference 7 (courtesy of The Concrete Society, London).

Design

of

Incrementally

Launched

Bridges

The dimensions for sented in Section 4.5.4 thickness, but the top thickness may have to the type of prestressing 7.9.4). 7.9.3

FIGURE 7.57. Shqhtwls 1 lowc~ Bridge, segment being launched from f’ormwork, from reference 7 (courtesy of’ The Concrete Society, London).

7.9

Design of Incrementally Launched Bridges

7.9.1 BRIDGE ALIGNMENT REQUIREMENTS

The designer must always remember that in order to construct incrementally launched bridges, the horizontal and vertical alignment must be either straight or constantly curved or twisted. This is generally not the case, as road planners are not bridge builders. As a matter of fact, it is the soffit ofthe bridge deck that has to be designed with a constant radius of curvature; the transverse cantilever of the deck flange can be varied to accommodate possible small deviations. 7.9.2

343 typical cross sections preremain valid for the web flange and bottom flange be increased, depending on layout adopted (see Section

SPAN ARRANGEME,VT AND RELATED PRI,VCIPLE OF CONSTRUCTION

The constant-depth requirement limits the economical use of this construction method to spans not longer than 160 to 200 ft (50 to 60 m). It is advantageous if all the spans are equal in length. However, much longer spans have been built by utilizing special techniques in conjunction with the basic principle of incremental launching. A three-span construction may be launched from both sides. In this way the center span can be twice the length of the edge spans without increase of the stresses in the deck. The span configuration then becomes: L-2L-L (see Figure 7.58). Champigny Bridge near Paris was the first structure of this type. Longer bridges are often launched from one side only (the record length is that of Olifant’s River Bridge in South Africa, in excess of 3300 ft). Auxiliary temporary devices are used to reduce the bending moments in the front portion of the deck (launching nose or tower stays)

TYPE, SHAPE AND DIMENSIONS OF SUPERSTRUCTURE

This method of construction requires a cross section with a constant depth, since the designer has to insure the resistance of the superstructure, under its own weight, at all sections as the launching proceeds. Economic considerations dictate a constant moment of inertia. Two types of cross section have been used to date: the box girder and the double T. The box girder provides a better stiffness and resistance to torsion and at the same time an easier placement of the prestressing tendons in the cross section. The depth of the box is usually one-twelfth to onesixteenth of the longest span, the first value applying to larger and the second to smaller spans. Table 7.1 summarizes the characteristics of several incrementally launched bridges.

FIGURE 7.58. Three-span symmetrical incrementally launched bridge.

TABLE 7.1. Characteristics of Incrementally Launched Bridges 1‘) pical Name

Year

Cross Section

Span

(t’t) Nuel Viaduct, France

~I‘otal Length (W

Launched Weight (t)

Vertical Curve

Horizontal Curve

1976

135

807

6,000 Slope 6%

R = 2,460 ft

Borriglione Viaduct, France

1976

135

807

6,000 Slope 5.5%

R = 2,460 ft

Kimonkro Bridge, Ivory Coast

1978

118

709

3,600

Straight

Tet Viaduct, France

141

660

Luc Viaduct, France

135

915

7,900 Slope 3.8%

Straight

1976

135

1,151

Slope 1.3%

Curve

1976

135

2,018

1972

131

345

1978

194

1,102

Paillon Bridge, France

Oli Viaduct, France

Marolles Bridge, France

Creil Bridge, France

344

15,000 Slope 5.85% R =

6,712 ft

345

Design of Incrementally Launched Bridges TABLE 7.1. (Continued) l‘otal .\ ‘II11C

1978

Var Viaduct, France

Bridge, Kufstein, German\

262

1,732

1976

138

1,107

1965

335

1,476

169

1,562

138

1,398

1967

108

469

1978

6 spans

t-

1

13,000 Slope 0.7%

R = 7,217 ft

9,700

Straight

\

Querlin Guen Bridge. German\

Ingolstadt Bridge, Danube Bridge, Gertnant

Horizontal Curve

\ L17.5’

Koches Valle) Bridge, Gerlnan\

Abeou Aqueduct, France

Launched Vertical Weight (t) Curve

46.3’

Gronachtal Bridge, Gerlnan!

Inn

Length (it)

l’eal-

197 to 377

as previously indicated in some of the examples described in this chapter. When the spans become too large, intermediate temporary bents are used. This was done for the first bridge over the Caroni River in Venezuela. The record span length for incrementally launched bridges was obtained by a structure over the Danube River designed by Prof. Leonhardt, the originator of the method, Figure 7.59. The cost of the temporary bents depends greatly ox the

2x 1,246

foundation conditions; it may be prohibitive if the bent height is greater than 100 ft (30 m) and soil conditions require deep piling. For very long bridges, intermediate expansion joints are needed, much the same as for cantilever bridges. The expansion joints are temporarily fixed by prestressing during launching and are released at the end of construction to allow for thermal expansion in the structure during service. A very ingenious variation of this principle was de-

346


FIGURE 7.59. Ih~ltrlx

Ki\-et.

Bridge,

.-\ustria.

veloped for the Basra Bridge in Iraq, where a concrete swing span was launched together with the approach spans as a single unit and later arranged to serve its purpose as a movable bridge over the navigation channel, Figure 7.60. 7.9.4

DESIGN OF LOlVGITUDINAL MEMBERS FOR FLEXURE AND TENDON PROFILE

During launching, the superstructure is subjected to continually alternating bending moments, so that any one section is subjected to a continual variation of bending moments, both positive and

negative, as shown in Figures 7.61 and 7.62. These bending moments are balanced by internal uniform axial prestressing. In the final stage, additional tendons are required to supplement the uniform axial prestressing in order to carry the service loads. Conventional solutions are applied to this problem, and in the present discussion we need only enlarge upon the specific problem of the axial prestressing. For this prestressing, tendons are so arranged that the compressive stresses are the same over the entire cross-sectional area. The required tendons are placed in the top and bottom flanges of the box section. They are usually straight, tensioned before launching, so couplers are needed at each joint between successive segments. Segment length may vary from 50 ft (15 m) to 100 ft (30 m). As noted in our discussion of- the progressive construction method, there are limitations to the deck’s capacity to carry its own weight during launching when the front part is in cantilever beyond a typical pier. To keep bending moments and stresses within allowable values, it is usually necessary to use a launching nose, a light steel member placed in front of the concrete structure to allow support from the next pier, rather than launching the concrete deck all the way with no support. Numerical values are given in Figures 7.61 and 7.62 for the critical maximum positive and negative moments during launching. Assuming the unit weight of the launching nose to be 10% of the weight of the concrete deck (a value somewhat lower than average), the critical

FIGURE

7.61. Critical negative moments during launching with nose. M, (W’L2/12)[6a’ + 6y(l - &)I. Multiplier: WL2/12. For y = 0.10:

N

P

M”

0.20 0.30 0.40 0.50

0.80 0.70 0.60 0.50

0.82 1.09 1.46 1.95

1 .oo

0.00

6.00

k yo/rp,n

=

.&

7.62. Critical positive moment during launching with nose. M, (WL”/12)(0.933 - 2.96#*). Multiplier = WL’l12. For y = 0.10: a M, P FIGURE

0.20 0.30 0.40 0.50

0.80 0.50 0.60 0.50

0.74 0.79 0.83 0.86

1.00

0.00

0.93

=

347

3

4

8


moments are as follows for various lengths of the launching nose: Nose Length, Percent of Typical Span 50 60 70 80

Maximum Moments Support OKJ

Span (Ml)

M&f,

1.95 1.46 1.09 0.82

0.86 0.83 0.79 0.74

2.27 1.76 1.38 1.11

Moment factor is WL2/12 (W = weight of concrete per unit length and L = span length)

FIGURE 7.65. Sathorn

Technologically, the uniform axial prestress may be installed in the superstructure in several different ways:

above the concrete deck with steel deviation saddles at intermediate joints. The three solutions above have their relative merits and disadvantages:

1.

2.

Straight tendons running through the top and bottom flange of each segment, joined by couplers at the joints between segments. Straight tendons running through the top and bottom flanges, anchored in block-outs inside the box girder, Figure 7.63.

3. Temporary curved tendons may be used to balance the final continuity tendons during construction. These tendons are outside the concrete section between supports, Figure 7.64. This method has been used for several large projects. Figure 7.65 shows the Sathorn Bridge in Bangkok, Thailand, with the temporary tendons installed

FIGURE 7.63. Lapped prestressing tendons. TEMPORARY

I

PRESTRESSING

F I N A L PRES’RESSING

SUPPORT5

I

FIGURE 7.64. Temporary external prestressing system.

Hr idge, Thailand.

1. The first solution may require local thickening of the concrete flanges for placement of the couplers. However, it is often preferred to increase the thickness of the flanges over the entire bridge length to simplify casting of the segments. Axial prestressing tendons are permanent and cannot be removed. They must be incorporated in the final prestressing layout. The joints between segments have to be carefullv designed, owing to the presence of couplers and concrete voids that may significantly weaken the section. 2. The main advantage of the second solution pertains to the removal and reuse of those tendons not required in the final prestressing layout. However, the cost and difficulty of providing a large number of block-outs offsets a significant part of the advantage of removing the temporary tendons. In order to obtain a satisfactory shear resistance from the webs, particularly during launching with alternating shear and bending stresses, the configuration of the box section and location of the upper and lower blisters must be carefully considered. This problem was mentioned in Chapter 4 as presenting potential difficulties. A satisfactory solution is shown in Figure 7.66, where upper and lower blisters are not in the same vertical plane. A sufficient amount of vertical prestress will insure the resistance of webs against shear during all construction stages. 3. The third solution is theoretically a satisfactory one, allowing the permanent prestress to be installed during construction and the temporary prestress to be designed only to counteract the un-

349

Design of Incrementally Launched Bridges

reinforcement must be made in an area already densely prestressed.

FIGURE 7.66. Offset lapped prestressing tendons. desired effects of the former during ‘moment reversals created by the successive launching stages. In practice, installation of the tendons passing from the inside to the outside of the box section is not particularly simple. An attempt should be made to reuse these temporary tendons to reduce the investment in nonproductive materials. A comparative analysis between the first two methods of temporary prestressing has been made for a typical railway bridge. Solution 2 requires 19% more conventional reinforcement than solution 1 because of the many blisters and more elaborate tendon layout. The total cost of materials (concrete prestress and reinforcement) is 9% higher for solution 2 than for solution 1. These results may be significantly different for highway bridges, where the ratio between girder load and superimposed dead and live loads is very different. 7.9.5 CASTING AREA AND LAC’,VCHING METHODS

The precasting area is located behind one abutment and has a length usually equal to that of two or three segments. T h e r e a r e t w o d i f f e r e n t launching methods: 1. The launching force is transmitted from the jacks bearing against the abutment face to the bridge by pulling tendons or steel rods anchored in the bridge soffit. 2. A launching device consisting of horizontal and vertical jacks is placed over the abutment. The vertical jack rests on a sliding surface and has a special friction gripping element at the top. The vertical jack lifts the superstructure for launching, and the horizontal jack pushes it horizontallv. The designer should be concerned with the following items: The first launching method applies high local forces to the concrete soffit where the pulling device is anchored. Careful design of the passive

The second launching method requires sufficient vertical reaction on the vertical jack. This could be critical at the end of launching, when the required launching force reaches its maximum with a corresponding small vertical reaction. A very precise geometry control is required during launching. The possibility of foundation settlement must be considered in the design. Whichever launching method is used, after completion of the launching procedure the deck must be raised successively at each pier so that the permanent bearings may be installed. This phase also calls for careful analysis.

7.9.6 LAUNCHING NOSE AND TEMPORARY STAYS

The large cantilever moments occurring in the front part of the superstructure that is being launched from pier to pier inevitably call for special provisions to keep the bending stresses and the temporary prestress within allowable and economically acceptable limits. Two methods have been used together and separately, as previously mentioned: nose: A steel member made either of plate girders or of trusses is temporarily prestressed into the end diaphragm of the concrete bridge, which is the front section of the deck during launching. Tower and stays: This method was described in Chapter 6 for progressive construction. Its application to incremental launching, however, needs a special approach, because the relative position of the tower and the stays changes constantly with regard to the permanent piers.

Launching

The advantage of the launching nose to reduce cantilever moments in the concrete superstructure was discussed in Section 7.9.4. It is important not only to select the proper dimensions of the launching nose but also to take into proper account the actual flexibility of the steel nose in comparison to that of the concrete span. This relative flexibility may be characterized by the following dimensionless coefficient: ++ c c


350

where E, and E, refer to steel and concrete moduli, and I, and I, are the moments of inertia of the steel nose and concrete superstructure. Figure 7.67 presents the results of a study analyzing the variation of the maximum support moment in the concrete deck for different launching stages with the relative stiffness K. This chart confirms the obvious fact that a flexible nose has only a limited efficiency in reducing the moments in the concrete deck. The following table gives the characteristics of several structures using a launching nose and serves as a reference for preliminary investigations of the optimum launching method.

Bridge

Launching Nose Length [ft (Ml

Wabash River Oli River Saone Roche

56 59 93.5 124.5

(17) (18) (28.5) (38)

Weight of Launching Nose (tons)

Stays

30 36 65 90

No Yes No No

For longer spans the launching nose is not necessarily the optimum solution, while temporary bents may also be expensive. A tower-and-stay system has been successfully used either alone or in conjunction with a launching nose to reduce the cantilever moments in the front part of the superstructure.

FIGURE 7.67. Variation of the maximum support moment.

To allow the method to be effective in all launching stages, it is necessary to constantly control the reaction of the tower applied to the concrete deck. When the tower is above one pier, it is totally efficient. When launching has proceeded for another half-span length, the tower and stays produce additional positive moments at midspan, exactly contrary to the desired effect. For this reason the tower may be equipped with jacks between the concrete deck and the tower legs, and the tower reaction may be constantly adjusted to optimize the stresses in the concrete superstructure. Figure 7.68 shows a device being successfully used for the first time in the construction of the Boivre Viaduct, near Poitiers, France. 7.9.7 PIERS AND FOUNDATIONS The loads applied to the piers and foundations during the incremental launching procedure are very different from those appearing during service. The static configuration of the piers is also

FIGURE 7.68. Boivre Viaduct. nwr I’oiliers. France.

351

Design of Incrementally Launched Bridges different. During construction, the bridge slides over the pier tops and the buckling length of the pier is larger than that during service. The horizontal force applied to the pier top is also higher than during service, thus requiring a close study of this construction phase. Lou& <4cting on the Piers The various systems of horizontal forces that may act on the piers depend on the following: Longitudinal profile of the superstructure Direction of’ launching E‘riction coef‘ficient of sliding bearings Notation: H= angle ot bridge superstructure lvith respect to the horizontal; tan 0 = r 4= angle of’ f’riction of sliding bearings; tan C$ = p R = total reaction of the superstructure on the piel-: \,ertical and horizontal components V and H, normal and tangential components A: and 7 The f’ollowing four cases will be considered (see Figure 7.69): 1. H > 4, upulard launching: Sliding starts on the bearings \vhen the inclination of the reaction R \\,ith respect to the vertical is: cY=t)++,

FIGURE 7.69. Reactions on piers during launching. (a) upward launching. (b) downward launching. F=N(tan8-tan+)

H = (r + p)V 2. H > 4, downward launching: Sliding starts Lvhen cy = 8 - 4. The horizontal force on the pier acts in the direction opposite to that of movement irith a value: H = V tan (0 - 4) For small values of the angles: H = (I - p)V Because p varies with environmental conditions (cleanness of the plates in particular), the launching equipment and the pier will be designed for H = 4’. The downward movement of’ the bridge is controlled by a restraining jacking force:

r

F=N(r-p)

For the same reasons as above, the safe value of F is equal to Nr. 3. 0 < $, upward launching: As above, the horizontal load applied to the pier is:

H = V tan (0 + 4)

For small values of 0 and 4:

o

H = (r + p)V 4. 0 < 4, downward launching: In this case the horizontal load on the pier is applied in the direction of the movement with a value of‘: H = (r - p)V Because of the possible variation in the angle of friction, it is safer to provide a braking system to control the movement of the bridge. Pier Cap Detailing The pier caps must be carefully detailed in order to provide room for the following devices: Temporary

sliding

bearings

Vertical jacks to lift the bridge after launching to install the permanent bearings Horizontal guiding devices during launching

352


Adjusting jacks for correction of the relative displacements between piers and deck

and the principle of the method are shown in Figure 7.70. The 900-ton structure had a width of 26 ft and the following spans: 46, 55, 55, 46 ft. The existing reinforcing did not provide the necessary strength to resist superstructure dead load during launching. Therefore, a rear launching-out tail 26 ft long was installed at the end opposite the direction of launching, while exterior post-tensioning tendons were placed above the deck to strengthen the structure. The bridge was lifted off its bearings 7 in. to install sliding bearings and lateral guiding devices in preparation f’or the operation. The whole operation was performed in 54 weeks as f’ollo\vs:

Moreover, to reduce the pier bending moments induced by launching, the sliding bearings are often eccentric. However, it is possible to reduce or balance this horizontal force by installing ties anchored in the ground. If the piers are very high, the horizontal force can be eliminated by using jacking equipment directly installed on the piers.

7.10

DemoEition of a Structure by Incremental Launching

We close this chapter with an unusual application showing the interesting potential of incremental launching. An overpass structure over the A-l motorwav north of Paris needed to be demolished for replacement by another structure as part of a highway relocation program. I‘he limited headroonl between the existing bridge soffit and the clearance diagram, together with the considerable traffic on the major motorway providing permanent access from Paris to Charles de Gaulle Airport, made all conventional methods of demolition extremely difficult and unadapted. A ver! simple scheme was devised whereby the deck was launched away from the traffic onto the approach embankment to be conventionallv demolished at leisure. The dimensions of the bridge

Design and preparation of the contract Mobilization and purchase of equipment Launching

rraffic \vas interrupted for only f’our nights between 10 P.SI. and 6 .a.~. The operation turned out to be a complete success in spite of its originality.

L4UNWlt4G

I

46’

1

55’

1

55’

: T

I

46’

t

T O T A L WEICUT

soot

PROCEDURE I/ L I F T T O T A L

2)

SCHEDULE

BRIOSC 7’

PLACE ROLLERS OVER

PIERS

A N D 48lJlMCNl5 3) INSTALL

APPROACH FILL AND

CONCRETE

A)

PLACE

T O T A L 5’/2 W E E K S

_ DESIGN & CONTRACT :

2

.MOB.

2

PURCUAsLS :

_ LAUNCHING

:

IV!?

5k

BEAM5 TRAFFIC INTERRUPTION :

PROVISIONAL

A REAR NOW

26

2 2 1; 5;

P/l A N D FT. LONG

FIGURE 7.70. Bridge over A-1,

(IO PM.

launching out.

TO

6

A.M.)

4

N;OUTs

References

2. :~non..

“First Incrementally Launched Post-Tensioned Box Girder Bridge to Be Built in the United Swtes.” Bridge Report, December 1976, Post-Tensioning Institute, Phoenix, Ariz. 3. .ir\ icl (;I.d1it. “Increment;il Launching of Concrete Stn~cturcs.” Jorrmnl of thr .-lvwriccl~r Courwtp Imtltutr, \‘()I. T”. s o . 8.

4 .

.-\ug”st

19i5.

.4IlOII..

“\‘a1 Restel Viaduct tol- the Provincial Road S o . 89 .Se;rr Ko\ereto. .I‘rento,” Prr.clrr.c,wd Couovtr .Strrlttrrtfs I)/ I/n/y 1970/l 974, .-\ssociarioIle Itdli;~na (:c.IllcIlto

.-\llllatc,

.\ ssoc-id/ione

E:

It~rliamr (.\I IX(:). Rome 1971.

l’l~ec.oIllpl~esM~

~cononica

(AIC.-\I’)

and

Del Cement0

353

.-h~Il., “Segmental Box Girder Bridges \lake the Big ‘I‘ime in U.S..” Engiuuwiug .\‘~!\-RPcoM/. \Iarch 2. 1978. 6. Xnon., “Wabash Rive]- Bridge. Covingtot). I ndian;l.” P o r t l a n d C e m e n t .-\ssociation, Bridge Repot“, SR201 .Ol E, lYi8, Skokie. 111. i. Xl. .\Iaddison, “Crossing the Cutting with Segments at Sonning,” Coucwtp, 7%~ Jou,-r~u/ of tha Corlo.rtr Socirt! (Lot~rlor~j, Yol. 12. S o . 2, Februar\ 19iH. 8. K. H. Best, R. H. Kingston, and 11. J. \Vhatle\, “lncremental Launching at Shepherd House Bridge,” Pwcfwfirrgc, Instztution of Cnfil Eugrnrfm, \‘ol. 64. Part I, Fehruar\ 1978. 5 .

8 Concrete Segmental Arches, Rigid Frames, and Truss Bridges

8.1 8.2

INTRODUCTION SEGMENTAL PRECAST BRIDGES OVER THE

MARNE RIVER, FRANCE 8.3 CARACAS VIADUCTS, VENEZUELA 8.4 GLADESVILLE BRIDGE, AUSTRALIA 8.5 A R C H E S B U I L T I N C A N T I L E V E R

8.6 RIGID-FRAME BRIDGES Bridge in Toulouse, France Saint Michel 8.6.1 8 . 6 . 2 B r i e s l e Maas B r i d g e , N e t h e r l a n d s 8.6.3 Bonhomme Bridge, France 8.6.4 Motorway Overpasses in the Middle East 8.7 TRUSS BRIDGES 8.7.1

8.5.1

Review of Concept; Summary of Structures with Temporary Stays

8.5.2 Neckarburg Bridge, Germany 8.5.3 Niesenbach Bridge, Austria 8.5.4 Kirk Bridges, Yugoslavia

8.1 Introduction An arch bridge, in a proper setting, is an elegant and graref‘ul structure with aesthetic appeal. Instinctivelv, a layman relates to an arch bridge as a form that follows its function. Long before prestressed concrete was developed as a technology, concrete arches were used for long spans, taking advantage ot the compressive stress induced b\ gravitational- forces into a curved tnetnbet- much as earlier generations of builders had done with niasotirv arches. Three bridges designed and built by Eugene Freyssinet between 1907 and 1910 in central France were to become a tnajor landmark in the development of concrete structures. In the \‘eut-dre Bridge, Figure 8.1, the three hinged reinforced concrete arches had a clear span of 238 ft (72.50 m) and an unusual rise-to-span ratio of l/15 dictated by the topography of the site and the sudden floods of the Allier River. The \‘enture Fvas an unqualified success both during load testing and after opening to traffic. As Freyssinet wrote in his memoirs: 354

Retrospect on Concepts for Concrete Truss Bridges 8 . 7 . 2 Mangfall B r i d g e , A u s t r i a 8.7.3 Rip Bridge, Australia 8.7.4 C o n c e p t f o r a C r o s s i n g o f t h e E n g l i s h C h a n n e l REFERENCES

Lord testing um (I triumph. 0~ the Gght bank, (1 hill oz~erlookirrg the bricige site UYS occupied by .\e-c~ewl thou,wd spectators do had trrken thei?- plclce c~lretrd~ crt dnulrl to ulatch the j
355

Introduction

FIGURE 8.1. Veurdre

first consequence seen in a structure of a phenomenon theretofore completely ignored: long-term concrete creep. Other beautiful concrete arches were also constructed in the same period. The Villeneuve Bridge over the Lot River in southwestern France, Figure 8.2, is an interesting example. The twin arch ribs are of plain concrete with a clear span of 316 ft (96 m) and a rise of 47 ft 4 in. (14.5 m). Each rib has a solid section 10 ft (3 m) wide and 4 ft 9 in. (1.45 m) deep built in at both ends into the concrete abutments. The reinforced concrete deck rests upon the arch ribs through a series of thin spandrel columns, faced with red brick. Construction began shortly before World War I and was interrupted for four years, fortunately not before the concrete arch ribs could be cast on a wooden falsework, Figure 8.3. Immediately upon completion, hydraulic rams were used at the midspan section to lift the concrete arches off the falsework and actively create the compressive stress in them, a technique from Freyssinet’s fertile mind that already contained the germ of the idea of prestressing.

FIGURE 8.2. \~illcncu\c HI itigc O\~I chc Lot Ki\cr.

Bridge.

The bridge was completed in 1919 and kept the world’s record for long-span concrete structures for several years. The photograph appearing in Figure 8.2 was taken by one of the authors in the summer of 1980; it shows that beautiful structure in a remarkable state after sixtvI vears of continuous operation under constant urban traffic. Another Freyssinet design, the Tonneins Bridge over the Garonne River, was built at the same time, and he considered it to be one of his nicest bridge structures, Figure 8.4. The Plougastel Bridge in Brittany, Figure 1.38, reached for longer spans with concrete arches. For the first time a box section was employed, calling on an ingenious method of construction in which a wooden falsework was floated into position and reused several times for the various arch ribs. Dimensions of the structure and typical details of the arches are shown in Figure 8.5, which is a facsimile of a document published in 1930. The three arches have a span length of 611 ft (186.40 m) and carry a single-track railroad and a two-lane highway. The reinforced concrete trussed double deck accommodates the train track on its lower level and the highway on the upper. Near the arch crown in each span, the train passes through the arch rib. The arch ribs were only slightly reinforced and the quantity of steel was 39 lb/y&’ (23 kg/m’{), in spite of the relatively thin walls used for the box section. The three arch ribs were constructed one after the other on a temporary wooden arch built on shore and floated into position for each of the three concrete arches, Figures 8.6 ,and 8.7. This wooden arch was 490 ft (150 m) long and weighed 550 tons (500 mt), including the two reinforced concrete end sections, which allowed the thrust created by the concrete arch ribs to be transferred

.

--

~. .,..

,I,.

-r

Segmental Precast Bridges Over the Marne River, France

357

know it today. It incorporated so many innovations in a single structure that it would not be out of place in today’s modern bridge technology. The single-span structure, Figure 8.8, is a double-hinged arch with a distance between hinges of 180 ft (55 m) and a very tight clearance diagram for river navigation that allowed only 4 ft 3 in. (1.30 m) below the finished grade of the roadway. Consequently, not only is the bridge structure very shallow, 4.16 ft (1.27 m), at midspan, but the riseto-span ratio of the arch is unusual: l/23. The bridge consists of three parallel box sections made up of precast segments 8 ft (2.44 m) long, connected after placement in the structure by precast slab sections at both top and bottom flanges, Figure 8.9. The bridge is prestressed in three directions:

to the arch springings completed earlier on the foundation caissons. Two barges and a temporary steel tie slightly above the water level, with the help of the large tidal range, allowed the transfer of this falsework from the construction area to the three positions of use and its final return after completion of the concrete structure. As this outstanding undertaking neared completion in 1930 after five years of uninterrupted effort, Freyssinet expressed his thoughts as follows: In Brittany light is like a fairy who constantly plays at covering nature with [many] changing coats, now of lead, noul of silver or of pearls, or of something immaterial and radmnt. Toward the evening oj the load testing of the bridge, she had spread her most sumptuous treasures on the roadstead and each line of the work, changed into a long rosary of unreal light, added another touch of beauty to the marvellous whole, proving in this way that the Fairy of the Roadstead had already adopted the child that men had imposed on her and had known how to weave for him garments magn$cent enough to hide all the imperfections of the work.

8.2

Segmental Precast Bridges over the Marne River, France

Located some 30 miles (50 km) east of Paris, the Luzancy Bridge represents probably the first application of truly segmental construction as we

The 4 in. (0.10 m) webs are vertically prestressed to resist shear. The longitudinal box girders are then prestressed to connect the precast segments and resist bending. The negative-moment prestressing tendons at the top flange level over the arch springings are located in grooves provided at the top surface of the precast segment upper flange and are ultimately embedded in a 2 in. (50 mm) concrete topping. This dense, high-quality concrete pavement provides the sole protection for the high-tensile steel wires and also serves as the sole roadway wearing course. In spite of the excellent behavior of this structure after more than 34 years of operation, it would probably be difficult to envisage duplicating it today. Transverse connection between the box girders and the connecting slabs is achieved by prestressing. There was no conventional reinforcing steel in the bridge superstructure except in local areas, such as the Freyssinet concrete hinges at the arch springings. The erection was just as remarkable as the conception of the bridge. Each box girder consisted of 22 segments, which were cast in a central yard at the rate of one a day (little progress has been achieved after thirty years). Afterward they were carefully aligned on concrete blocks to take the profile of the finished structure with proper provision for camber. The 2 in. (20 mm) wide joints were dry packed to allow segment assembly by prestressing. In fact, the 22 segments of each box girder were assembled at this stage in three units: two side units made up of three segments each, and the center unit incorporating the remaining 16

L E GENIE CIVIL

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FONT A.LOUPPE,EN BQTONARMG, SUR L'ELORN,PRlj:S DEPLOUGASTEL (FINIST~RE)

re du tabher

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Fig 8 Coupe’de kc montrant la dqositmn des armatures

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FIGURE 8.5. Plougastel Bridge, dimensions of the structure and details of the arches, a facsimile of a document published in 1930.

~,.swn.CU*o”~rUn*ar *.ms

Segmental Precast Bridges Over the Mame River, France

segments with a length of 170 ft (52 m) and a maximum weight of 134 tons (122 mt). All three units were assembled on the bridge centerline immediately behind one abutment, while the deltashaped sections representing the arch springings were cast in place over the abutment in their final location in the structure. A special aerial cableway made up of two steel towers resting on both banks and properly anchored to the rear, a system of suspended winches, and a unique elliptical drum allowed the transfer of the precast girder units from their assembly po-

Pou 1 re

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FIGURE 8.8. Luzancy

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cli

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FIGURE 8.9. Luzancy Bridge, concrete dimensions.

Blitlgc.

359

360

Concrete Segmental Arches, Rigid Frames, and Truss Bridges

sition on the banks to their final location in the structure. In spite of a seemingly involved concept, the operations were carried out safely and rapidly; a center beam was placed in only eight hours and a complete arch including all preparatory and finishing operations was assembled in 120 hours, Figure 8.10. Another interesting feature of this structure was the incorporation at both arch springings of Freyssinet flat jacks and reinforced concrete wedges between the arch inclined legs and the abutment sills, to adjust and control the arch thrust and the bending moments at midspan. The bridge was opened to traffic in May 1946 after successfully proving its structural adequacy through a comprehensive series of static and dynamic load tests, following a custom still in use today in several European countries. Figure 8.11 gives a view of the finished structure. This first precast segmental arch bridge was followed a few years later by a series of five other structures, all of the same type and in the same geographical area, the valley of the Marne River,

FIGURE 8.12. One of the five hlarne River Bridges:

Esbly, Anet, Char@, ‘Trilbardou, and Ussy. Figure 8.12, at the following locations: Esbly, Anet, Changis, Trilbardou, and Ussy. All five bridges have the geometric dimensions shown in Figure 8.13: Distance between hinges: 243 ft (74 m) Rise of the central axis at the crown over the abutment hinge: 16.3 ft (4.96 m) Depth at crown: 2.82 ft (0.86 m) Deck width: 27.5 ft (8.40 m) The deck structure is made up of six precast girders, each consisting of: Two precast delta-shaped sections at the springings Thirty-two precast segments 6.8 ft (2.07 m) long and weighing from 2 to 4.2 tons (1.8 to 3.8 mt).

FIGURE 8.10. Luzancy Bridge, erection of central

section.

FIGURE

rise.

8.11.

Luzancy Bridge, view shmving flat arch

The same design and construction principles used at the Luzancy Bridge were repeated for this series of five bridges, except for some improvements commensurate to the experience gained from the first structure and taking into account the importance of the project. Precasting of the 960 segments was achieved in a factory completely enclosed and using the most modern concrete manufacturing techniques of that period. Each segment was fabricated in two stages in heavy steel forms. Top and bottom flanges were cast first, with high-strength steel stirrups embedded in both units. After strength was achieved, a set of steel forms equipped with jacks was placed between the flanges, which were jacked apart to stress the web pretensioned stirrups. Then the web was cast between the flanges. There was no need for any conventional reinforcing steel in the precast segments. The concrete was vibrated with high-frequency external vibrators, then compressed for maximum

FIGURE

8.13.

Marne River Bridges, typical longitudinal and CTOSS sections.

362


compaction and steam cured for a fast reuse of the forms. The equivalent 2%day cylinder strength was in excess of 6500 psi. Near the precast factory, an assembly yard allowed the segments to be carefully aligned and assembled by temporary prestressing into sections, which were transferred into barges to be floated to the various bridge sites. Each longitudinal girder was thus made up of six sections: The two delta springing sections Two

intermediate

Two

center

five-segment

ten-segment

sections

sections

Handling of these various sections was performed by the Luzancy cableway properly rearranged for the purpose. The stability of the side sections, at both ends, was obtained by temporary cantilever cables anchored in the abutments, while the two center sections were suspended to the cableway until casting of the wet joints was completed and longitudinal prestressing installed to allow the arches to support their own weight. Figures 8.14 through 8.16 show the various sequences of the arch construction, while one of the finished bridges is shown in Figure 8.17. The quantities of materials for the superstructure were very low, considering the span length and the slenderness of the structure:

FIGURE 8.15. tral section.

,tIanre

FIGURE 8.16. tral section.

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Precast concrete: 353 yd:’ (270 m3) Reinforcing steel: 13.2 tons (12 mt) Prestressing steel: 13.2 tons (12 mt) For a deck area of 6540 ft2, the quantities per square foot were:

Precast concrete: 1.46 ft 3/ft 2 Reinforcing steel: 4.0 lb/ft2 Prestressing steel: 4.0 lb/ft2

FIGURE

tion.

8.14.

YI,II IIC Ki\ (‘1 RI iclga, Ed ectetl end scc-

As in the Luzancy Bridge, the high-density concrete placed over the exposed longitudinal prestressing tendons was also used for the roadway

Caracas

FIGURE 8.17.

Viaducts,

Marne River Bridges, completed strut-

Venezuela

363

ture.

FIGURE 8.19. C;lr-atas Viadutr\,

wearing course. The behavior of these bridges has been excellent f’or thirty years.

length as shown in Table 8.1 .2 Preliminary investigations indicated that adequate soil material would probably be found irregularly at great depths. Construction of abutments to resist large bending moments under these conditions would be difficult if not impossible. The decision was therefore made that the abutments would resist only the centered thrust of the arches and that the bending moments applied to the abutment would be reduced, as far as practical, to zero. This required that hinges be located as near as possible to the points of origin of the arches. Because of consideration of long-term creep deformation on buckling of the arch and possible consequences of abutment displacement as might be caused by an earthquake, the decision was made to eliminate a crown hinge, thus resulting in two hinged arches.’ Although the bridges vary considerably in dimensions, they are quite similar in appearance. Because of the valley profile, it was possible to use the same basic design for all three structures. All were designed for AASHO H20-44 loading. Wherever possible, the elements were standardized in order to minimize design and maximize precasting and prefabrication. Pilasters were placed at each end of the arch in Bridge 1 so as to avoid an unpleasant appearance of a change without transition from the main structure to the approach viaducts.

8.3 Caracas Viaducts, Venezuela In Venezuela in 1952 a highway was being constructed between Caracas and La Guaira airport. Alignment of this highway necessitated crossing a gorge at three locations with relatively large bridges. These structures were designed and constructed under the direction of Eugene Freyssinet.’ Although the three bridges are similar in appearance, Figures 8.18 and 8.19, they vary in

FIGURE 8.18. ~&acas

Viaducts, Bridge 1.

Britlgcs 2 ,mct 3.

TABLE 8.1. Caracas Viaduct Arches Height from

Bridge 1 2 3

Total Length

Bed of Gorge

1013 ft (308.8 m) 830 i’t (253 m) 700 ft (213.4 m)

230 ft (70.1 m) 240 ft (73.2 m) 170 ft (51.8 m)

Main Span 498 it (151.8 m) 478 it (145.7 Ill) 453 tt (138.1 m)

364


FIGURE 8.20. Caracas Viaducts, elevation of Bridge 1, from reference 1 (courtesy of Civil Engineering-AXE) An elevation of Bridge 1, Figure 8.20, shows the principal dimensions and foundations of the arch. The three bridges have identical cross sections, Figure 8.2 1. The poured-in-place concrete deck topping varies in thickness from 2 in. (50 mm) at the edges to 7$ in. (190 mm) at the center to provide a transverse slope of 1.5% for drainage. Each deck span, except at the crown, consists of eight precast prestressed I girders. Variations in span length of the deck girders are accommodated by adding or removing standard form units. Identical transversely prestressed precast stay-in-place deck slabs span transversely between the deck girders. Continuity of the deck girders is accomplished by longitudinal tendons placed in a groove in the top of the top flange of the girders.’ Approach piers and spandrel columns over the arches consist of three I-shaped columns of a standard cross section shown in Figure 8.21. A five-segment precast cap beam on the columns re-

Cross section of pier A-A

ceives the eight deck I girders. A perspective of the deck over the piers is shown in Figure 8.22. The precast deck girders, cap beams, and slab are supported on the cast-in-place piers, and the whole assembly is prestressed vertically, transversely, and longitudinally. The center span consists of three parallel double-hinged arch ribs 27 ft 6 in. (8.4 m) on center, Figure 8.21. Each arch rib is a box with a width of 10 ft 6 in. (3.2 m) and a slightly varying depth from 9 ft 6 in. (2.9 m) to 10 ft (3.05 m) at the supporting points of the deck. To provide increased capacity to resist end moments developed by horizontal loads, the width of the ribs is increased to 17 ft (5.18 m) at the spring lines. The 5 in. by 5 in. (127 mm x 127 mm) fillets provided at each inside corner of the box are to reduce the concentration of torsion stresses. Thickness of the bottom flange of the box rib was kept to a minimum to reduce weight on the falsework. The

FCross

5’3” 4

section of arcqly ’

1

7”

itkA

FIGURE 8.21. Caracas Viaducts, typical cross section, from reference 1 (courtesy of Civil Engineering-ASCE).

Caracas Viaducts, Venezuela

365

FIGURE 8.22. Caracas Viaducts, perspective of deck over piers, from reference 1 (courtesy of Civil Engineering-ASCE)

thicker top flange provides the box rib with the required area and moment of inertia for resisting thrust and live-load moments. Design of these structures considered a design wind pressure of 50 psf (2.4 kN/m*). The arch ribs carry part of the wind pressure to which they are directly subjected; the remainder is transmitted to the deck structure by bending of the spandrel columns and the connection of the arch rib to the deck at the crown. The arches were assumed to be transversely fixed in the foundations, the end moment developed in the springings resulting in a slight transverse displacement of the pressure line.* Thus, the deck structure was chosen as the principal member to resist wind loads, requiring the exclusion of all joints in the deck from abutment to abutment. The condition of deck continuity led to the attachment of the deck to the arch on both sides of the arch crown. This was accomplished by prestressing the continuous cables provided over the top flange of the girders and anchoring them into the arch. Six girders were connected to the arch in this manner: the two intermediate girders that do not rest directly on the arch were lengthened to the crown, Figure 8.21.* During construction, an open joint was provided at the crown. In this joint Freyssinet flat jacks staggered with concrete wedges were inserted, acting as a hinge for the arches to adjust the pressure line during different phases of construction. Expansion and contraction of the deck due to temperature, creep, and shrinkage take place over

an approximate length of 1000 ft (305 m), developing approximately symmetrically on both sides of the arch crown. Free movement of the deck structure over the pilasters was accommodated by providing two concrete rockers over each transverse wall of a pilaster. The rockers consisted of a 3 ft 6 in. (1.07 m) high continuous wall throughout the width of the bridge with a continuous Freyssinet-type concrete hinge at both the top and bottom. Approach piers were fixed in the deck at the top and hinged at their footings. Because of their height, these piers have sufficient flexibility to allow movement of the deck without developing appreciable bending moments, the exception being the short stiff piers next to the abutment, which were hinged both top and bottom.* We shall describe the construction procedure for the superstructure of Bridge 1, which was also used for the other two bridges. Because the cableway did not have the capacity to transport the deck girders across the canyon, precasting operations were established at both ends of the bridge. During construction of the foundations, precasting operations were started at both sites at either end of the bridge. When the foundations for the approach piers were completed, the cableway transported and positioned the precast Freyssinet pier hinges to their respective locations, where they were grouted to their respective foundations. Pouring of the piers then commenced, using special steel forms attached to the hinge blocks. Two sets of forms were used in leap-frog fashion to maintain a pouring

366


rate of 5 ft (1.52 m) per day. Because of the hinge at the base of each pier column, the piers required temporary support until the deck girders could be placed. The first 25 ft (7.62 m) lift of each column in each pier was supported by a light steel scaffolding that surrounded each column; the scaffoldings, in turn, were braced together. Succeeding 25 ft (7.62 m) lifts were braced to the previous lift by light timber trusses. As the columns in the piers rose, steel reinforcement was placed: at the same time, holes for vertical prestressing tendons were cast in the concrete by the insertion of l+ in. (38 mm) steel tubes, which were withdrawn lf lio~~t-s after concrete placenlent.3 Upon completion of the three columns of an approach pie r, precast segments of the cap beam were placed atop the columns and prestressed vertically to them as indicated in Figure 8.22. The two intermediate cap beam segments were placed by the cableway and temporaril!; held in position by steel brackets. Four prestresslng tendons were then placed through the cap beam segments and the four vertical 14 in. (38 mm) joints between the segments were packed with a rich mortar. After eight to ten hours the longitudinal tendons in the cap beam \vere stressed and anchored to complete a pier bent, which was then readv to receive the deck girders and slabs. The 137 ft (41.75 m) high pilasters at each end of the arch are four-celled hollow boses 20 by 80 ft (6.1 s 24.4 m) in plan with all walls 4: in. (120.65 mm) thick. They were constructed in lifts with special steel forms that were leap-f‘rogged. ‘l‘en vertical prestressing tendons anchored into the foundation provided stabilit\ against wind forces.3 Upon completion of the abutments and the first approach piers, erection of the bridge deck girders and slabs commenced. It was accomplished with a 126 ft (38.4 m) long structural steel lattice girder gantry, 60 f’t (18.3 m) of which extended as a cantilever. One 48 ft (14.6 m) span, consisting of eight precast beams and 112 precast slabs, required nine working days and a crew of 16 men. When the approach viaduct decks were in place, they were prestressed longitudinally by prestressing tendons placed in the grooves of the top flange of the deck girders, which were anchored at one end into the abutment and at the other end over the arch pilasters. The three arch ribs of the main span were cast in place on a light wooden falsework, which was reused almost in its entirety for the two other bridges. Basically, the system adopted was to erect the timber formwork for casting the arch ribs by

the cantilever method, this formwork being placed by the overhead cableway and held in place by a system of cable stays. Thus, the arch rib was essentially constructed to the quarter-points. The center half-span formwork was constructed as a light wooden trussed arch assembled at the bottom of the canyon and winched into position from the ends of the quarter-span cantilevers. .I‘he timber falsework truss was wedged against the concrete arch ribs already erected. It acted as an arch under the weight of the bottom flange concrete, transmitting its thrust to the cantilevered arch sections previously erected. Later the timber falsework acted compositely with the hardened bottom flange concrete to support the webs and top Hange of the hollowbox arch ribs when they were placed.’ The following discussion describes the erection sequence of the center-span arch ribs.” The first falsework unit in the quarter-span for each arch rib consisted of a timber platform 31 ft (9.45 m) in length with a width of 27 ft 8 in. (8.43 m) at the spring line and a width of 17 ft 2 in. (5.23 m) at the opposite end, Figure 8.23 (Phase 1). This platform was constructed of 3 x 10 in. (76.2 x 254 mm) timbers on edge at l@ in. (267 mm) centers covered on the upper face with 1 in. (12.7 mm) thick plywood. It provided the form for the bottom of the arch rib. For the first section of the quarterspan, three of these units (one for each rib) were placed by the cableway, supported by cable stays A and B, and their position adjusted by hydraulic jacks at the ends of the anchor cable stays. Next four precast Freyssinet hinge blocks were positioned at the spring line and assembled into one hinge block by prestressing them together. Forms were then erected on the falsework for the webs of the arch rib, and placement of concrete commenced, Figure 8.23 (Phase 2). As the weight of each increment of concrete came onto the forms, the cable stays elongated and the geometry of the arch-rib soffit had to be carefully adjusted by the hydraulic jacks. Upon completion of the concreting for the first section of the quarter-span, falsework section 2 was attached to it and supported by two more cable stays, C and D. After geometry adjustment, concreting continued, Figures 8.23 (Phase 3) and 8.24. As a result of the position of the cable stays and the concreting sequence, angular deformations were possible between falsework sections 1 and 2. Therefore, a temporary concrete hinge was placed in the lower flange of the arch rib, which would allow angular deformation but transmit the thrust to maintain equilibrium. When the concreting of

PHASF3

3As.u /

/’ /

-. -.._c

/ /

/. i’

;rF---_ ‘\(

,/ /I /

FIGURE 8.23. Caracas Viaducts, erection and construction sequence, from refet-ence (courtesy of Civil Engineering-ASCE).

3 367

368


FIGURE 8.24. Caxcas Viaducts, comtruction of wch

springings on suspended scaffolding. the second portion of arch rib was completed and geometry adjustment made, the temporary hinge was blocked and the two sections were prestressed together. In the same manner, temporary hinges were used for the remaining sections of the quarter-span arch rib and at each end of the central half-span arch section. The first two sections of arch rib thus became a continuous member supported at the outer end by cable stays, and during construction of the rest of the arch its geometric position was adjusted by cable stay D. The next operation was the erection of the third falsework unit consisting of a trusswork. Its weight was such that it could not be accommodated by the cableway, Therefore, it was assembled at the bottom of the canyon below its position in the arch. The outer end was lifted by the cableway and the inner end by a winch located at the end of the previously concreted section of the arch, Stay cables E passing over the pilaster were attached, and the bottom flange of the new arch rib section was cast, Figures 8.23 (Phase 4) and 8.25, In like manner the next section of trussed falsework was positioned and supported by cable stay F. Next, concrete for the bottom flange of the rib was placed, including small concrete brackets

which protruded below the bottom flange to take the thrust of the 267 ft (81.4 m) central falsework after its positioning, Figure 8.23 (Phase 5). In the last phase of the quarter-span concreting, the vertical webs were formed and concreted, as well as a few narrow strips across the top to provide stiffness to the arch-rib members, which at this stage had a U-shaped cross section, Figure 8.23 (Phase 6). The anchor stay cables were again adjusted to bring the 125 ft (38 m) quarter-span into its proper position. The central 267 ft (81.4 m) falsework span had been assembled at the bottom of the canyon below its final position in the arch, Figure 8.26. The ends of the timber falsework arches were tied together by steel cables acting as ties to keep the arch falsework rigid. The whole central falsework was hoisted into position by winches located at the ends of the cantilevered quarter-span units, Figure 8.27. Once the central falsework was in place and the location of the crown exactly positioned, cement mortar was packed in the gap between the ends of the central falsework and the quarter-span falsework, and extra-flat sand boxes were embedded in the joint for subsequent stripping of the central falsework. After two days, the steel tie cables on the central falsework were released and the winches support-

GENERAL

,/ - -n

DETAIL OF JOlNT OF TOP MEMBER AND DIAGONALS

ELEVATION

OF

FALSEWORK

I A’ ._ - \ a,,II\


ing the section were loosened. At this point the combination of the central trussed falsework and the concreted quarter-span units acted as a complete arch from abutment to abutment. Next, the bottom flanges of the arch ribs were concreted, in a previously arranged sequence, up to the crown on each side, and temporary crown hinge blocks were placed. The other temporary hinges between elements of the quarter span were blocked and the cable stays up to stay D removed. ‘rhe combination of timber falsework and partly built concrete arch ribs continued to be held in position by stays D, E, and F, with a temporary hinge at F onlv. The vertical webs of the arch ribs over the central section were then concreted up to the crown hinge; cable stav D was released; crown concrete was completed; the remaining construction joints were tied with prestressing tendons; and the last cable stays E and F were released. At this point the

concrete arch ribs, less the top flange over the center 260 ft (79.25 m) section, carried themselves as well as the dead load of the entire falsework. Next, the cement joints at the ends of the falsework were destroyed, sand boxes emptied, and, after the steel cable ties had been retightened, the central section of falsework was lowered, Figure 8.28. Falsework elements in the quarter-spans were lowered by hand winches. Spandrel columns were constructed next. Then, following a carefully worked out sequence, the top flanges of arch ribs over the central section were concreted. Upon completion of the arch ribs the deck beams and slabs were placed, in the manner previously described for the approach viaducts, in a symmetrical and simultaneous manner on both sides of the crown. After the deck had been prestressed transversely, it was prestressed longitudinally in the same manner as the approach viaducts. Finished Viaduct 1 is shown in Figure 8.29. In 1973, twenty-one years after the construction of these arches, they were reevaluated to see how they would now be designed and constructed. Figures 8.30 and 8.3 1 compare the actual project constructed in 1952 with the structure as it would have been designed in 1970 (two boxes) and in 1973 (single box). The three-arch-rib and eight-beam superstructure would be replaced by a variabledepth box section (cantilever construction using precast segments) supported on slip-formed piers. The arch remains an appealing and aesthetic structure and might still prove to be competitive; but perhaps the construction technique suggested in the Neckarburg Bridge (Section 8.5.2) might be more appropriate today, either cast in place or precast.

FIGURE 8.28. Caracas Viaducts, lowering center falsework.

FIGURE 8.29. Caracas Viaducts, finished Viaduct 1.

FIGURE 8.27. Caracas false\\x)rk.

Viaducts,

lifting

center

Gladesville Bridge, Australia

371

As constructed in 1952

Possible alternative in 1973

FIGURE

8.30.

Caracas

Viaducts, comparison of longitudinal sections.

8.4 Gladesville Bridge, Australia This precast segmental arch bridge, completed in 1964, spans the Parramatta River between Gladesville and Drummoyne and serves a large section of the northern area of the Sydney Metropolis, Figure 8.32. After award of contract the contractors submitted an alternative design. They proposed that the arch be built on fixed falsework, whereas in the original design part of the arch was to be built on floating falsework and towed into position. The original design called for an arch span of 910 ft (277.4 m). The alternate design increased the clear span of the arch to 1000 ft (305 m) and eliminated the necessity for deep-water excavation for the arch foundations on the Gladesville, or northern, side of the river.4 Total bridge length between abutments is 1901 ft 6 in. (579.6 m). The 1000 ft (305 m) clear span arch consists of four arch ribs, Figure 8.33, supported on massive concrete blocks, known as “thrust blocks,” founded on sandstone on each side of the river. Roadway width is 72 ft (22 m) with 6 ft (1.8 m) wide sidewalks on each side. The roadway has a grade of 6% at each end, and the grades are connected by a vertical curve 300 ft (91.4 m) in ngth over the center portion of the structure. he arch has a maximum clearance, at the crown,

of 134 ft (40.8 m) above the water and not less than 120 ft (36.6 m) above water level for a width of 200 ft (61 m) in the center of the river. Construction of the bridge involved the following main operations4: 1. Excavation for foundation of: a. Arch thrust blocks on each side of the river at the shoreline and partly below water. b. Abutments at the ends of the bridge. Shore pier columns of the approach spans C. on each side of the river. 2. Concreting of the arch thrust blocks, the abutments and columns. 3. Driving of falsework piles in the river and erection of steel falsework to support the hollow concrete blocks and diaphragms forming each of the four arch ribs. 4. Casting of the box-section segments of the arch and diaphragms and the erection of the four arch ribs one at a time. 5. Jacking each rib to raise and lift it off the falsework. 6. Casting of concrete deck beams on each side of the river. 7. Erection of the deck beams to form the roadway over the arch. 8. Paving of the concrete roadway and final completion of the structure.

As constructed in 1952

-t--

20.50 --wft

I

II

I

1 R

-.-

Possible alternative in 1970

-

20.50 Possible alternative in 1973

FIGURE 372

8.31.

Caracas Viaducts, comparison of cross sections.

FIGURE 8.32. Gladesville Bridge, aerial view, from reference 4.

The roadway deck is supported on pairs of prestressed concrete columns, Figure 8.33. The wall thickness is 2 ft (0.6 m), except in the tall columns above the arch foundation where the wall thickness is increased by 6 in. (152 mm). At the top of each pair of columns there is a reinforced concrete cap beam to support the deck girders. During construction it was necessary to provide falsework to support the box segments and diaphragms that make up each of the four arch ribs in the arch. The falsework was made up of steel tubular columns on steel tubular pile trestles carrying spans of steel beams 60 ft (18.3 m) long and a steel truss span of 220 ft (67 m) over a navigation opening in the Gladesville (northern) half of the

FIGURE 8.34. Gladesville Bridge, arch rib falsework and positioning of arch rib segment, from reference 4. falsework. These falsework units were tied together and anchored at each end to the thrust blocks, Figure 8.34. Piling was taken down to rock in the river bed. Steel columns, braced together, formed a tower extending transversely the full width of the bridge at the center of the falsework. Transverse mem-

FIGURE 8.33. Gladesville Bridge, schematic of four arch ribs, columns, and deck, from reference 4.

374


bers, extending the full width of the bridge, above the waterline connected the pile trestles, Figure 8.34. The balance of the falsework was of sufficient width to support one arch rib. Upon completion of erection of an arch rib, the falsework was moved transversely on rails on the transverse members of the pile trestle to a position to enable erection of the adjacent arch rib, until all arch ribs were erected. Equipment installed on the central tower lifted the arch box segments and diaphragms from water level and positioned them. The tower also served as a lateral bent to stabilize the individual arch ribs after they were self-supporting and until they were tied together. 4 The hollow-box segments and diaphragms were cast 3 miles (4.8 km) downstream from the bridge site. The casting yard was laid out to accommodate the manufacture of one arch rib at a time. Each arch rib consists of 108 box segments and 19 diaphragms. Each arch-rib box segment is 20 ft (6 m) wide, with depths decreasing from 23 ft (7 m) at the thrust block to 14 ft (4.3 m) at the crown of the arch, measured at right angles to the axis of the arch. The length of the box segments along the arch varies from 7 ft 9 in. (2.36 m) to 9 ft 3 in. (2.82 m). After the box units were manufactured, they were loaded on barges and transported to the bridge site. The box segments and diaphragms were lifted from the barges to the crown of the arch falsework and winched down to their proper position, Figure 8.34. Diaphragms are spaced at intervals of 50 ft (15.24 m), serving not only to support the slender columns that support the roadway above but also to tie the four arch ribs together. When the units were located in position on the falsework, a 3 in. (76 mm) joint between the precast segments was cast in place. At two points in each rib, four layers of Freyssinet flat-jacks were inserted, with 56 jacks in each layer. The rib was then jacked longitudinally by inflating the jacks with oil one layer at a time, the oil being replaced by grout and allowed to set before the next layer was inflated. Inflation of the jacks increased the distance between the edges of the segments adjacent to the jacks and thus the overall length of the arch along its centerline. In this manner a camber was induced into the arch rib, causing it to lift off the supporting falsework. The falsework was then shifted laterally into position to support the adjacent arch rib and repeat the cycle. Figure 8.35 is a view of the completed four arch ribs, and Figure 8.36 shows the completed bridge.

FIGURE 8.35. Gladesville Bridge, complctrd four arch ribs, from reference 4.

8.5 Arches Built in Cantilever Until the appearance of the concrete cable-stay bridge starting in 1962 (see Chapter 9), long-span concrete bridges were the domain of the arch type of structure. Until 1977, with the completion of the Brotonne Cable-Stay Bridge in France with a span of 1050 ft (320 m), the record length for a concrete bridge had always been held by an archtype bridge. When the Kirk Bridges in Yugoslavia were completed in 1980, the larger arch with a span of 1280 ft (390 m) once again regained for the arch the record of longest concrete span.

FIGURE 8.36. Gladesville Bridge, \ ICI\ of completed bridge.

Arches Built in Cantilever

375

Here is a brief chronology of record concrete arch spans up to 1964: 1930, Plougastel Bridge, France: three spans of 611.5 ft (186.40 m) 1939, Rio Esla, Spain: 631 ft (192.4 m) span 1943, Sando, Sweden: 866 ft (264 m) span 1963, Arrabida, Portugal: 886 ft (270 m) span 1964, Iguacu, River Parana, Brazil: 951 ft (290 m) span 1964, Gladesville, Sydney, Australia: 1000 ft (305 m) span The concrete arch bridge does not enjoy the favor it once did. Modern methods of bridge construction utilizing prestressing, cable stays, and segmental construction have all but eliminated it from contention as a economical bridge type. However, with the application of these modern methods to the older form, and given the proper site conditions, concrete arches may regain some of their lost popularity. 8.5.1 REVIEW OF COSCEPT; SUMMARY OF STRUCTURES WITH TEAZIPORARY STz4YS

The use of temporary stays to facilitate the construction of arch bridges began, perhaps, with the Plougastel Bridge. Temporary prestress tendons were used to provide stability to the short arch cantilever sections emanating from the arch foundations (see Figure 8.5). Prestressing tendons were used to support the f-alsework of the Rio Esla Bridge and were incorporated into the structure. However, the more novel method, which is the birth of today’s technology, was employed in the construction of the Saint Clair Viaduct at Lyon, France, by M. Esquillan. The stability of precast segments was obtained by the use of temporary stays. In the construction of the Caracas Viaduct, Freyssinet extended this concept by using temporary stays to support the falsework and construct a much longer cantilever section of the arch. This same stay system was then used to accommodate the forces produced by lifting the center arch section falsework (see Section 8.3). This concept was partially recaptured for the construction of the Iguacu Bridge in Brazil, where the falsework of the central portion of the arch was supported by temporary stays. The first arch bridge to be constructed using the concept of supporting segmental sections of the

FIGURE 8.37.

fc) Concrete arches built in cantilever with

temporary stays. (n) With stays and pvlons. (h) With stays. spandrel columns, and pylons. (c) With spandrel colunins, tie diagonals and stay’s. arch by temporary stays is the Sibenik Bridge in Yugoslavia. Falsework f-or an approximate length of 88.6 ft (27 m) was supported on Bailey trusses, which were in turn supported by temporary stays, Figure 8.376, consisting of a combination of cables and structural steel rolled shapes. This arch was constructed in nine sections, four on each side and the central closure section. A modification of this concept was used for a second Yugoslav bridge at Pag with a 634 ft (193.2 m) span constructed in seven sections. A further modification was used for the Van Staden Bridge in South Africa, Figure 8.37a, with a span of 656 ft (200 m). A somewhat different concept is where, with the assistance of spandrel columns, the stays act as temporary diagonals during construction, Figure 8.37~. In this manner, the structure is built as a variable-depth Pratt truss. This concept was used for the Kirk Bridges in Yugoslavia. In some instances these temporary diagonal stays may be incorporated into permanent diagonals such that in the final configuration the structure is a truss and not an arch (see Section 8.7.3). In summarizing the construction methods using temporary cable stays, we find two basic categories:

I I I

i I

I I I

I ’

i I

__~ Longitudinal

section

(a)

Erection

at

scheme

a-a approaches

b-b at arch

Cross-sections 376

Cc)

Arches Built in Cantilever

377

FIGURE 8.38. (Opposite) Neckarburg Bridge, erection scheme and sections, from reference 5. (a) Longitudinal section. (h) Erection scheme. (c) Cross section.

Where the arch is supported directly by the temporary stays Where the temporary stays act as diagonals of a Pratt truss during construction Characteristics of the arch bridges using this concept of temporary stays during construction are presented in Table 8.2. 8.5.2 NECKARBURG

BRIDGE, GERMANY

This unique and contemporary arch-supported structure, some 50 miles (80 km) southwest of Stuttgart, crosses the Neckar River near Rottweil, Germany. It is a part of the federal expressway A-81 from Stuttgart to the west of Bodensee with a connection to Zurich, Switzerland. The original scheme proposed by German authorities consisted of a steel girder structure supported on tall piers. Designer-contractor Ed. Zublin, Stuttgart, developed an alternative design consisting of twin concrete arches to support the roadway. The proposal was to construct the arches segmentally by the cantilever method and construct the twin single-cell trapezoidal box girders for the roadway by the incremental launching technique (see Chapter 7). The Austrian method called the Mayreder system was used to construct the arches without scaffolding.5,6 The roadway of this 1197 ft (364.98 m) long structure is approximately 310 ft (94.7 m) above the Neckar River, Figure 8.38. The 507 ft (154.4 m) arch span, Figure 8.39, has a rise of 164 ft (49.85 m). Total roadway width is 102 ft (31.0 m). The

FIGURE 8.39. Neckarburg Bridge, completed arch (courtesy of Willhelm Zellner).

FIGURE 8.40. Neckarburg Bridge, arch just before closure (courtesy of Willhelm Zellner).

structure is constructed as two independent parallel structures with a 1.8 ft (0.54 m) gap in the median. Roadway spans are 98 ft (30 m) in the approach sections and 72.6 ft (22.14 m) over the arch. Each independent arch rib is a two-cell box. The arch ribs were constructed in symmetrical halves, Figure 8.40. The cuEved formwork was 43 ft (13.1 m) long, the first 23.3 ft (7.1 m) of the form clamped to the previously constructed arch segment and the remaining 19.7 ft (6 m) remained to cast the next segment increment. The first 23.3 ft (7.1 m) of arch segment at the arch foundation was constructed by conventional forming methods. There are 14 segments on each side of an arch rib and a closure segment at the crown of each arch. The exterior dimensions of each two-cell arch rib are 21.3 ft (6.5 m) wide by 9.8 ft (3.0 m) deep. Exterior webs vary in thickness from 10 to 11 in. (260 to 280 mm), and the interior web is 6.3 in. (160 mm) thick. The arch rib was cast in two operations -first the bottom flange and second the webs and top flanges.” Piers supported by the arch or independent foundations are of a constant section and slipformed by conventional methods. Sliding bearings are used at the abutments and the short stiff piers 1 and 13. The remaining piers are hinged to the superstructure deck such that the elastic piers can follow the superstructure movement.5’6 During construction, as each half-rib was cantilevered out from its foundation, it was supported by a temporary system of Dywidag bar stays, Figures 8.38, 8.41, and 8.42. After completion of the arch, the temporary stays were removed, except those required to stabilize the arch during the incremental launching of the superstructure deck. Dywidag bar stays were anchored either to a pier foundation or to Dywidag rock anchors in the side of the valley.5

TABLE Year Name Sib&k

Yugfelavia

01

8.2 Span,

(:onwuctlon

(In)

I Y64-66

x07 (246)

Characteristics of Arch Bridges Constructed with Cable Stays It. Stiay Mrthod

Arch C~,tlSt~U~tlOtl

Staic Arc II Type

Nine wction\ on lalsework of X8.6 and Y5 It (27 and

Simple ,,rrCa\t glrders ol 76.4 tt (23.30 Ill) spat, rmde continuott9

2Y 111)

Vat1 Statlen

Niesenbachbrucke

South Al&a

Austria

I Y66-67

634 (193.2)

A b o u t I970

656 (200)

I Y73

394 ( 120)

‘I‘hrer stay\ Iormed from rolled \trel shape\ and cables, auxiliary pylon, the longer stq being addttxmally su}q”“ted on toIUrnllS

I brce-cell ICC 1611gulnr box

Segment\ 19.6 It

( 2 0 m) l o n g

Multif~le stays wpporting the arch dire< tly with the aid of an auxiliary pykm

De< k ‘I’ypc

Mobile forms pet-mittinK thr successive con struction of 21 11 ( 6 . 5 1x1) tong seg-

Arch Scheme Possibility 01 carI-ertmg the thrust at the crown by a battery 01 hydraulic jacks

Pos5ibiltty 01 carrecting the thrust at the c~owtt by a battery ol hydraulic jacks

Simple precast girders 01 55 tt (16.8 m) span mnde ~OIIIIIIUOU\

Fixed

Continuous double-T \pan 01 65.6 It (20 111)

Fixed

Hollow slab of 2 It ( 0 . 6 0 m) thtc knes5; 5 0 I t ( I5 m) continuous span\ < r,nstru
Hmgrd ‘it t h e two qwingmgs of thr arch

“,e”tS Hvkawaru

Japan

1973-74

55x ( 170)

Formwo~-k parttally tufq>orted by a stay for- the firs1 SC< lion between the abutment and the hrst Fpandrel column. Alterward, ~onstrution by successive cantilevers by \cgments 01 I I I t (3.4 111) Ienfgh.

Rectangular twocell box. Near the springings, the width increases Itnearly from 26 to 52 It (X to I6 m) to ,mprwe the tatera1 stahdttv

Remarks

Horinmtal curvature of deck is R = 1092 It (332.8 m)

Krummharhbr-ilrke

Switxrland

1976-77

407 (124)

Constructrd I” segmentc rd 20.5 tt (6.25 m) length

‘fwo parallel arches. dlaphragm at the columns. t:ach arch IS a solid r-e< tangular rib 3 6 x 6.6 It (1.1 x 2.0

Double ‘I’ wth continuous cpns varymg between 32.8 ‘I”1 65.6 II ( IO to 20 111)

Fixed

m) Neckarhurg

Schwar/waswhrtrke

(&many

Switzerland

I977

1977-79

507 ( 154.4)

(:onstrurted of

374

I’wo p”‘“llel arches. each
IGxed

Succesive curltilever segments 16.4 to 17.7 It (5 to 5.4 m) 111 length

Rectangular \lah

E‘ixed

Railroad bridge

4I3

Constructed

Arch c onsistc

Fixed

Railroad bridge

(126)

mohile l.orms such that a complete panfzl was cast, inc tuding the arch. the column, and the de< k

wccessive segments 19.7 It (6.0 rn) In length

(114)

Akayagwa

K i r k Bridge (smaller arch)

JZ3ptl

Yugoslavia

197R

197x

x00 (244)

st‘lys uwtl ‘IS diagonals ot a Pratt tn,s\

with

of a

rwr,-
rectangular thin slab between two COlUmtIS

ue< t‘rnp I‘N three-cell box

Fixed: posslhltity ot correcting the thrust at the crown hy a hattery ot hydraulic jacks

Rectanpdar three-cell box

(Same as ;tbove. for smaller arch)

FIGURE 8.41. TSecharburg

Bridge, temporaq

Dvwi-

dag bar stays supporting cantilevered arch rib (courtesy of Willhelm Zellner).

FIGURE

8.43.

Neckahur-g Bridge, hudling

of’ deck

girder.

FIGURE 8.44. Neckarburg Bridge, close-up of

launching nose.

FIGURE 8.42. Neckarburg Bridge, temporary Dywi-

dag bar stays supporting cantilevered arch rib (courtesy of Willhelm Zeliner). The trapezoidal box girders of the superstructure deck were constructed behind the Singen abutment and incrementally launched “downhill” toward the Stuttgart abutment, Figure 8.43. A close-up of the launching nose is shown in Figure 8.44. Overall girder width is 48.8 ft (14.9 m) with a constant depth of 7.5 ft (2.3 m). Girder segments were cast in lengths of 65.6 ft (20 m). The lift and push combination of hydraulic jacks (see Chapter 7) launched the girder in 10 in. (0.25 m) increments. To maintain deformations of the arch and

piers, resulting from the horizontal forces of the incremental launching operations, within allowable limits, the tops of the piers were tied back to the abutments and the arch was tied back by the temporary stays used during the arch construction. An innovation introduced by Zublin on this project was the use of bearings for the incremental launching that remained as permanent bearings. Prior procedure had employed a system of temporary bearings for the incremental launching and then a transfer to permanent bearings5 8.5.3

NIESENBACK

BRIDGE,

AUSTRIA

This is a two-rib arch structure utilizing the free cantilever construction method for each half-arch, Figure 8.45. The arch has a span of 394 ft (120 m) with a rise of 123 ft (37.5 m). Each arch rib is a two-cell box with exterior dimensions of 16.4 ft (5 m) wide by 8.2 ft (2.5 m) deep. The roadway consists of a concrete slab and girder system with an overall width of 57.7 ft (17.6 m). Although the longitudinal axis of the arch is in a straight line, the

Arches Built in Cantilever Structure during construction

WJO

FIGURE

8.45.

-

Final structure

I I

Hll‘SPYlO”

590

!=

s&m

cl

Niesenback Bridge, elevation, plan, and cross section, from reference 7.

roadwav it supports has a centerline radius, in plan, of 1092 ft (332.8 m). The curved roadway structure has spans of 65.6 ft (20 m) over the arch and is supported by two 3.3 ft (1.0 m) square piers, one on each arch rib. At the arch foundations, roadway support is by a wall pier with dimensions of 4.6 ft (1.4 m) by 33.8 ft (10.3 m).

Each two-cell box arch rib is constructed by the cantilever method, using a 41 ft (12.5 m) long traveling form. The form clamps to the preceding construction such that a 19.7 ft (6.0 m) segment can be cast. A crew of seven men was able to cast a segment on a weekly cycle. To keep moments in the cantilevering arch to a minimum during construction, the cantilevered

382


portion of the arch was supported by a system of Dywidag bar stays, Figure 8.45. Stay stresses are monitored at each stage of construction to maintain a nearly moment-free condition in the arch. Dywidag bars used in the stays were 1 in. (26.5 mm) diameter and were used because they were easily coupled and could be reused.’ 8.5.4 KIRK BRIDGES, YUGOSLAVIA

These structures connect the mainland with the Island of Kirk in the Adriatic Sea. In between is a small rocky outcropping known as St. Mark, such that from the mainland to St. Mark is the world’s longest concrete arch with a span of 1280 ft (390 m) and from St. Mark to Kirk is the seventh longest concrete arch with a span of 800 ft (244 m), Figures 1.40 and 8.46. Because the distance between the shores of the mainland and St. Mark is 1509 ft (460 m), the arch support is partially founded in the sea, Figure 8.47. The arch reaction of approximately 15,400 tons (14,000 mt) is accommodated by the inclined pier in the sea, which takes 9900 tons (9000 mt) to the rock, while the nearlv horizontal box structure

1/3.00

above sea level takes the other reaction component of 6600 tons (6000 mt). A system of temporary stavs was used to support the arch as it was progressively cantilevered out from the springings, Figure 8.48. These temporar) stays were used as the top chord and diagonals of a temporary variable-depth Pratt truss during construction, Figures 8.48 and 8.49. The arch rib consists of a three-cell rectangular precast box, which was cast in segment lengths of 16.4 ft (5 m) and assembled with cast-in-place joints, Figure 8.48. .4 view of the completed arch with spandrel columns is given in Figure 8.50. 8.6 Rigid-Frame Bridges

Another bridge type that lends itself to the contemporary segmental concept is the rigid-frame bridge. Unfortunately, segmental construction has not often been applied to this type of structure. The reason is probably that the segmental concept is associated with the conventional girder type bridge, and designers have given little consideration to applying this method to the rigid-frame bridge. Hopefully, the few examples that follow will stimulate thinking about this type of structure.

!

Section 1

Section 2

ELEVATION

FIGURE 8.46. Kirk Bridges, elevation and sections.

390m

1

L

I

/

300 t

0 0 00 ~

, -'lo.

-

,

33 50

1

-

00

, 0 00 -

-19OQ

33 50

FIGURE 8.47. Kirk Bridge, foundation detail. 383

FIGURE

tion.

8.48.

Kirk Bridge, erection of first arch sec-

FIGURE 8.50. Kirk Bridge, completed arch.

FIGURE

FIGURE 8.49. Kil k Bridge, erection apploarhing

pleted

crown.

8.51.

structure.

Saint Xlichael 131 itigc, LI~‘M ot the com-

7L’O’

_-

-

__- _-__. .--

FIGURE

8.6.1

8.52.

I-.--“..

Saint

--_

Michael

2Tr;LO’---

--.

Bridge,

partial

SAINT MICHEL BRIDGE IN TOULOUSE, FRANCE

This beautiful structure, Figure 8.51, appears as a succession of arches with inclined legs, crossing the two branches of the Garonne River in the southern city of Toulouse, France. Typical dimensions of a rigid frame are presented in Figures 8.52 and 8.53. Because the bridge replaced an obsolete structure resting on masonry piers, it was possible to

longitudinal

section

construct the inclined legs on suspended scaffolding using temporary ties anchored to the masonry piers before they were demolished, Figure 8.54. The longitudinal girders were cast in place between the legs to complete the rigid frame. Over each pier an expansion joint with laminated bearings is provided in the roadway slab, Figure 8.54. Another view of the finished bridge is presented in Figure 8.55.

FIGURE 8.53. Saint Michael Bridge, typical section.

9 4

@oprmc bearing

Q

4 Q

4

‘. . ‘A Extcling FIGURE 8.541.

Saint Michael Bridge, construction sequence at typical pier.

FIGURE 8.55. Saint Michael Bridge, finished structure.

moaonry pier

FIGURE 8.56.

Briesle Maas Bridge, general view.


386 8.6.2 BRIESLE MAAS

BRIDGE, NETHERLANDS

The Briesle Maas Bridge near Rotterdam, completed in 1969, is a distinctive structure with its V-shaped piers, Figure 8.56. This bridge, crossing the Meuse River, is situated in an area reserved for pleasure boating and recreational purposes. It was therefore considered essential to maintain a high degree of bridge aesthetics. Although the design is perhaps not the most economical, it was chosen to meet the aesthetic requirements. The three-span superstructure consists of a 369 ft (112.5 m) center span with end spans of 264 ft (80.5 m). Transversely, the superstructure consists of three precast single-cell boxes, joined at their flange tips by a longitudinal closure pour and transversely prestressed, Figure 8.57. The hollow inclined legs of the V piers are structurally connected to the deck structure by post-tensioning, and the V pier is supported at its base through neoprene bearing pads on the pile cap foundation, Figures 8.58 and 8.59. The superstructure, with the exception of a few cast-in-place closure joints, is composed of precast segments. Shear forces, mainly concentrated in the webs, normally are transferred to piers or columns by a diaphragm. Prefabrication prevented this solution

as-25

485 CROSS

FIGURE

8.57.

in this project, however, as the additional weight in the pier segments would have increased intolerably. Shear stresses were maintained at an acceptable level by increased web thickness and by triaxial prestressing. At the moment that the midspan closure pour of the center span is consummated, the bending moment at this joint is zero. With time this moment increases, as a result of creep, to a significant percentage of what would occur if the bridge were built as a continuous structure on falsework. Prestressing to accommodate both conditions cannot be given maximum eccentricity, and it becomes both difficult to execute and expensive. .4 considerable amount of prestressing was saved by eliminating the condition of zero stress at closure and therefore preventing creep. This was accomplished by inducing an upward reaction under segments 7 and 72, Figure 8.59, after joint closure. Simultaneously with the increase of these reaction forces, prestressing tendons in the central span were stressed. Upon completion of the end spans the induced forces were released automatically by prestressing the end spans. Segments were produced at an existing casting yard 68 miles (110 km) from the bridge site. A long-line precasting bed (see Figure 11.37) was

3525

i

485

Lc.75

/

SECTION

Briesle Maas Bridge, transverse cross section.

LONGITUDINAL SECTION WITH CABLE PROFILE

FIGURE

8.58.

Briesle Maas Bridge, longitudinal section with tendon profile.

Rigid-Frame Bridges

A= B= C= D= E =

Steel frame Jacks R u b b e r bearing p a d s lolnts Counter weight

387

F = Joint G = Temporary support H = Scaffolding J = Joint

FIGURE 8.59. Briesle Maas

used with a length equal to a half-span-that is, one cantilever. Three sets of segment forms were employed to cast a total of 234 segments, averaging 78 reuses. Segments were transported to the bridge site by barge. The various stages of erection are indicated in Figure 8.59. A special structural steel frame was used to position the inclined precast hollow-box legs of the piers and to support the seven precast roadway girder segments before casting the joints at the corners of the delta pier portion of the structure. This frame was also utilized to balance the pier during erection of the remainder of the roadway girder segments and to adjust, by means of jacks, the loads in the inclined legs of the pier during various stages of erection. Upon completion of the balanced cantilever erection about both piers, temporary supports were placed under segments 7 and 72 (the extreme end segments of the partially completed end spans) so that the temporary steel frames under the piers could be removed. At this point both halves of the structure were in an unstable equilibrium condition, therefore, counterweights were placed over the supported segments, Figure 8.59, to prevent the half-structures from toppling over. Jacks atop the temporary supports were used to adjust the position of the bridge halves with respect to one another and to induce the upward vertical reaction forces previously discussed. Also, differences in elevation between the three box girders

Bridge, erection sequence. were adjusted by these jacks. After casting the center-span closure joint and stressing in the center span, the remaining segments in the end spans were placed on falsework, Figure 8.60; closure joints were cast; and longitudinal and transverse prestressing was completed. All segments in the balanced cantilever portion of the structure were placed by a floating crane. Because of the crane’s small reach, it could not place the last five segments needed to complete the end span. Therefore, it placed them on a small dolly installed on top of the falsework, which would roll them into their final positions. To avoid dismantling the falsework after completing one girder and reinstalling it under the next, it was constructed so that it could be lowered and moved transversely into position, Figure 8.60. A close-up of the piers of the finished structure is shown in Figure 8.61. 8.6.3 BONHOMME BRIDGE, FRA.VCE

The Bonhomme Bridge over the Blavet River in Brittany, France, was designed and built between 1972 and 1974, Figure 8.62. This three-span slant-leg portal-frame bridge has a center span of 481 ft (146.7 m) and end spans of 223 ft (67.95 m), Figure 8.63. The span between the foundations of the slant legs is 611 ft (186.25 m). A tubular steel framework was used to support the slant legs temporarily until closure at midspan, Figures 8.64 and


I

smcu A-.

-k--J FIGURE

span.

8.60.

Briesle Maas

Bridge, erection falsework for last five segments in the end 8.65. This structure was built by the cast-in-place balanced cantilever method. For adjusting the geometry of the bridge, flat jacks were placed under the legs and at midspan. A detail of the adjusting jacks placed on top of the temporary support is shown in Figure 8.66. Flat jacks and sand boxes were used both to adjust the geometry of the bridge before closure was achieved at midspan and later to release the energy stored in the legs of the temporary supports, which were loaded with the full weight of the bridge.

FIGURE 8.61. Briesle Maas Bridge, close-up \iew of V piers.

FIGURE 8.62.

KERVIGNA&

&ORIENT L N

Ronhomme Bridge over Blavet River.

67.95

282.60 146.70

67.95 /

FIGURE 8.63. Bonhomme Bridge, elevation.

TW'FII FRS FOR CIP CONSTRUCTlON A

FIGURE 8.64. Bonhomme Bridge, construction stages.

+

++

++

++

+

+ +

+ + ++ +i \

FIGURE 8.65. Bonhomme Bridge, temporary support 389


390

The scheme is a very satisfactory one in terms of both the aesthetics of the finished structure and simplicity of construction. However, it may be used only when site conditions allow the foundations of the temporary supports to be established safely at a reasonable cost. Figure 8.67 shows the temporary supports during the balanced cantilever construction of the bridge.

8.6.4 MOTORWAY OVERPASSES IN THE MIDDLE EAST

CONI’RFTF

CAF

_.

;

,.

13'-

FIGURE 8.65. (Continued)

i5

'

The use of precast segmental construction for the Alpine Motorways in southern France was described in Section 3.15. It was shown how mass production could be applied to the construction of a large number of similar overpasses. This experience was repeated recently in a middle eastern country for the construction of 17 overpass structures over an existing freeway, Figure 8.68. To minimize disturbance of freeway traffic, it was felt that a three-span rigid-frame structure with inclined legs would be an attractive solution. Dimensions are shown in Figures 8.69 and 8.70. The total deck length of 252 ft 3 in. (77 m) is divided into 32 precast segments for each of the twin box girders. Deck width of the overpasses is either 36 ft (11 m) or 46 ft (14 m). The same box section is used for all structures, and the cast-in-place longitudinal closure strip varies as required. The slant legs are precast in the same plant where the deck segments are produced. The typical erection sequence is shown in Figure 8.71. A temporary bent founded at the edge line of the new freeway is used to place and adjust the precast legs on either side of the bridge. Segments are placed in balanced cantilever from the special segment located atop the slant legs. A light temporary bent in the short side spans is used to reduce the bending moment in the slant legs during construction. After completion of the deck and removal of all temporary supports, the structure is in effect a two-hinged arch with vertical restraints at both ends. The bridges were analyzed for earthquake and large thermal variation loads (seasonal variation of 120°F and temperature gradient between top and bottom flange of 18°F). Figure 8.72 shows a detailed view of the inclined legs and the temporary support during construction.

FIGURE 8.66. Bonhomme Bridge, details of bearing of concrete cantilever on temporary support.

b . , h

392


FIGURE 8.67. Honl~ornrne

construction.

Bridge, during cantilever

I I--/ : r] ;-y j /

+----: : ‘--: I //j, / ,

FIGURE 8.70. Motorway Overpass Frames, cross section and elevation of inclined legs.

FIGURE 8.68. Motorway Overpass Frames, general view.

8.7.1

As with rigid frames, segmental construction has seldom been applied to truss bridges. Once again the designer must realize that the principles of segmental construction, together with imagination, can be applied to bridge structures other than the conventional girder bridge.

149.3 I

FIGURE

RETROSPECT ON CONCEPTS FOR CONCRETE TRUSS BRIDGES

Trusses were used in all long-span cantilever steel bridges, and it was logical to conceive of the application of this type of structure to prestressed concrete. An interesting example of such an approach is presented in Figure 8.73, in which an original sketch made in 1948 by Eugene Freyssinet for the design of a precast prestressed concrete truss is reproduced. The studies were applied to two specific examples:

8.7 Truss Bridges

3Segments-8’0”

1 Plain concrete

,9’-0”

19’-0”

B $1

I

“:

I I

8.69. Motorway Overpass Frames, longitudinal section.

51’-6”

A I

Truss Bridges

393

FIGURE 8.71. Motorway Overpass Frames, erection sequence. (a) Stage 1. (b) Stage 2. (c) Stage 3. (d) Stage 4.

6

A bridge over the Hanach River near Algiers, Algeria, with a clear span of 400 ft (123 m), Figures 8.74 and 8.75. A major crossing of the Rhine River at Pfaffendorf, Germany, with a main span of 600 ft (180 m)

FIGURE 8.72. Motorway Overpass Frames, detail of inclined leg and temporary support.

These studies were very encouraging from the viewpoints of both economy of materials and simplicity of construction. The deck was to be entirely precast, with members assembled by prestressing. Construction would proceed in balanced cantilever from the main piers until reaching midspan closure, where adjustment of the deck geometry and loads in the members was provided by jacks.

FIGURE 8.73. Original sketch of E. Freyssinet for a concept of prestressed precast concrete truss (1948).

__ ELEVATION -

FIGURE

8.74.

Concept of a truss bridge.

l/2 COUPE B-B

l/2 C O U P E A - A

FIGURE 8.75.

Concept of a truss bridge.

The use of I girders at 7 ft (2 m) spacing for the precast deck would not be considered today as the optimum design. One of the authors, who was involved in the studies with E. Freyssinet, remembers also that many technological problems such as the connection details between diagonals and chords were not completely solved. Neither of these two designs reached the construction stage, and the concept was rapidly forgotten before its potential could be objectively ascertained. Oddly enough, the designers of steel structures followed a similar path. Abandoning prematurely the concept of truss structures, which had allowed such outstanding structures as the Firth of Forth Bridge to be built all over the world, they turned to web girder structures and closed box sections with all the critical problems they entailed, such as elastic stability. Perhaps it is time to reassess some major design approaches in both steel and concrete for very long spans. 8.7.2 MANGFALL

anced cantilever, as construction started at one abutment and proceeded to the opposite abutment by progressive placement. Temporary intermediate piers were used as required to reduce the cantilever stresses. Figure 8.77 shows an interior view. The lower flange is used as a walkway for pedestrians and for bicycles. The railing in the center surrounds an opening in the bottom flange where stress conditions do not require the concrete area. Figure 8.78 is an interior view looking through one of the floor openings, and Figure 8.79 is another interior view. 8.73 RIP BRIDGE, AUSTRALJA

The recently completed Rip Bridge, Figure 8.80, north of Sydney, Australia, has a center span of

BRIDGE, AUSTRIA

The Mangfallbriicke in Austria, Figure 8.76, on the autobahn between Munich and Salzburg was constructed in 1959. This structure is perhaps best described as a large box girder with the webs being a trusswork. Total length is 945 ft (288 m) from abutment to abutment; the center span is 354 ft (108 m) with side spans of 295.5 ft (90 m). It was constructed as cast-in-place segmental using the free cantilever method. However, it, was not bal-

FIGURE 8.76. Mangfallbriicke, general view.

396


FIGURE 8.77. hl;~t~gf;llIbt.ucke, trusswork.

interim view showing FIGURE 8.80. Rip Bridge, general view.

FIGURE 8.78. Mangfallbl.iicke, through floor opening.

FIGURE

8.79.

interior view looking

M,tngfallbrticke, general intertor view.

600 ft (182.88 m). The identical cantilever trusses, which sit symmetrically on either side of the crossing, reach out 240 ft (73.56 m) toward each other to support a 122 ft (37 m) drop-in simple span at their extremities, Figure 8.8 1. The erection scheme is illustrated in Figure 8.82. Note that cable stays were used as diagonal members during construction to support the arch segments. Temporary falsework bents were used at each panel point of the truss on the landward side of the main piers. Precast concrete elements were delivered from a precasting plant some 80 miles (130 km) from the site. Each panel of the lower chords of the truss was assembled from five precast I-shaped elements with a 1 ft (0.3 m) longitudinal pour strip between the flange tips. Similarly, the upper chord was assembled from five rectangular two-cell precast members. Erection of one of the lower chord members is shown in Figure 8.83. The exterior two I-shaped lower chord members are supported by the diagonal stays, while the interior three elements of the lower chord are supported by a transverse beam arrangement from the exterior two during construction. Each diagonal member was assembled from longitudinally split halves, which, when brought together, encase the diagonal prestress tendon stays, incorporating them into the structure by concrete poured in place between the two halves. The upper chord or deck members are erected after the vertical members along with temporary falsework to support the deck panels, while the cast-in-place concrete is placed between the deck elements and transversely prestressed.

Truss Bridges

397

8.81. Rip Bridge, elevation and cross sections.

FIGURE

Prestress cable to support lower member

f

FIGURE 8.83. Rip Bridge, erection of knver chord.

Abutment I.

,

The deck performs as a prestressed concrete tension member. As construction proceeds, additional prestress is progressively added to ensure that the deck remains in compression. 8.7.4 Location of

<-ggri.n

CONCEPT FOR A CROSSING OF THE ENGLISH CHANNEL

Certain projects for crossings, such as of the English Channel between France and Great Britain, the Straits of Messina, and even the Straits of Gibraltar, have exerted a powerful fascination on the minds of the great engineers of this century. FIGURE 8.82. Rip Bridge, erection sequence.

FIGURE 8.84. E‘t-eyssinet’h c-oncept of‘ preconfinetl Channel with a set-ies of‘ 2000 f’t (612 111) slxtns.

concrete at-ch crossing the English

Refmences Eugene Freyssinet was no exception, and he spent the last years of his long professional career studying the crossing of the English Channel with a series of 2000 ft (612 m) long prestressed concrete spans. The many worthwhile ideas contained in this concept are not likely to be developed soon, or even by the turn of the century. Figure 8.84 presents an elevation of a typical 2000 ft (612 m) span, which was contemplated as a prestressed concrete composite truss. Major members of the truss were not of conventional prestressed concrete, because such high stresses had to be accepted to keep the weight of the span within acceptable limits. A new material to be used for that purpose had occupied Freyssinet’s mind for several years and had even been laboratory tested for confirmation of the concept. When a concrete member is completely confined in an envelope that creates permanently biaxial transverse compressive stress, it will resist safely much higher stress than if subjected to a monoaxial stress or reinforced conventionally with untensioned transverse reinforcing (such as spirals in a circular column). From a technological point of view, the permanent active restraint creating the biaxial transverse compressio:l is easily achieved in a member that has a circular cross section by confining it in a high-strength steel pipe or within a continuous spiral of prestressing steel wires, which are prestressed at the time the concrete is cast. This material, which could be called “preconfined concrete,” has extraordinary properties such as total absence of brittleness and a capability

399

to sustain several times as much longitudinal compressive stress as a reinforced concrete member without excessive strains, provided it is initially loaded to offset the initial strain. Such a project and such a material could not be developed in a short period of time. They are mentioned here at the close of this chapter as a conceptual heritage, which it is our duty to make functional.

References 1. E. Freyssinet, “Largest Concrete Spans of the Americas-Three Monumental Bridges Built in Venezuela,” Cizd Engineering-ASCE, March 1953. 2 . J e a n Muller, “ L a r g e s t C o n c r e t e S p a n s o f t h e Americas-How the Three Bridges Were Designed,” Civil Engineering-ASCE, March 1953. 3.

R o b e r t Shama, “Largest Concrete Spans of the .Americas-How They Were Built.” C/zlil Et/g/tff’f’rrng--AXE, M a r c h 1 9 5 3 .

4. Anon.,

“ N e w B r i d g e o v e r P a r r a m a t t a R i v e r at Gladesville,” Main Roaak, Journal of the Department

of Main Roads, New South Wales, December 1964. 5. Anon., “Talbriicke Rottweil-Neckarburg,” ZublinRundschau, Heft 7/8, Dezember 1976, Stuttgart, Germany.

6. “Arch Slipformer Shuns Ground Support to Cross Valley,” Engineering LXrews-Record, June 1, 1978. 7. Anon., “Niesenbachbriicke, Bogen im Freien Vorbau,” Austria 1970-74, York.

FIP Congress 1974, New

.

9 Concrete Segmental Cable-Stayed Bridges

9.1

INTRODUCTION 9.1.1

9.2 9.3 9.4 9.5 9.6

Historical

Review

9.1.2 Advantages of Concrete Cable-Stayed Bridges 9.1.3 Structural Style and Arrangement LAKE MARACAIBO BRIDGE, VENEZUELA WAD1 KUF BRIDGE, LIBYA CHACOlCORRIENTE.5 BRIDGE, MAINBRiiCKE, GERMANY TIEL BRIDGE, NETHERLANDS

9.1

ARGENTINA

Introduction

The concept of supporting a beam or bridge by inclined cable stays is not new, and the historical evolution of this type of structure has been discussed in the literature.‘-‘j Although the modern renaissance of cable-staved bridges is said to have begun in 1955, with steel as the favored material, in the last two decades a number of cable-stayed bridges have been constructed using a reinforced or prestressed concrete deck system. In recent years several concrete cable-stayed bridges have been built in the long-span range. In at least four current projects, alternative designs in concrete and steel have been prepared for competitive bidding. Cable-stayed bridges are extending the competitive span range of concrete bridge construction to dimensions that had previously been considered impossible and reserved for structural steel. To date, approximately 2 1 concrete cable-stayed bridges have been constructed, and others are either in design or under construction. A tabular summary of concrete cable-stayed bridges is presented in Tables 9.1 and 9.2. 400

9.7 9.8 9.9 9.10

PASCO-KENNEWICK

BRIDGE,

U.S.A.

BROTONNE BRIDGE, FRANCE DANUBE CANAL BRIDGE, AUSTRIA NOTABLE EXAMPLES OF CONCEPTS 9.10.1

Proposed Great Belt Bridge, Denmark

9.10.2 Proposed Dame Point Bridge, U.S.A. 9.10.3 Proposed Rock-A-Chucky Bridge, U.S.A. REFERENCES

9.1.1 HISTORICAL REVIEW

Since the beginning of the cable-stay renaissance in 1955, whether for technical or other reasons, structural steel has been the preferred construction material. In 1957, however, considerable excitement was generated when Prof. Riccardo Morandi’s prize-winning design of a prestressed concrete 1312 ft (400 m) center span cable-stayed bridge for the Lake Maracaibo crossing was announced. Regrettably the Lake Maracaibo Bridge was not constructed as originally conceived. The modified structure, built in 1962, is generally considered to be the first modern cable-stayed bridge. However, the Lake Maracaibo Bridge was preceded by two little-known concrete cable-stayed structures. The first concrete structure to use cable stays was the Tempul Aqueduct crossing the Guadalete River in Spain. ’ Designed by the famous Spanish engineer, Prof. Torroja, who has introduced many original concepts in prestressed concrete, this structure has a classical three-span symmetrical cable-stayed bridge configuration with two pylons.

401

Zntroduction TABLE

9.1.

Bridge

8 9 10 11 12 13 14 15 16 17 18 19 20 21 22 23 24 25 26 27 28

Tempul Benton City Lake Maracaibo Dnieper River Canal du Centre Polcevera Viaduct Magliana Danish Great Belr Danish Great Belt” Pretoria Barwon River Mount Street Wadi Kuf Richard Foyle Mainbrticke ChacolCorrientes River Waal Barranquilla Danube Canal Kwang Fu Pont de Brotonne Carpineto Pasco-Kennewick M-25 Overpass Ruck-A-Chucky’ Dame Poinr East Huntington” Weirton-Steubenville’

Concrete

Cable-Stayed

Bridges-General

Location

Type

Guadalete River, Spain Yakima River, Wash., U.S.A. Venezuela Kiev, U.S.S.R. Obourg, Belgium Genoa, Italy Rome, Italy Denmark Denmark Pretoria, S. Africa Geelong, Australia Perth, Australia Libya Londonderry, N. Ireland Hoechst, West Germany Parana River, Argentina Tiel, Holland Barranquilla, Columbia Vienna, Austria Taiwan Normandy, France Province Poetenza, Italy State of Wash., U.S.A. Chertsey, England Auburn, California, U.S.A. Jacksonville, Florida, U.S.A. East Huntington, W.Va., U.S.A. Weirton, W.Va., U.S.A.

Aqueduct Highway Highway Highway Pedestrian Highway Highway Highway & rail Highway & rail Pipe Pedestrian Pedestrian Highway Highway Highway & rail Highway Highway Highway Highway Highway Highway Highway Highway Rail Highway Highway Highway Highway

Data

Spans (ft)d 66- 198-66 [email protected]@57.5 525-5@771-525 216.5-472-2 16.5 2@220 282-664-689-460 476-176 multispans 1132 multispans 1148 2@93 180-270-180 2e116.8 320-925-320 230-689 485.6-308 537-803.8-537 312-876-312 228-459-228 182.7-390-182.7 220-440-440-220 471-1050-471 100-594-100 406.5-981-406.5 2e180.5 1300 650-1300-650 158-300-900-608 820-688

Year

Completed

1925 1957 1962 1963 1966 1967 1967 Delayed by funding Delayed by funding 1968 1969 1969 1971 Project abandoned 1972 1973 1974 1974 1974 1977 1977 1977 1978 1978 Design completed Design completed Under constt-uction In design

“Design by White Young and Partners. bDesign by Ulrich Finsterwalder. “Alternative design with structural steel. 1 ft = 0.305 m.

The stays were introduced to replace two piers that were found to be too difficult to construct in deep water. Thus, the stays were introduced to provide intermediate support in the main span. On July 5, 1957, a stayed structure crossing the Yakima River at Benton City, Washington, was opened to traffic. Designed by Homer M. Hadley, the structure has a total length of 400 ft (122 m) with a center span of 170 ft (51.9 m) flanked on each side by two continuous spans of 57.5 ft (17.53 m) each. A 60 ft (18.3 m) central drop-in span of 33 in. (0.84 m) deep steel beams is supported by transverse concrete beams, supported in turn by structural steel wide-flange stays. Continuous longitudinal concrete beams comprise the remainder of the structure and receive support at their extremity, in the center span, from the transverse concrete beams and steel stays.4*8 In the more than half-century that has elapsed since Torroja’s Tempul Aqueduct, 2 1 cable-stayed bridges have been constructed (Table 9.1). Thirteen, or 62%, of these structures have been con-

strutted in the past decade. In the last five years nine have been completed, representing 43% of the total. Within the last three years the span of 1000 ft (300 m) has been exceeded, and a current design contemplates a span of 1300 ft (400 m). It has taken almost a quarter-century to reach a span contemplated by Prof. Morandi in his original design concept for the Lake Maracaibo Bridge. Be that as it may, it is obvious from the statistics that in recent years the concrete cable-stayed bridge has been accepted as a viable structure.

9.1.2 ADVANTAGES OF CONCRETE CABLE-STAYED BRIDGES

As engineers, we are aware that no particular concept or bridge type can suit all environments, considerations, problems, or site conditions. The selection of the proper type for a given site and set of circumstances must take into account many parameters. The choice of material, in addition to

Concrete Segmental Cable-Stayed Bridges

402 TABLE

9.2.

Concrete

Cable-Stayed

Bridges-Dimensional

Parameters

P)hl

Bridge I 2 3 4 5 6 7 8 9 10 I1 12 I3 14 15 16 17 18 19 20 21 22 23 24 25 26 27 28

Tempul Benton City Lake Maracaibo Dnieper River Canal du Centre Polcevera Viaduct Magliana Danish Great Belt” Danish Great Belt” Pretoria Barwon R i v e r Mount Street Wadi K u f River Foyle Mainbrticke Chaco/Corrientes River Waal Barranquilla Danube Canal Kwang Fu Pont de Brotonne Carpineto Pasco-Kennewick M-25 Overpass Ruck-A-Chuck) Dame Point East Huntington WeIrton-Steubenville

Stay Planes 2 2 2 2 2 2 2 3 2 2 2 1 2 1 2 2 2 2 2 2 1 2 2 2 2 2 2 2

No. stay Stays Arrangement 1 1 1 3 4 1 1 2 16 2 2 2 1 2 13 2 2 1 1 2 21 1 18 2 20 21 15116 24

Height Above Deck (f0

Pylon Heightto-Span Ratio’

14.1 139.4 95 65.6 148 111.5

0.07

315 41 43 49 177.5 360 172 155 151.8

0.27 0.44 0.16 0.42 0.19 0.52 0.38 0.19 0.17 -

52.5 -

0.15 -

Radiating Fan

231 94.75 220 71

0.22 0.16 0.22 0.39

Harp Radiating Radiating

302 279.4 333.2

0.23 0.31 0.41

Radiating Radiating

Radiating Harp Radiating Fan Harp Harp Radiating Radiating Radiating Fan

0.18 0.20 0.30 0.21 0.23

Deck Width cf.0

Girder Depth (ft)

6.9 57 5.87 59 79 51.75’ 46 15.8 6 15.75 42.5 98 101.5 47 101 37 51.8 67 63 41.3’ 79.8 39 54 105.75 41 103.5

SpantoDepth Ratio”

3.25 16.4 4.8 1.94 15 9.8-13.2 23.5 2.95 3 7 2 11.5-23 11.5 8.5 11.5 11.5 10 9.2

28.7 52.3 46.7 98.75 113 46 36 48 390 31 38.5 58.4 70 60 57 70 76 46 42.5

12.5 11.5 7 9 8.5 5-6 5 8.5

84 52 140 20 153 260 180 96.5

Girder Construction Typed CIP CIP CIP/PC d-i-s PC PC CIP/PC d-i-s CIP/PC d-i-s PC segments CIP segments CIP CIP CIP CIPiPC d-i-s PC segments CIP PC/CIP d-i-s PC and CIP CIP segments PC and CIP PC PC and CIP CIP PC segments CIP PC segments CIP and PC Composite Composite

“Design by White Young and Partners. hDesign by Ulrich Finsterwalder. ( See Table 9.1 for major span dimensions. “CIP = cast-in-place, PC = precast, d-i-s = drop-in-span. ’ Form hyperbolic paraboloid in space. ‘Per single-cell box. 1 ft = 0.305 m.

material properties, depends on availability and the prevailing economics at a particular time as well as the specific location of the site. The process of weighting and evaluating these parameters for various types of bridges under consideration is certainly more an art than a science. In evaluating a concrete cable-stayed bridge, the designer should be aware of the following advantages: 1.

The main girder can be very shallow with respect to the span. Span-to-girder-depth ratios vary from 45 to 100. With proper aerodynamic streamlining and multistays the deck structure can be slim, having span-to-depth ratios of 150 to 400, and not convey a massive visual impression.

2.

3.

4.

5.

Concrete deck structures, by virtue of their mass and because concrete has inherently favorable damping characteristics, are not as susceptible to aerodynamic vibrations. The horizontal component of cable-stay force, which causes compression with bending in the deck structure, favors a concrete deck structure. The stay forces produce a prestress force in the concrete, and concrete is at its best in compression. The amount of steel required in the stays is comparatively small. A proper choice of height of pylon with respect to span can yield an optimum solution.g Live-load deflections are small because of the live-load-to-dead-load ratio, and therefore

Introduction

6.

403

concrete cable-stayed bridges are applicable to railroad or mass-transit loadings. Erection of the superstructure and cable stays is relatively easy with today’s technology of prestressing, prefabrication; and seg&&tal cantilever construction.

9.1.3

STRUCTURAL STYLE AND ARRANGEMENT

Many of the concrete cable-stayed bridges have been designed by Morandi or have been strongly influenced by his style. Commencing with the Lake Maracaibo Bridge, of the 12 bridges constructed, excluding pedestrian and pipe bridges (see Table 9.1), six have been designed by Morandi, Figures 9.1 through 9.6. A third prize winner in the 1967 Danish Great Belt Bridge Competition was the Morandi-style design proposed by the English consulting firm of White Young and Partners, Figure 9.7. The ChacoKorrientes Bridge, Figure 9.8, very much resembles the Morandi style.

FIGURE 9.3. Magliana Viaduct (courtesy of L’Industria Italiana de1 Cemento).

FIGURE 9.4. Wadi Kuf Bridge, general consrruction ’ view (courtesy of Prof. R. Morandi).

FIGURE 9.1. Lake Maracaibo Bridge, general view, from reference 11 (courtesy of Julius Berger-Bauboag Aktiengesellschaft).

FIGURE

9.2.

Polcevera

Creek Bridge, general view.

FIGURE 9.5. Barranquilla Bridge (courtesy of L. A Garrido).

404


FIGURE 9.6. Carpineto Viaduct (courtesy of L’Industria Italiana de1 Cemento).

FIGURE 9.7. Danish Great Belt Bridge, artist’s ering (courtesy of White Young and Partners).

These structures, with the exception of the Magliana, Barranquilla, and Carpineto bridges, are typified by the A-frame pylon positioned in the plane of the stays and an auxiliary X frame or inclined struts to support the deck structure at the pylon. They are statically determinate systems so as to preclude any possible damage from differential settlements of the bridge piers and pylons or from light seismic shocks. A simple schematic of the structural scheme is shown in Figure 9.9, which consists of a series of independent balanced systems, each carried by an individual pier and pylon. These systems are then connected by drop-in girders, which are simple span girders spanning between independent systems.‘O The cantilever girder is supported at two points (C and D) by a pier system and elasticafly supported at two points (B and E) by the cable stays, thus producing a three-span girder with cantilevers on each side. The stays are supported by a pylon portal frame that is independent of the pier system supporting the girder. Another entry in the 1967 Danish Great Belt Competition by Ulrich Finsterwalder, of the German firm Dyckerhoff 8c Widmann, deviated from the Morandi style and was awarded a second prize. Finsterwalder’s design proposed a multiple-span, multistay system using Dywidag bars for the stays, Figure 9.10. The deck was envisioned as being constructed by the cast-in-place balanced cantilever

rend-

FIGURE 9.9. Schematic of Morandi-style structural scheme, from reference 10 (courtesy of the American Concrete Institute).

FIGURE 9.8. ChacoiCorr ientes Bridge, general \iew, from reference 13 (courtesy of Normer Gray).

FIGURE 9.10. Da&h Gre,lr Belt Bridge, .I1 list’\ rrndering (courtesy of Ulrich Finsterwalder).

Lake Maracaibo Bridge, Venezuela segmental method, each segment being supported by a set of stays. This concept was later to be consummated in the Main Bridge and in the design of the Dame Point Bridge. The choice of geometrical configuration and number of stays in a cable-stayed bridge system is subject to a wide variety of considerations. If cable stays are few, they result in large stay forces, which require massive anchorage systems. A relatively deep girder is required to span the large distance between stays, producing span-to-depth ratios varying from 45 to 100 (see Table 9.2). Depending upon the location of the longitudinal main girders with respect to the cable-stay planes, large transverse cross girders may be required to transfer the stay force to the main girder. A large number of cable stays, approaching a continuous supporting elastic media, simplifies the anchorage and distribution of forces to the girder and permits the use of a shallower girder, with span-to-depth ratio varying from 150 to 400 (see Table 9.2). The construction of the deck can be erected roadway-width by free cantilever methods from stay to stay without auxiliary methods or stays. If the depth of the roadway girder can be kept at a minimum, the deck becomes, more or less, the bottom chord of a large cantilevering truss; it needs almost no bending stiffness because the inclined stays do not allow any large deflections under concentrated loads.” In the 55 years since Torroja’s Tempul Aqueduct the concrete cable-stayed bridge has evolved from basically a statically determinate structure with one stay on each side of the pylon to a highly indeterminate system with multistays. As demonstrated by the Danish Great Belt Bridge Competition, the Pasco-Kennewick Bridge, and the Pont de Brotonne, spans of approximately 1000 ft (300 m) are practical and have been accomplished. The practicality of spans of 1300 ft (400 m) is demonstrated by the Dame Point Bridge, and spans approaching 1600 ft (500 m) are considered technically feasible. Leonhardt” has projected that with an aerodynamically shaped composite concrete and steel deck a span of 2300 ft (1500 m) can be achieved. With today’s technology of prefabrication, prestressing, and segmental cantilever construction, it is obvious that cablestayed bridges are extending the competitive span range of concrete bridges to dimensions that had previously been considered impossible and int.o a range that had previously been the domain of structural steel. This technological means exist; they only require implementation.

405

9.2 Lake Maracaibo Bridge, Venezuela This bridge, Figure 9.1, has a total length of 5.4 miles (8.7 km). Five main navigation openings consist of prestressed concrete cable-stayed structures with suspended spans totaling 771 ft (235 m). The cantilever span is supported on four parallel X frames, while the cable stays are supported on two A frames with a portal member at the top. There is no connection anywhere between the X and A frames, Figure 9.11. The continuous cantilever girder is a three-cell box girder 16.4 ft deep by 46.7 ft wide (5 m by 14.22 m). An axial prestress force is induced into the girder as a result of the horizontal component of cable force, thus, for the most part, only conventional reinforcement is required. Additional prestress tendons are required for negative moment above the X-frame support and the transverse cable-stay anchorage beams. l1 The pier cap consists of the three-cell box girder with the X frames continued up into the girder to act as transverse diaphragms, Figures 9.12 and 9.13. After completion of the pier, service girders were raised into position to be used in the construction of the cantilever arm. Owing to the additional moment, produced during this construction stage by the service girder and weight of the cantilever arm, additional concentric prestressing was required in the pier cap, Figure 9.13. To avoid overstressing of the X frames during this operation, temporary horizontal ties were installed and tensioned by hydraulic jacks, Figures 9.13 and 9.14.

FIGURE 9.11. Lake Maracaibo Bridge, pier cap with

X frames, from reference 11 (courtesy of Julius Berger-Bauboag Aktiengesellschaft).

IL

I

I

II

FIGURE 9.12. Lake Maracaibo Bridge, main span tower and X-f’rames, reference 11 (courtesy of Julius Berger-Bauboag Aktiengesellschaft).

I---,

Worklnq

Service qirder

’

\

\

/I I

from

I

I

for

FIGURE 9.13. Lake Maracaibo Bridge, pier cap of a main span and service girder, from reference 11 (courtesv of Julius Berger-Bauboag Aktiengesellshaft). 406

407

Wadi Kuf Bridge, Libya

FIGURE 9.14. Lake Maracaibo Bridge, brace mernbers bear against X frames after being tensioned by hy-

draulic ,jacks, from ref‘e rence 11 (courtesv of’ Julius Berger-Bauboag Aktiengesellschaft). In the construction of the cantilever arms, special steel trusses (service girders) were used for formwork. They were supported at one end by the completed pier cap and at the other end by auxiliary piers and foundations, as shown in Figure 9.15. The anchorages for the cable stays are located in a 73.8 ft (22.5 m) long prestressed inclined transverse girder. The reinforcing cages for these members were fabricated on shore in a position corresponding to the inclination of the stays. They

FIGURE 9.15. Lake Maracaibo Bridge, placing service girder for forming cantilever girders, from reference 11 (courtesy of Julius Berger-Bauboag Aktiengesellschaft).

weighed 60 tons and contained 70 prestressing tendons, Figure 9.16. The cable stays are housed in thick-walled steel pipes, Figure 9.1’7, which were welded to steel plates at their extremities and were encased in the anchorage beam. A special steel spreader beam was used to erect the fabricated cage in its proper orientation. The suspended spans are composed of four prestressed T sections. 9.3

Wadi

Kuf Bridge, Libya

The Wadi Kuf Bridge in Libya, designed by Prof. Morandi, consists of two independent balanced

FIGURE 9.16. Lake >fal-acaibo Bridge, fabrication of anchorage beam, from reierence 11 (courtesy of Julius Berger-Bauboag Aktiengesellschaft).

408

Concrete

Segmental

Cable-Stayed

Bridges

9.4 ChacolCorrientes Bridge, Argentina

FIGURE 9.17. L&e hla~.rc&o

BI idge, bousing for cable stays, from reference 11 (courtesy of Julius Berger-Bauboag Aktiengesellschaft).

cable-stay systems having their ends anchored to the abutment by a short hinge strut. The cable-stay systems are connected by a simply supported drop-in span, Figure 9.4. This structure consists of only three spans. The center span is 925 ft (280 m) long and the two end spans are each 320 ft (97.5 m), for a total length of 1565 ft (475 m). The simply supported drop-in center portion of the main span consists of three double-T beams 180 ft (55 m) in length; each beam weighs approximately 220 tons (200 mt).12 The A-frame towers are 459 ft and 400 ft (140 and 122 m) high and the roadway deck is 597 (182 m) above the lowest point of the valley beneath the structure. l2 The superstructure is a single-cell box girder that varies from 13 ft (4.0 m) to 23 ft (7.0 m) at the pylons. The single-cell box is 24 ft (7.4 m) wide and with cantilever flanges forms a 42.7 ft (13 m) deck. The contractor made good use of traveling forms to construct the box girder and deck, using the balanced cantilever technique to build on both sides of the pylons at the same time. Traveling forms were used because extreme height and difficult terrain made other conventional construction methods impossible or too costly. The deck was constructed by progressive cast-in-place segments, attached to the previously completed segments by means of temporary prestress ties and subsequent permanent post-tensioning Dywidag bars. The procedure adopted required temporary cable stays to support the cantilever arms during the construction sequence as the superstructure progressed in both directions from the pylon. When the superstructure extended sufficiently, the permanent stays were installed, and the structure was completed in the same manner.

The ChacoKorrientes Bridge (also referred to as the General Manuel Belgrano Bridge) crosses the Parana River between the provinces of Chaco and Corrientes in northeast Argentina and is an important link in one of the highways between Brazil and Argentina, Figure 9.8. It has a center navigation span of 803 ft 10 in. (245 m), side spans of 537 ft (163.7 m), and a number of 271 ft (82.6 m) approach spans on both the Chaco and Corrientes sides of the river. The vertical clearance in the main spans above flood level is 115 ft (35 m).i3.14 The superstructure of this bridge consists of two cast-in-place concrete A-frame pylons, which support a deck of precast segmental post-tensioned concrete. The pylons are flanked by concrete struts, which reduce the unsupported length of the deck, Figure 9.18. Although the pier cap section of the deck (between inclined struts) is cast in place, the cantilever portion consists of precast segments. The drop-in spans are cast in place. The deck structure consists of two longitudinal hollow boxes 8 ft 2% in. (2.5 m) wide and with a constant depth of 11 ft 6 in. (3.5 m), which support precast roadway deck elements, Figure 9.19. The precast girder elements were match-cast on the river bank in lengths of 13 ft 1% in. (4.0 m), with the exception of shorter units at the point of stay attachment, which contain an inclined transverse anchorage beam, Figure 9.20. Units were cast by the long-line method on a concrete foundation with the proper camber built in. Each unit was cast with three alignment keys, one in each web and one in the top flange. The units were erected as balanced cantilevers with respect to the pylon to minimize erection stresses. After a unit was hoisted, an epoxy joint material was placed over all of the butting area; then the unit was placed against the already erected unit and tensioned.13 To eliminate the need for falsework, the tnclmed struts and pylon legs were supported by horizontal ties at successive levels as construction proceeded, Figure 9.21. The legs were poured in segments by cantilevering the formwork from previously constructed segments. When deck level was reached, the girder section between the extremities of the inclined ties was cast on formwork. To further stiffen the pylon structure, a slab was cast between box girders at the level of the girder bottom flanges. This slab is within the limits of the cast-in-place box girders and inclined struts and serves as an additional element to accept the horizontal thrust from the cable stays. The upper portion of the pylon was

409

ChacolCorrientes Bridge, Argentina

k-

Precast construction --+I- Cast-irt-place 4 f Precast construction + 369 ft 1 in. (112.50 ml -4 369ft 1 in.(112.50m)

~803 ft 10 in (245.00 m) =I= Center span

537 ft 0 in. (163.70 rn)M Side span

FIGURE 9.18. Chaco/Corrientes Bridge, longitudinal geometry, from reference 14 (courtesy of Civil Engineering-ASCE).

8; I”. (22cm)

9 ft 2: in n

’

(2.80 m)

8: i n . I I ’ (22 cm1 nl

9 ft 2f in.

27 ft 3 in. (8.30 m) Cast-w-place concrete7 I

(2.8Om)

i in.

.’ ; 11 ft 5; in. Ii( 3 . 5 0’

/+8 ft 2f in.&-11 (2.50 m)

ft 3: in. (3.45 ml

11 ft 3: in. -& 8 ft 2f in.4 (3.45 m)

ml

(25-30 c m )

(2.50 m)

FIGURE 9.19. Chaco/Corrientes Bridge, deck cross section, from reference 14 (courtesy of Civil Engineering-AXE).

1 each rade of cable 4 ,n. anchor bolt

,. .,i.,...,.:,.. : ‘.. . . :.:..,;.;y.. ,; ,,. . Box girder -i: . . .! ‘. .:.-:. ,.. . . ,?.:..: .._ . ..*. .., . ,. :. 2, T.‘..

FIGURE 9.20. ChacoKorrientes Bridge, cable anchorage at girder, from reference 14 (courtesy of Civil Engineering-AXE).

then completed, using horizontal struts to brace the legs until they were connected at the apex, Figure 9.21.r3*14 The precast box girder units, with the exception of those at the cable-stay anchorage, were cast 13 ft 1% in. (4 m) in length by the long-line, match-cast procedure. The soffit bed of the casting form had the required camber built in. Alignment keys were cast into both webs and the top flange. Match casting and alignment keys were required to ensure a precise fit during erection. Each 44 ton (40 mt) unit was transported by barge to the construction site and erected by a traveling crane operating on the erected portion of the deck. Since each box was lifted by a balance beam, four heavy vertical bolts had to be cast into the top flange of each box. The lifting crane at deck level allowed longitudinal

410


2. 3. 4. 5. 6. 7. 8.

1

2

P---

FIGURE 9.21. ChacoiCorrientes Bridge, erection sequence of pylon, front reference 14 (courtesy of Civil Engineering-ASCE).

movement of the suspended box. Upon erection to the proper elevation, the unit was held to within 6 in. (150 mm) of the mating unit while epoxy joint material was’applied. Bearing surfaces of the unit were sand-blasted and water-soaked before erection. The water film was removed before erection and application of the epoxy joint material. The traveling deck crane held the unit in position against its mating unit until it could be post-tensioned into position. The crane was slacked off without waiting for the joint material to cure.13*14 To minimize overturning forces and stresses in the pylon, it was necessary to erect the precast box units by a balanced cantilever method on both sides of the centerline of the pylon. The erection schedule demanded simultaneous erection at each pylon, although the pylons are independent of each other. When four precast box units were erected in the cantilever on each side of the pylon, temporary stays were installed from the top of the pylon to their respective connections at deck level. After installation of the temporary stays, cantilever erection proceeded to the positions of the permanent stays, and the procedure was repeated to completion of the installation of the precast box units.13 The erection sequence may be outlined as follows: 1. Erect precast boxes and post-tension successively.

Erect diaphragms between lines of boxes and post-tension. Place temporary and permanent stays as erection proceeds. Remove temporary stays. Remove temporary post-tensioning in the cantilever sections. Place precast deck slabs between box girders. Concrete the three 65 ft 8 in. (20 m) drop-in spans. Place asphalt pavement, curbs, and railings.

9.5 Mainbriicke, Germany

The Main Bridge near Hoechst, a suburb of Frankfort, constructed in 1971 is a prestressed, cast-in-place, segmental, cable-stayed structure that connects the Fabwerke Hoechst’s chemical industrial complex on both sides of the River Main in West Germany, Figure 9.22. It carries two three-lane roads separated by a railway track and pipelines. This structure, a successor to Finsterwalder’s Danish Great Belt Bridge proposal, represents the first practical application of the Dywidag bar stay. l5 The bridge spans the river at a skew of 70” from the high northern bank to the southern bank, which is 23 ft (7 m) lower. The center navigation span is 486 ft (148.23 m) with a northern approach span of 86 ft (26.17m) and southern approach spans of 55, 84, 95, and 129 ft (16.91, 25.65, 29, and 39.35 m), Figure 9.23. Railroad track and pipelines are in the median between the two cantilever pylon shafts and are supported on an 8.7 ft (2.66 m) deep torsionally stiff box girder, Figure 9.24. The centerline of the

FIGURE 9.22. .1Ltinbt

nuke, from reference 16.

Mainbriicke, Germany

. . . .\~; _\ \ _ .LI~P .\\.~:

: FIGURE

411

9.23.

Mainbriicke, elevation and plan, from reference 16.

FIGURE 9.24. Ll;ti~~l~l-iicke,

cl-ass sections, f’rotl~

longitudinal webs of the box girder coincides with the centerline of the individual cantilever pylons, and they are 26.25 ft (8 m) apart. Transverse cross beams at 9.8 ft (3 m) centers form diaphragms for the box and cantilevers, which extend 39 ft (11.95 m) on one side and 36 ft (11 m) on the other side of the central box to support the two roadways, Figure 9.25. The cross section of the towers consists of an anchoring web in the center, sandwiched by two flatplate flange elements, Figure 9.26. In a transverse elevation of the pylons, the width of the pylon increases from the top to just below the transverse strut, where it decreases to accommodate clearance requirements for both modes of traffic, Figure 9.26. The stay cables (Dywidag bars) are in pairs, horizontal to each other in the main span and vertical in the side span, thus simplifying the anchorage detail at the pylon, Figure 9.26.1fi FIGURE 9.25. Mainbriicke, view of deck at pylon (courtesy of Richard Heinen).

rcf’evellcc\ l(i.

Ii]

VERANKERUNG DER SCHR;LiGSEILE IM PYLON

FIGURE 9.26. Mainbticke, pylon and cable configuration, from refer_

ence 16.

Construction of the bridge superstructure was by the cast-in-place segmental method, Figure 9.27. Segments in the river span were 20.7 ft (6.3 m) in length, corresponding to the spacing of the stays. Segments in the anchor span were 19 ft (5.8 m) in length. Segments in the anchor span were concreted before the corresponding segment in the river span to maintain stability. The pylon segments were associated with the superstructure segments, and each pylon segment was slipformed. Figure 9.28 shows the partially completed structure and the falsework necessary to install the stays. Each stay is composed of twenty-five 16 mm (5/s in.) diameter Dywidag bars encased in a metal duct, which is grouted for corrosion protection similar to post-tensioned prestressed concrete construction.

FIGURE 9.27, Mainbriicke, casting of deck segments (courtesy of Dyckerhoff & Widmann).

9.6

Tie1 Bridge, The Netherlands

The Tie1 Bridge, l7 Fi gures 9.29 and 9.30, crosses the Waal River, which, together with the Maas and the Rhine, flowing east to west, divides the

parCall\ FIGURE 9.28. Mainbticke, structure (courtesy of Richard Heinen).

complered

VIADUC

D’ACCF5

APPROACH VIADUCT

FIGURE

9.29.

‘l‘iel

OUVRAGE

Bridge,

PRINCIPAL

general

layout.

MAIN BRIDGE

414

FIGURE


9.30.

l‘iel Hriclgc,

main

splls.

Netherlands into northern and southern parts. This structure provides a needed traffic link between the town of Tie1 and the south of the country and is a major north-south route. The structure has an overall length of 4656 ft (1419 m) and consists of a 2644 ft (806 m) curved viaduct on a 19,685 ft (6000 m) radius, which includes ten continuous 258 ft (78.5 m) long spans and a 2008 ft (612 m) straight main structure comprising three stayed spans of 3 12, 876, and 3 12 ft (95, 267, and 95 m) and two 254 ft (77.5 m) side spans. The cross section consists of two precast concrete boxes, each supporting two vehicular and one bicycle lane. The total width of the superstructure, which is 89 ft (27.2 m) in the access viaduct, Figure 9.31, is enlarged to 103 ft (31.5 m) over the main structure so as to accommodate the pylon supporting the stays. The structure crosses not only the Waal River but also a flood plain, which is under water during the winter months. Navigation requirements dictate a horizontal clearance of 853 ft (260 m) and a vertical clearance of 30 ft (9.1 m).

The ten-span 2648 ft (806 m) long access viaduct is continuous over its entire length. The superstructure is supported on the piers by sliding teflon bearings, except at the three center piers where it is supported on neoprene bearings, having a thickness such that they ftx the viaduct at these piers. Expansion joints are located at piers 1 and 11. The superstructure in the access viaduct consists of two precast rectangular boxes of a constant depth of 11.5 ft (3.5 m) and width of 21 ft 8 in. (6.6 m). The top flange including cantilever overhangs has a width of 44 ft (13.44 m). The overall width of the approach viaduct deck is 89 ft 3 in. (27.2 m), including a longitudinal pour strip. The viaduct was constructed by the precast balanced cantilever method with cast-in-place closure pours at the midspans. To accommodate the cantilever compressive stresses in the bottom flange over the piers, the thickness of the bottom flange is linearly increased from a minimum of 8 in. (200 mm) to 24 in. (600 mm) over a length of 33 ft (10 m) on each side of the pier. Each pier segment contains a diaphragm. Because of the potential flooding of the river from April through December and the consequent loss or damage of falsework and loss of time, it was decided to build the access viaduct utilizing precast segments in the balanced or “free” cantilever construction. The segments could be cast during flooding and placed in storage. Erection of the segments, which would take less time than the casting, could be accomplished after the flood had subsided. The precast segments, weighing 132 tons (120 mt), were cast in movable forms on a casting bed having the length of one span (by the long-line method, see Section 11.6.2). Segments were stored by and parallel to the casting bed and handled by a 130 ft (40 m) span gantry crane, Figure 9.32. They were transported to the site (access viaduct abutment) by means of a 132 ton (120 mt) capacity trolley and then placed in the structure by the same gantry crane used in the precasting yard for handling, Figure 9.33; The trolley was used to transport the segments because the gantry was usually engaged in the precasting yard or in placing segments in the viaduct. The gantry crane was such that it spanned over the twin boxes in the superstructure and the trolleyway used to transport the segments. Segment joints are of the epoxy-bonded type (see Section 11.5). Cantilever imbalance is accommodated by a temporary support ad.jacent to the pier, Figure 9.33. Five temporary prestress bars


415

FIGURE 9.32. Precasting plant. (1) Casting bed, (2) I-e-bar storage, (3) segment storage, (4) concrete batch plant, (5) office, (6) gantry crane, (7) bridge approach.

are used as provisional prestressing to hold the segments in position until permanent prestress tendons can be threaded into the ducts and stressed. The symmetrical box girder main structure

consists of a 254 ft (77.5 m) side span, a 312 ft (95 m) side stayed span, and a 33 1 ft (101 m) section of stayed center span cantilevering toward the center of the bridge. The center section between the stayed cantilever ends is made up of four 213 ft (65 m) suspended lightweight concrete girders. Two alternatives were considered for the cablestay pylons: a single pylon located on the longitudinal centerline of the bridge or a portal-type pylon. To simplify the project, the portal-type pylon was selected. The portal pylon is fixed to the pier and passes freely through the superstructure, Figure 9.34. The superstructure is fixed at the pylon piers except for rotation. It is allowed to move longitudinally at succeeding piers. Two alternatives were also considered for the stay system: a multiple stay system supporting the deck almost continuously and a system consisting of a few large stays. As prestressed concrete stays had been selected, the second solution became somewhat mandatory. Construction of prestressed

FIGURE

9.33.

FIGURE

9.34.

Free passage of pylon through deck.

concrete stays is a costly operation requiring extensive high scaffolding, Figure 9.35; thus it is advantageous to reduce the number of stays. The short stays of the bridge have a slope of 1: 1 and the long stays a slope of 1:2. Their points of anchorage to the deck are respectively at 156 ft (47.5 m) and 3 12 ft (95 m) on both sides of a pylon. The long stays have a cross section of 3 by 3.3 ft (0.9 by 1 .O m) and are prestressed by 36 tendons on the bank side and by 40 tendons on the river side, because of the larger load on that side, Figure 9.36a. The effect of the different loads on the stays introduces a flexural moment into the pylon. The short stays have a cross section of 2.13 by 3.3 ft (0.65 by 1.0 m) and are prestressed by 16 tendons

Placing of segments by gantry crane.

416


Three loading conditions were considered for the stays from a statics point of view: 1.

2.

3. FIGURE 9.35. Falsework fbr stay construction.

on the bank side and 20 tendons on the river side, Figure 9.366. The concrete of the stays has a 2%day strength of approximately 8700 psi (60 MPa). Its function is not only to protect the tendons, but also to increase the rigidity of the stays, which is four times that of the tendons alone.

Long stays 40/36 cables E?!l

Short stays 20/16 cables 65

8,: (b)

t I

FIGURE 9.36. Cross section of stays.

For the self-weight of the stays and dead load of the superstructure, the deck is considered as supported on nonyielding supports, which are the stay anchorage points, and the load in the stays results from the reactions at these points. For design live load, the deck is considered as supported on yielding supports, the rigidity of which is determined by the rigidity of the prestressed stays. The prestress of the stays was calculated with a safety factor against cracking of 1 .l for dead load and 1.3 for live load, without allowing any tension in the concrete. The ultimate load safety factor is 1.8. For the load condition between cracking and collapse the stay rigidity is reduced to the rigidity of the tendons alone. Their excessive elongation, in case they yielded, would lead to an excessive deflection of the box girder and a premature collapse before the proposed safety limit. Therefore, it was necessary to reduce the initial stress of the tendons to 40 to 45% of their ultimate strength in order to keep them in the elastic range up to ultimate load determined by the safety factor of the structure as a whole.

The sag of the long stay is 2.3 ft (0.70 m) in a length of 328 ft (100 m) under dead load. Under live load the sag is reduced to 1.8 ft (0.55 m). The cross section of the stays at their extremities is increased slightly to resist bending stresses. These stresses were calculated by the method of finite differences. In the longitudinal direction the girders are prestressed primarily by the horizontal components of the stay forces. The unstayed end spans are prestressed with 54 tendons. In the other spans additional prestressing is provided by 10 tendons that overlap each other at the supports. These tendons were required until such time as the stay forces were applied and, at completion, to provide safety against cracking and collapse. The deck slab is prestressed transversely by tendons spaced at 12 to 17 in. (0.30 to 0.44 m). The suspended 213 ft (65 m) span is composed of four precast lightweight concrete girders with a 6500 psi (45 MPa) concrete. The cast-in-place deck slab is increased from a thickness of 9.8 in. (250 mm) in the box girders to 12.6 in. (320 mm), owing to the smaller restraint of the slab in the one web girders.

417


The following restraints and conditions were considered in the determination of the construction procedure for the main spans of the structure: The exclusion of falsework from the river because of ndvigation requirements. 2. The potential for flooding. 3. The presence of the precasting plant on the north bank. 4. The possibility of adjusting the attachment points of the stay to the deck.

PHASE 1

1.

Construction was executed in increments limited by the attachment points of the stays to the deck. The stays were prestressed progressively, by increasing the number of stressed tendons as the load in the stays increased. However, during certain construction phases when the load in the stays decreased, some of the tendons were detensioned or slacked off. Using the north side (access viaduct side) as an example, the construction was divided into the following phases, Figure 9.37: Phme 1:

Construction oj the outer spans-that is, the stay-supported side span andjanking span

a.

b. c.

Phase 2:

Superstructure from pier 11 to pier 12 and a 72 ft (22 m) cantilever into the next span Extension up to temporary support 12A Extension up to pier 13 with a 26 ft (8 m) cantilever into the center span; simultaneous construction of the pylon

Construction qf the$rst section over the river and the shortfbrestay.

Phase 3: Construction of the second section ouer the ri-iw a11d the long,fowstay.

The external spans on the north side were constructed on falsework during the dry season. Utilizing the precast plant on the north side, precast segments 16.7 ft (5.10 m) long weighing 132 tons (120 mt) were assembled on the falsework. Segments were joined by f in. (5 mm) cast-in-place joints. Placing of the segments was carried out by the same gantry crane as for the access viaduct. On the south bank, where there was no precasting plant, the external spans were cast in place on falsework. The cantilever river spans were built on 157 ft (48 m) long steel falsework, consisting of four 10 ft

PHASES DE L’OUVRAGE

CONSTRUCTION PRINCIPAL

DE

MAIN BRIDGE LONSTRUCTION P H A S E S

FIGURE 9.37. Main bridge construction phases.

(3 m) deep girders on 23 ft (7.10 m) centers. This falsework was suspended at one end by prestressing strands from the top of the pylons. At the lower end, the temporary support strands were anchored in a cross beam that supported the steel falsework by four 350 ton (315 mt) jacks. The 3 ft (1 .O m) stroke of the jacks allowed adjustment of the level of the suspension points, and the jacks were used also to release the temporary prestress suspension strands when the final stays were installed. At the opposite end, the steel falsework was hinged. The horizontal force component on these hinges was transmitted directly to the completed part of the deck, and the vertical component was taken by 1 in. (26 mm) bars. In Phase 3, the temporary stays were deflected by means of 95 ft (29 m) booms. This provided the advantage of maintaining the angles at the lower connection equal to that of Phase 2 and keeping approximately the same force level in the temporary stay. The falsework used in Phases 2 and 3 was carried on a barge; it was positioned by two derricks located on the completed part of the deck and by a floating crane. After the box girders were cast, the level of the falsework was adjusted, the last joint

418

Concrete

Segmental

was cast, and the concrete was prestressed. The next steps were constructing the stays, prestressing them, releasing the temporary stays, and removing the falsework. In order to reduce creep and shrinkage, the stays were made of 17 ft (5.15 m) long segments with protruding reinforcement and 16 in. (0.4 m) cast-in-place joints. The building of the falsework for the stays and the handling of the precast segments were carried out with the help of a 16 ton (15 mt) tower crane 2 13 ft (65 m) high, running on the deck. The precast 213 ft (65 m) suspended span girders weighed 468 tons (425 mt) and were transported by barge.

9.7 Pasco-Kennewick Bridge, U.S.A. The first cable-stayed bridge with a segmental concrete superstructure to be constructed in the United States is the Pasco-Kennewick Intercity Bridge crossing the Columbia River in the state of Washington, Figure 9.38. Construction began in August 1975 and was completed in May 1978. The overall length of this structure is 2503 ft (763 m). The center cable-stayed span is 981 ft (299 m), and the stayed flanking spans are 406.5 ft (124 m). The Pasco approach is a single span of 126 ft (38.4 m),

FIGURE 9.38. Pasco-Kennewick (courtesy of Arvid Grant).

Intercity

Bridge

Cable-Stayed

Bridges

while the Kennewick approach is one span at 124 ft (37.8 m) and three spans at 148 ft (45.1 m).4*15*18,1s The girder is continuous without expansion joints from abutment to abutment, being fixed at the Pasco (north) end and having an expansion joint at the Kennewick (south) abutment. The concrete bridge girder is of uniform cross section, of constant 7 ft (2 m) depth along its entire length and 79 ft 10 in. (24.3 m) width. The shallow girder and the long main spans are necessary in order to reduce roadway grades to a minimum, to provide the greatest possible navigation clearance below, and to reduce the number of piers in the 70 ft (21.3 m) deep river. The bridge is not symmetrical. The Pasco pylon is approximately 6 ft (1.8 m) shorter than the Kennewick pylon, and the girder has a 2000 ft (610 m) vertical curve that is not symmetrical with the main span. Therefore, the cable-stay pairs are not of equal length, the longest being 506.43 ft (154 m).‘s There is no attachment of the girder at the pylons, except for vertical neoprene-teflon bearings to accommodate transverse loads. The girder is supported only by the stay cables. There are, of course, vertical bearings at the approach piers and abutments. It is estimated that the natural frequency of the girder, where it will respond to dynamic acceleration (i.e., earthquake), is 2 cycles per second. If the situation occurs where the longitudinal acceleration exceeds this value, the vertical restraint at the Pasco (north) abutment is designed to fail in direct shear, thus changing the structure frequency to 0.1 cycles per second, which renders the system insensitive to dynamic excitation. The three main spans were assembled from precast, prestressed concrete segments, while the approach spans were cast in place on falsework, Figures 9.39 and 9.40. Deck segments were precast about 2 miles (3.2 km) downstream from the bridge site. Each segment weighs about 300 tons (272 mt) and is 27 ft (8.2 m) long, Figure 9.41. The segment has an 8 in. (0.2 m) thick roadway slab, supported by 9 in. (0.22 m) thick transverse beams on 9 ft (2.7 m) centers, and is joined along the exterior girder edges by a triangular box which serves the function of cable anchorage stress distribution through the girder body, Figure 9.42. 6 Each match-cast segment required approximately 145 yd3 (11 lm3) of concrete, continuously placed in a previously adopted sequence within six hours. After initial curing in the forms, the girder segments were wet cured for two weeks in the storage yard, air cured for an additional six months, prestressed transversely,

Brotonne

F I G U R E 9 . 3 9 . l’r~w~-K~~~~w~ I& Intu Lit\ HI ldge, precast segments in main spans (courtesy of Arvid Grant).

FIGURE 9.40. Paaco-Kenne\\ic 1, Intel city Bridge, appreach spans cast in place on falsework (courtesy of Walter Bryant, FHWA Region 10).

Bridge,

France

419

cleaned, repaired, completed, loaded on a barge, and transported to the structure site for installation in their final location. For possible unpredicted developments a shimming process was held in reserve for maintaining the assembled girder geometry correctness, but it was not used. There are no shims in the segmentally assembled, epoxy-joined prestressed concrete girder.‘“+18*‘” The sections were barged directly beneath their place in the bridge and hoisted into position, Figure 9.43. Fifty-eight precast bridge girder segments were required for the project. The stays are arranged in two parallel planes with 72 stays in each plane-that is, 18 stays on each side of a pylon in each plane. They are held at each p y l o n top, 180 ft (55 m) above the bridge roadway, in a steel weldment, Figure 9.44. Stay anchorages in the bridge deck are spaced at 27 ft (8.2 m) to correspond with the segment length. The stays are composed of + in. (6 mm) diameter parallel high-strength steel wires of the BBR type. The prefabricated stays, manufactured by The Preston Corporation, arrived on the job site on reels, Figure 9.45, and contained from 73 to 283 wires, depending upon their location in the structure. They were covered with a # in. (10 mm) thick polyethylene pipe, and after installation and final adjustment were protected against corrosion by pressure-injected cement grout. The outside diameter of the pipe covering varies from 5 to 7 in. (0.12 to 0.17 m). Design stress level for the stays is 109 ksi (751.5 MPa). Stay anchorages are of the epoxy-steel ball (HiAmp) fatigue type produced by The Preston Corporation. This structure was designed by Arvid Grant and Associates, Inc., of Olympia, Washington, in professional collaboration with Leonhardt and Andra of Stuttgart, Germany.

9.8 Brotonne Bridge, France

FIGURE 9.41. Pasco-Kennewick Intercity Bridge, precast segments in casting yard (courtesy of Arvid Grant).

The Pont de Brotonne, designed and built by Campenon Bernard of Paris, crosses the Seine River downstream from Rouen in France. Because of increased navigation traffic in the area, a second crossing over the Seine River was urgently needed between the two harbors of Le Havre and Rouen. The first one, the steel suspension bridge of Tancarville, was opened to traffic in 1959. The second, the Brotonne Bridge, the world’s largest cablestayed prestressed concrete bridge, was opened to traffic in June 1977. 2o A model of the structure is

I

22.50 m

I-

cl

CROSS

-SECTION

OF

7

CONCRETE

BRIDGE

/ eoprene sleeve

SECTION. ELEVATION B - B FIGURE 9.42. Pasco-Kennewick Intercity cables (courtesy of Prof. Fritz Leonhardt).

FIGURE 9.43. Pasco-Kennewick Intercity Bridge, erection of precast segments from barge (courtesy of Arvid Grant).

SECTION A - A Rridge, cross section and anchorage of sta?

FIGURE 9.44. I’asco-Kennewick lntercity Bridge, pylon and stay attachment steel weldment at top (courtesy of Arvid Grant).

Brotonne Bridge, France

FIGURE 9.45. l’asco-E;enne~~,ick prefr~bricated cable stay on reel.

Intcrcity

Bridge,

shown in Figure 9.46 and the general layout in Figures 9.47 and 9.48. The box girder carries four lanes and replaces ferry service between two major highways that run north and south of the Seine. Because large ships use this section of the river to approach the inland port of Rouen 22 miles (35 km) to the east, vertical navigation clearance is 164 ft (50 m) above water level, which results in a 6.5% grade for its longer approach.15*21 Total length of structure is 4194 ft (1,278.4 m), consisting of the main bridge and two approach viaducts. The main crossing has a span of 1050 ft (320 m). On the right bank, the transition between the main span and the ground is quite short because of a favorable topography where limestone strata slope upward to a relatively steep cliff. On the left bank, the terrain is flat and occupied by meadows. With an allowable maximum grade of 6.5% and a maximum height of fill of 50 ft (15 m), a nine-span viaduct was required to reach the main bridge. In a structural sense, the bridge is divided into two sections separated by an expansion joint at a point of contraflexure in the left-bank viaduct span adjacent to the cable-stayed side span, Figure 9.48.‘O

FIGURE 9.46.

Model of the Pant dc Rrotonne.

421

FIGURE 9.47. Artist’s rendering of the Pont de

Brotonne.

The prestressed segmental concrete deck consists of a single-cell trapezoidal box girder with interior stiffening struts, Figures 9.49 and 9.50. In the approach spans, web thickness is increased from 8 in. (200 mm) to 16 in. (400 mm) near the piers, and the bottom flange thickness is increased to a maximum thickness of 17 in. (430 mm). The only portion of the segment that was precast is its sloping webs, Figure 9.51, which were precast at the site. The other portions of the cross section, including top and bottom flanges, interior stiffening struts, and cable-stay anchorages (in the main structure only), were cast in place. Each segment is 9.8 ft (3 m) long. Extensive use of prestressing was made in the deck to provide adequate strength to this light structure. To resist the extreme shear stresses it was decided to place vertical prestressing in the webs. Pretensioned units were stressed on a casting bed, Figure 9.52, and equipped with specially designed button heads, thus producing a combination of pretensioning and anchorage plates. This system has the advantage of ensuring a perfect centering of the prestressing force together with a very rapid transfer of this force at both ends. Intensive rupture tests proved that an extremely high resistance to shear was created by this system.20 Finally, prestressing was also used as follows, Figure 9.53:20 1. Transversely in the top flange to provide flexural strength to the thin 8 in. (200 mm) slab. 2. In the inclined internal stiffeners, to accommodate tensile forces created by the transfer of loads from the box girder to the stays.

-+s_s_sO,s8>~_+

5850

1

5850

1 5850

4

5850

1

5850

i

5850

(5850

i

32000 69750 127840

FIGURE 9.48. General layout of Brotonne Bridge.

_ _ _ _ _

sl

b -’,, 1.50 (5’) .,

6.50

t I

I- 1.60 (5’) . 1.60 (5’) _,

6.50

1 1.50 (5’) * I

I

5.60

1

4.00

I

4.00

(13’)

(18’) FIGURE

9.49.

1

5.60

II

(18’)

(13’)

Cross section of Brotonne Bridge.

FIGURE 9.50. Interior view of deck, Brotonne Bridge.

FIGURE

coupler

for

9.51.

Precast webs, Hrotonnc

tensioning jack 36 am dia tension

Dyuidag -mat -..

FIGURE 9.52. Casting bed for pretensioned webs.

131 ItlgC.

distribution beam

FIGURE

9.53.

Various prestressing systems in the box girder.

3. Transversely in the bottom flange, to counteract tensile forces created by the stiffeners. 4. Longitudinally near the center of the main span, to allow for a reasonable margin of the order of 300 psi (2 MPa) of compressive stress in view of creep and secondary tensile stresses. Before erection of the superstructure, the bridge’s 12 approach piers were slip-formed, nine on the left bank and three on the right. The pier shafts have an octagonal curvilinear cross section inscribed inside a 13 by 29 ft (4.0 by 8.75 m) rectangle, Figure 9.54. The same section was used for all the approach-span piers, whose height varied from 40 to 160 ft (12 to 49 m). The shape of the piers did not substantially increase costs but did increase the aesthetic appeal of the piers. The piers bear through a reinforced concrete footing on four rectangular slurry trench walls used as piles with a maximum length of 60 ft (18 m), Figure 5.17. The pylon pier shafts also have an octagonal curvilinear shape inscribed inside a 30 ft (9.2 m) square to produce equal bending resistance about both principal axes. They are supported on foundation shafts having a diameter of 35 ft (10.86 m) with a maximum wall thickness of 6 ft 8 in. (2.03 m). The foundation shafts transfer the loads to a limestone stratum at a depth of 115 ft (35 m) below ground level. Foundation shafts were built inside a circular slurry trench wall, which was used as a cofferdam for dewatering.*” When slip-forming of the piers reached deck level, the piers were prestressed to their foundation so as to stabilize them for erection of the deck segments. As the precast deck units were erected,

!I/ *J

.

FIGURE 9.54. Pier and foundation of approach

spans.

wl

c 1.50 ,,

, 1 . 6 0 ,. 1 . 6 0 A _ (5’) - (5’)

6.50

*I (5’) T

6.50

(21’) ROADWAY

; 1.50 1

I

(d

I

FIGURE

9.49.

Cross section of Brotonne Bridge.

FIGURE 9 . 5 0 . I n t e r i o r v i e w o f d e c k , B r o t o n n e

FIGURE 9.51. Precast webs, Brotonnc

Hr~tlgc.

Bridge.

bulkhead . . ___-_- - steel i. forma -.

_-.--

coupler for ., tcnaion rods

___i..

tensioning __.a-__--_ 36 am dia tendon _ ._--.-

jack .--._ .rodr; .

dlatributiw ,-- ___.i- :’

-

bean

adjustable _-bracketa

11-36 IUS ..--Rwidag --.tenrion bars

i preatrcrsin~ _ _ _ __tendons __ ._-._

-- soffit ..__ -

FIGURE 9.52. Casting bed for pretensioned webs.

FIGURE 9.53.

Various prestressing systems in the box girder.

3. Transversely in the bottom flange, to counteract tensile forces created by the stiffeners. 4. Longitudinally near the center of the main span, to allow for a reasonable margin of the order of 300 psi (2 MPa) of compressive stress in view of creep and secondary tensile stresses. Before erection of the superstructure, the bridge’s 12 approach piers were slip-formed, nine on the left bank and three on the right. The pier shafts have an octagonal curvilinear cross section inscribed inside a 13 by 29 ft (4.0 by 8.75 m) rectangle, Figure 9.54. The same section was used for all the approach-span piers, whose height varied from 40 to 160 ft ( 12 to 49 m). The shape of the piers did not substantially increase costs but did increase the aesthetic appeal of the piers. The piers bear through a reinforced concrete footing on four rectangular slurry trench walls used as piles with a maximum length of 60 ft (18 m), Figure 5.17. The pylon pier shafts also have an octagonal curvilinear shape inscribed inside a 30 ft (9.2 m) square to produce equal bending resistance about both principal axes. They are supported on foundation shafts having a diameter of 35 ft (10.86 m) with a maximum wall thickness of 6 ft 8 in. (2.03 m). The foundation shafts transfer the loads to a limestone stratum at a depth of 115 ft (35 m) below ground level. Foundation shafts were built inside a circular slurry trench wall, which was used as a cofferdam for dewatering.2u When slip-forming of the piers reached deck level, the piers were prestressed to their foundation so as to stabilize them for erection of the deck segments. As the precast deck units were erected,

FIGURE 9.54. Pier and foundation of approach spans.

Bccticm D-D

Lalgitullim1bctic0A-A FIGURE 9.55. Half center span and pylon.

longitudinal

cocaectioaktweea

rection

Pyloncrdl Plcr

------c I c I __ . . . _ - 2 * l

w O

. N

i ._

” +

FIGURE 9.56.

.i.

-

.._..

Pzo.

Connection between pylon, deck, and pier.

_ . . -. . . . .

the pylon was constructed by conventional methods. Two single-shaft pylons carry a system of 21 stays located on the longitudinal axis of the structure, Figure 9.55. The reinforced concrete pylons required limited cross-sectional dimensions to preclude an unnecessary increase of the deck width while providing sufficient dimension to accommodate bending stresses from a transverse wind direction. Total pylon height above the deck is 23 1 ft (70.5 m). Construction of the pylon required Ieapfrog forms with 10 ft (3 m) lifts. An interesting feature is the total fixity of the pylon with the box girder deck. Because the bending capacity of the pylon pier and foundation had to be such as to accommodate unsymmetrical loads due to the cantilever construction, a decision was made to take advantage of this requirement in the final structure to reduce the effect of live load in the deck. Therefore, the pylon was constructed integral with the deck at its base, both pylon and deck being separated from the pier by a ring of neoprene bearings, Figure 9.56.20

FIGURE 9.57. C.;;rl,lc-SI;I>

mhor;1gc.

All deck loads are carried to the pylon piers by 21 stays on each pylon. Each stay consists of 39 to 60-0.6 in. (15 mm) strands encased in a steel pipe, which is grouted after final tensioning. Stay length varies from 275 to 1115 ft (84 to 340 m). Anchorage spacing of the stays at deck level is every 19.7 ft (6 m), every other segment, where the inclined stiffeners in the deck segments converge, Figures 9.53 and 9.57. A special deck anchorage block was designed to accommodate the variable number of strands in the stay as well as to allow full adjustment of the tension in the stays by a simple anchoring nut, Figure 9.58. The anchorage of the stays is such that it is possible at any time during the life of the structure to either readjust the tension in the stay or replace it without interrupting traffic on the bridge. Permanent jacks are incorporated into the anchorage, Figure 9.59, such that by tensioning the stay the adjusting nut can be slacked off. Stays are continuous through the pylon where they transfer load to the pylon by a steel saddle. The pipe wall thickness is increased near the anchorage points and near the pylon so as to improve fatigue resistance of the stays with regard to bending reversaIs.20 In constructing the deck girder, the operation was to extend the bottom flange form from a traveling form at the completed segment, placing the precast web units that form the basic shape and act as a guide for the remaining traveling form. After placement of the precast webs the interior steel form was jacked forward to cast the bottom flange struts and the top flange. Tower cranes at the pylon placed, as far as they could reach in both directions, the precast webs, Figure 9.60. Beyond the range of the tower cranes, gantry cranes running on rails on the top flange and extending 9.8 ft

FIGURE 9.58. Jacking of stay.

427

Danube Canal Bridge, Austria

FIGURE 9.59. Permanent stay anchorage.

FIGURE 9.60. llZain pier, pylon, anti deck during construction, from reference 20.

(3 m) beyond the end of the completed section were used to place new elements. The structure is shown at the start of main span construction in Figure 9.61, before closure of the main span in Figure 9.62, and completed in Figure 9.63.20

FIGURE 9.6 1. reference 20.

Stnrt of mnill \pdn Loll$tructiotl.

FIGURE 9.62. erence 20

&fore closure of mn111 SP;III, f 1oln Ed-

FIGURE 9.63. Aerial view of the Hroronne from reference 20.

9.9

flom

Hrldge,


This structure is located on the West Motorway (Vienna Airport Motorway) and crosses the Danube Canal at a skew of 45”. It has a 390 ft (119 m) center span and 182.7 ft (55.7 m) side spans,


428

182.7 ft

*

55.7 m

FIGURE 9.64.

canal.

182.7 ft

110m

55.7 m

I

Elevation of the Danube Canal Bridge.

Figure 9.64. It is unique because of its construction technique. Because construction was not allowed to interfere with navigation on the canal, the structure was built in two 360.8 ft (110 m) halves on each bank and parallel to the canal, Figure 9.65. Upon completion the two halves were rotated into

FIGURE 9.65.

390 ft

Construction of half-bridge on bank of

final position and a cast-in-place closure joint was made, Figures 9.66 through 9.69. In other words, each half was constructed as a one-time swing span. The bridge superstructure is a 5 1.8 ft (15.8 m) wide trapezoidal three-cell box girder, Figure 9.70. The central box was cast in 25 ft (7.6 m) long segments on falsework, Figure 9.7 1. After the precast inclined web segments were placed, Figure 9.72, the top slab was cast. Each half-structure has two cantilever pylons fixed in a heavily prestressed trapezoidal crosshead protruding under the deck with a two-point bearing on the pier, Figure 9.73. At the deck level the stays attach to steel brackets connected to prestressed crossbeams, Figures 9.74 and 9.75. Each stay consists of eight cables, two horizontal by four vertical. At the top of the pylons each cable is seated in a cast-iron saddle. The cable saddles are stacked four high, Figure 9.76, and are fixed to each other as well as to those in the adjacent plane. The cables were first laid out on the deck, fixed to a saddle, and then lifted by a crane for placement at the top of the pylon. The cables were then pulled

FIGURE 9.66. Plan of Danube Canal Bridge during construction and final state.


429

FIGURE 9.67. Danube Canal Bridge during rotation.

FIGURE 9.69. Closure joint, Danube Canal Bridge. FIGURE

9.68.

Ihnubc Canal Bridge during rotation.

FIGURE 9.70. Cross section, Danube Canal Bridge.

at each extremity by a winch rope to their attachment point at the deck level. During rotation of the two half-bridges, the deck and pylon sat on a bearing consisting of five epoxy-glued circular steel plates. The top plate was coated with teflon, sitting in turn on a reinforced concrete block that sat on a sand box. After rota-

tion the structure was lowered to permanent bearings by emptying the sand box. At the canal-bank end the deck had a concrete wall on its underside, bearing on a circular conCrete sliding track, Figure 9.77. The bearing between the wall and the track was effected by two concrete blocks clad with steel plates, under which

430


FIGURE 9.71. Con~tl

llctiotl OII

htnk, I),~nubc

C,~nnl

Bridge.

FIGURE 9.73. ‘Trapezoidal crosshead, Danube Canal

Bridge. FIGURE 9.72. P~t.c,~st

\jcbr, Danube Canal Bridge.

teflon-coated neoprene pads were introduced during the rotation movement (similar to the incremental launching method). The pivoting was accomplished by means of a jack pulling on a cable anchored in a block located near the sliding-track end. After rotation the two halves of the structure were connected by a cast-in-place closure joint, and The continuity tendons were placed and stressed.22 final structure is shown in Figure 9.78.

9.10 9.10.1

Notable Examples of Concepts PROPOSED GREAT BELT BRIDGE, DENMARK

The competition for a suitable bridge design in Denmark produced many new concepts and architectural styles. The design requirements specified three lanes for vehicular traffic in each direction and a single railway line in each direction.

FIGURE

9.74. Jacking of stays, Danube Canal Rridge.

The rail traffic was based on speeds of 100 mph (161 km/hr).23 Navigational requirements stipulated that the bridge deck be 220 ft (67 m) above water level, and the clear width of the channel was to be 1130 ft (345 m). A third prize winner in this competition was the Morandi-style design proposed by the English consulting firm of White Young and Partners, Figure

Notable Examples of Concepts

431

FIGURE 9.75. Cable-stay attachment, Danube Canal Bridge.

FIGURE 9.77. Circular concrete sliding track, Danube Canal Bridge.

FIGURE 9.78. Completed Danube Canal Bridge.

FIGURE 9.76. Stay saddles at pylon, Danube Canal Bridge.

9.7. This design embodied the principles of a cable-stayed bridge combined with conventional approaches of girders and piers with normal spans. The principal feature of this bridge design is the three-plane alignment of cable stays. This feature may become more important in urban areas, where trends in the future may dictate multimodal transportation requirements and an increase in the

number of automobile traffic lanes. The deck consists of two parallel single-cell prestressed concrete box’girder segments, Figure 9.79. The rail traffic is supported within the box on the bottom flange and the road traffic is carried on the surface of the top flange. The box girder contemplated a depth of 23.5 ft (7.2 m) and width of 27.75 ft (8.45 m) with the top flange cantilevered out 12 ft (3.7 m) on each side. The piers and towers were to be cast-in-place construction to support the deck segments, which were to be precast at various locations on shore and floated to the bridge site for erection. The maximum weight of a single box segment was estimated at 2200 tons (2000 mt). All segments of the superstructure were to be of reinforced and prestressed concrete. Up to this point in time, when the competition for this structt.re was conducted, all the concrete cable-stayed bridges had been either designed by

432

Concrete Segmental Cable-Stayed Bridges -e--- 3.675 m

8.45 m

b ’<

Morandi (Lake Maracaibo, Wadi Kuf, and so on) or strongly influenced by his style (Chaco/ Corrientes). They were typified, for the most part, by the transverse A-frame pylon with auxiliary X-frame support for the girder. However, an entry in the Danish Great Belt Competition by Ulrich Finsterwalder of the German firm of Dyckerhoff & Widmann deviated from this style and was awarded a second prize. Finsterwalder proposed a multiple span, multistay system using Dywidag bars for the stays, Figure 9.10. This proposal contemplated a spacing

between pylons of 1148 ft (350 m) and a spacing of the stays at deck level of 32.8 ft (10 m). Pylon height above water level was 520 ft (158.5 m). In a transverse cross section the deck was 146 (44.5 m) wide with two centrally located vertical stay planes 39 ft 4 in. (12 m) apart to accommodate the two rail traffic lanes, and three automobile traffic lanes in each direction outboard of the stay planes, Figure 9.80. The solid concrete deck had a thickness of 3 ft (0.9 m) in the transverse center portion, under the rail traffic, and tapered to a 1.3 ft (0.4 m) thickness -k! 58,M

22,50

f 0.00 i 175.w-j-350,oo / 175,oo

-J

‘,.

-_-_-_--------_-__- _______r____

/

r _ _ . r ____ - -----‘-‘-, -------, j 1 Ir;

I

Bahn

lu~+~ !

r ________,/__ ------

n

1

L.

‘-----:- -.---7 --__ -_,_ _ _ -q-i-’

(

~

‘C sam&

..I’ !

’ ON ----______ _ - - - - - - - - - - ------_-k

15,25 ------&2,00?c--- 10,00------+2,00MO

15,25 -

-

:o,llo -

FIGURE 9.80. Danish Great Belt Bridge, elevation and cross section (coultesv Dyckerhof’f’ & Widmann).

of

;p2,50

Notable Examples of Concepts

at the edges. The deck was to be constructed by the cast-in-place balanced cantilever segmental method, each segment being supported by a set of S&F s.

The proposed Dame Point Bridge over the St. Johns River in Jacksonville, Florida, as designed b) the firm of Howard Needles Tammen & Bergendoff, is a cable-staved structure with a concrete and a steel alternative. .4n artist’s rendering of the concrete cable-staved bridge alternative is shown in Figure 9.81. Navigation requirements dictate a 1250 ft (381 m) minimum horizontal opening and a vertical clearance of 152 ft (46.3 m) above mean high water at the centerline of the clear opening. I‘he proposed concrete cable-stayed main structure will have a 1300 ft (396 m) central

FIGURE 9.81. lhmc Point l3ritigc , artist’s rendering (cotll-tc’sv of’ Ho\vxcl Needles I‘a~nrncn ,Y- Bergendoff).

433

span with 650 ft (198 m) flanking spans. The layout of the main structure is shown in Figure 9.82.‘4 Structural arrangement of the bridge deck is shown in Figure 9.83. The bridge deck, which will carry three lanes of traffic in each direction, will span between longitudinal edge girders on each side. The longitudinal edge girder is in turn supported by a vertical plane of stays arranged in a harp configuration. The concrete deck and edge girders take local and overall bending from dead and live load in addition to the horizontal thrust from the stavs.25 The stav cables are anchored in massive vertical concrete pvlons, two at each main pier, which carry all loads to the foundations, Figure 9.84. In the center span, at each edge of the deck, the stavs are in a single plane spaced 30 in. (0.76 m) vertically, Figures 9.84 and 9.85. Stavs in the side spans, along each edge, are in two planes spaced 30 in. (0.76 m) transversely. Spacing of pairs of stavs along the edge beam is approximately 30 ft (9.1 m). Preliminary design contemplates 7 to 9 Dlwidag bars per stay, li in. (31.75 mm) in diameter, the number of bars per stay being a function of stress in the stay. The Dywidag bars are to be encased in a metal duct. During erection the fabricated length of duct is left uncoupled. After final adjustment the lengths of duct are coupled and pressuregrouted. Thus, the steel encasing tube will then be composite for live load and secondary dead load.‘” Construction proceeds bv conventional methods from the top of the pier bases at elevation 15.0 ft (4.6 m) to the level of the roadway at elevation 144.6 ft (44 m). At this point, a fixed formtable is secured and the first elements of the pylon and edge girders are cast. Erection of the deck is bv the

FIGURE 9.82. Dame Point Bridge, concrete cable-stayed alternative, from reference 23 (courtesy of Howard Needles ‘l‘ammen & Bergendoff).

105’-10” (32 3m)

Cast-in-situ

.Precast T B e a m

Beam

FIGURE 9.83. Dame Point Bridge, structural arrangement of bridge deck. from reference 24 (cowtesy of Howard Needles I‘ammen & Bergendoff).

Tower -.,

SIDE V I E W

FRONT VIEW

FIGURE 9.84. Dame Point Bridge, pylon arrangment, from reference 24 (courtesv of Howard Needles Tammen & Bergendoff). 434

;OMETRIC VIEW OF ERECTION SEQUENCE x *’

,’

15,

/’

/

PLATE 17

,? /

<,,de’,

15

436


balanced cantilever method. Two pairs of traveling forms are then used for sequential casting of 17.5 ft (5.3 m) lengths of edge girders on each side of the pylon. The bridge deck consists of single-T precast floor beams spanning between longitudinal edge girders and a cast-in-place topping. The precast T’s are pretensioned for erection loads. After erection the entire deck is post-tensioned to provide positive precompression between edge girders under all conditions of loading, Figure 9.85.24*25 A hinge expansion joint is provided at the centerline of the main span to allow for changes of superstructure length due to temperature, creep, and shrinkage. Similar joints are provided at the end piers, and link connections are used to prevent vertical movement of the superstructure. 9.10.3 PROPOSED RUCK-A-MUCKY BRIDGE, U.S.A.

The site for the proposed Ruck-A-Chucky Bridge designed by T. Y. Lin International, Figure 9.86, is approximately 10 miles (16 km) north of the proposed Auburn Dam and about 35 miles (56 km) northeast of Sacramento, California, crossing the middle fork of the American River. The river at this location is about 30 ft (9 m) deep and 100 ft (30.5 m) wide; however, upon impounding of the water behind the proposed dam, the river will become 450 ft (137 m) deep and 1100 ft (335 m) wide.26 In order to provide a 50 ft (15 m) vertical clearance above high reservoir water level, a bridge length of 1300 ft (396 m) will be required between the hillsides, which rise at a 40” angle from the

horizontal. Two existing roads parallel the canyon faces; a straight bridge across the river would require extensive cuts into the rock faces of the canyon to provide the necessary turning radius at the bridge approaches. This would be not only expensive but would also be damaging to the environment. Conventional piers in the river provide prohibitive design constraints, not only because of the 450 ft (137 m) water depth, but also because of the seismicity of the area. The hydroseismic (seiche effect) forces provide a formidable design load. After extensive studies, the proposed final solution was that of a hanging arc, Figures 9.87 and 9.88. The geometric configuration of this structure is such that the stays are tensioned to control the stresses and strains, in order to balance all the dead load with zero deflection; the curved girder carries the traffic and absorbs the horizontal component of the stays as axial compression. The stays are anchored on the slope according to the design formation to control the line of pressure in the girder. Thus, an ideal stress condition is achieved with almost no bending or torsional moments. After numerous studies and trade-offs a final radius of curvature was selected at 1500 ft (457 m).26 Two alternative designs have been prepared for this structure, one with a steel box girder and one with a lightweight concrete box girder. The concrete box girder, Figure 9.89, is fixed at the abutments and has no hinges or expansion joints in the 1300 ft (396 m) span. Depth of this box girder is 8.5 ft (2.6 m), so as to provide vertical stiffness and to distribute live load and construction loads on the deck to a sufficient number of adjoining cables. Stay anchorage at the girder is at 30 ft (9 m) intervals, based on construction and aesthetic considerations. 26

FIGURE 9.86. Ruck-A-Chucky ering (courtesy of T. Y. Lin).

Bridge, artiht‘s

rend-

/

) I LAY CAe‘ES

/ + L\

TAY CANES sourn 0U%5l0h

PEDESTALS SOUTH INSlOE

FIGURE 9.87. Kuck-A-Chucky Bridge, plan of bridge with concrete alternate, from reference 26.

.

ABtJWENT

FIGURE 9.88. Ruck-A-Chucky from reference 26.

Bridge, elevation of bridge with concrete alternative,

References 54’-0”

FIGURE 9 . 8 9 .

Ruck-A-Chuckv from reference 26

Bridge, cross section of concrete box girder alternative,

10.

References 1. A. Feige, ” The Evolution of German Cable-Stayed B r i d g e s - A n O v e r a l l S u r v e y , ” Acier-Stahl-Steel, N o . 12, December 1966 (reprinted in AISC Engineering Journal, July 1967). 2.

H. Thul, ‘Cable-Stayed Bridges in Germany,” Proceedings, Conzerence on Structural Steelwork, Institution oJ Civil Engineer.s, September 26 to 28, 1966, London.

3. W. Podolny, Jr., and J. F. Fleming, “Historical Development of Cable-Stayed Bridges,” Journal of the Structural Division, ASCE, Vol. 98, No. ST9, September 1972. 4. W. Podolny, Jr., and J. B. Scalzi, ‘Construction and Design of Cable-Stayed Bridges,” John Wiley & Sons, Inc., New York, 1976. 5.

M. S. Troitskp, “Cable-Stayed Bridges-Theory and Design,‘* Crosby Lockwood Staples, London, 1977.

6. F. Leonhardt,

11. .Anon.. The Bridge Spanning Lake ,21aracn1bo W i e s b a d e n . B e r l i n . Bauverlag ezurla, 1963.

B r i d g e i n .4rgentina,” 13. N. Gray, “ChacoiCorrientes Municipal Engineers Journal, Paper No. 380, Vol. 59, F o u r t h Q u a r t e r , 19i3. 14. H. B. Rothman, and F. K. Chang, “Longest PrecastConcrete Box-Girder Bridge in Western Hemisphere,” Civil Engineering, ASCE, March 1974. 15.

W. Podolny, Jr., “Concrete Cable-Stayed Bridges,” Transportation Research Record 665, Bridge Engineering, Vol. 2, Proceedings, Transportation Research Board Conference, September 25-27, 1978, St. Louis, M O . , National Academy of Sciences, Washington, D.C.

16.

H . S c h a m b e c k , “The Construction of the Main Bridge-Hoechst to the Design of the 365 m Span Rhein Bridge Dusseldorf-Flehe,” Cable-Stayed Bridges, Structural Engineering Series No. 4, June 1978, Bridge Division, Federal Highway Administration, Washington, D.C.

uf

8. H. M. Hadley, “Tied-Cantilever Bridge-Pioneer Structure in U.S.,” Civil Engineering, ASCE, January 1958. 9.

F. Leonhardt and W. Zellner, “Vergleiche zwischen Hangbrucken und Schragkabelbrucken fur Spannw.eiten iiber 600 111,” International Association for Bridge and Structural Engineering, Vol. 32, 1972.

in VU(;mbH..

12. Anon., “Longest Concrete Cable-Stayed Span Cant i l e v e r e d o v e r T o u g h T e r r a i n , ” Engineering ‘VewsRecord, July 15, 1971.

“Latest Developments of CableS t a y e d B r i d g e s f o r L o n g S p a n s , ” Saetryk Bygoningsstatzske Meddelelser, V o l . 4 5 , N o . 4 , 1 9 7 4 Denmark).

7. E. Torroja, Philosophy of Structures, English version by J. J. Polivka and Milos Polivka, University of California Pt-ess, Bet-kelev and Los Angeles, 1958.

R. Morandi, “Some Types of Tied Bridges in Prestressed Concrete,” First International Symposium, C o n c r e t e B r i d g e D e s i g n , AC1 Publication SP23, Paper 23-25, American Concrete Institute, Detroit, 1969.

17. A n o n . , “Tie1 Bridge,” Freyssinet Bulletin, March-April 1973. 18. Arvid

Grant,

International, STUP

“Pasco-Kennewick

Bridge-The

440

Concrete Segmental Cable-Stayed Bridges Longest Cable-Stayed Bridge in North America,” AXE, V o l . 4 7 , N o . 8 , A u g u s t 1977.

Freyssinet

19.

20.

Arvid Grant, “Intercity Bridge: A Concrete Ribbon o v e r t h e C o l u m b i a R i v e r , W a s h i n g t o n , ” CableStayed Bridges, Structural Engineering Series No. 4, June 1978, Bridge Division, Federal Highway Administration, Washington, D.C. C. Lenglet, “Brotonne Bridge: Longest Prestressed Concrete Cable Stayed Bridge,” Cable-Stayed Bridges, Structural Engineering Series No. 4, June 1978, Bridge Division, Federal Highway Administration, Washington, D.C.

2 1. Anon., “Cable-Stayed Bridge Goes to a Record with Hvbrid G i r d e r D e s i g n , ” Engznwring Nrw.+R~cod. October 28, 1976. 22. Anon.,

“The

Danube

Canal

Bridge

(Austria),”

International,

STUP

Bulletin,

May-June,

1975.

Civil Engineting,

23.

Anon., “Morandi-Stvle Design Allows Constant Susp e n d e d S p a n s , ” C’otr~wlting Enginret ( L o n d o n ) . March 1967.

2 4 . H . J . G r a h a m , “Dame Point Bridge,” Cable-Stayed Bridges, Structural Engineering Series No. 4, June 1978, Bridge Division, Federal Highway Administration, Washington, D.C. 2 5 . A n o n . , “Dame Point Bridge,” Design Report, Howard Needles Tammen SC Bergendoff, November 1976. 26. T. Y. Lin, Y. C. Yang, H. K. Lu, and C. M. Redfield, “ D e s i g n o f Ruck-A-Chucky Bridge,” Cable-Stayed Bridges, Structural Engineering Series No. 4, June 1978, Bridge Division, Federal Highway Administration, Washington, D.C.

10 Segmental Railway Bridges

10.1

INTRODUCTION TO PARTICULAR ASPECTS OF RAILWAY BRIDGES AND FIELD OF APPLICATION

10.8

10.2

LA VOULTE BRIDGE OVER THE RHONE RIVER, FRANCE

10.9 10.10

10.3

MORAND BRIDGE IN LYONS, FRANCE

10.4

CERGY PONTOISE BRIDGE NEAR PARIS, FRANCE MARNE LA VALLEE AND TORCY BRIDGES FOR THE NEW EXPRESS LINE NEAR PARIS, FRANCE

10.5 10.6

CLICHY FRANCE

RAILWAY BRIDGE NEAR PARIS,

10.7

OLIFANT’S

10.1

Zntroduction to Particular Aspects of Railway Bridges and Field of Application

10.11 RIVER BRIDGE, SOUTH AFRICA

Construction of segmental post-tensioned bridges for railway structures started in France in 1952 with a bridge crossing the Rhone River at La Voulte, Figure 10.1. It has been used extensively since that time in many countries. Precast segmental construction was introduced in railway structures in France with the Marne la Vallee Viaduct and in Japan with the Kakogawa Bridge, while incremental launching was adopted for several large railway crossings including the world’s longest bridge of this type: the Olifant’s River Bridge in South Africa (see Section 7.5). The major characteristic distinguishing railway bridges from highway bridges is the magnitude and application of loading. Live loading on a railway structure is two to four times larger than that applied to a highway bridge of comparable size. Every time a train crosses a railway bridge, the actual load applied to the structure is much closer to design live load than for a highway bridge, where even dense truck traffic usually represents only a

INCREMENTALLY LAUNCHED RAILWAY BRIDGES FOR THE HIGH-SPEED LINE, PARIS TO LYONS, FRANCE SEGMENTAL RAILWAY BRIDGES IN JAPAN SPECIAL DESIGN ASPECTS OF SEGMENTAL RAILWAY BRIDGES 10.10.1 Magnitude of Vertical Loads 10.10.2 Horizontal Forces 10.10.3 Bearings 10.10.4 Stray Currents 10.10.5 Durability of the Structure 10.10.6 Conclusion PROPOSED CONCEPTS FOR FUTURE SEGMENTAL RAILWAY BRIDGES

moderate proportion of the design load. Fatigue and durability of railway structures, therefore, are essential problems and need careful consideration, particularly in view of the fact that maintenance and repair of railway structures under permanent

FIGURE 10.1. La Voulte Bridge. view of the completed structure. 441

Segmental

Railway

Bridges

traffic is a very critical operation that can lead to unacceptable disturbance in a railway network.

10.2

La Voulte Bridge over the Rhone River, France

This first segmental prestressed concrete railway bridge is a notable structure and a landmark in the development of prestressed concrete. Constructed in 1952, it carries one railway track over the Rhone River near la Voulte, 80 miles (128 km) south of Lyons, in the southeastern part of France. The structure has five spans, each 164 ft (50 m) long. Each pier is made up of two inclined legs, and each span is an independent frame supported by an inclined leg at each end. Between the inclined legs on each pier, the deck is supported by a small beam resting on simple bearings. Construction proceeded using the cantilever scheme, with poured-in-place segments. The form travelers were supported by a temporary steel truss bridge, Figure 10.2. The cantilevers were built symmetrically in one span, the unbalanced moments being taken care of by temporary post-tensioning connecting the two inclined legs and the independent beam on one pier. The segments were 9 ft (2.75 m) long. The bending moments of each completed frame were adjusted by jacks placed at midspan and by continuity posttensioning tendons, Figure 10.3.

10.3

Morand Bridge in Lyons, France

This structure is a combined highway and masstransit bridge over the Rhone River in Lyons,

FIGURE 10.2. La Voulte Bridge, aerial view of the

deck

under

construction.

L,a Voulte Bridge, cantllcver- deck conprogress.

FIGURE 10.3.

struction

in

France’s third largest city. It is a three-span continuous structure with span lengths of 160, 292, and 160 ft (49, 89, and 49 m), resting on two river piers and two end abutments, which allow the transition of highway and railway traffic on both banks. The deck is made up of two parallel box girders carrying at the upper level three lanes of highway traffic including sidewalks. Inside each box girder is a railway track for the mass-transit system, Figure 10.4. This final scheme proved to be significantly less expensive and more efhcient in terms of the layout of the railway system than did the initial proposal, which contemplated a submerged tunnel for the railway crossing and a separate bridge for the highivay traffic. Dimensions of the structure in cross section are shown in Figure 10.5. The railway clearance of 13 ft 5 in. (4.12 m), including ballast and rail, calls for a 15 ft (4.56 m) structural height in excess of the normal requirements for a maximum span length of 292 ft (89 m). A constant-depth girder could thus be maintained throughout the river crossing except in the vicinity of the river piers, where short straight haunches allow the depth to be increased to 22 ft 7 in. (6.90 m). Over the piers a strong transverse diaphragm connects the two box girders, and the additional height over the pier allows the continuity of the diaphragm over the height of the haunch while the full clearance of the trains is maintained inside the box girders. The deck was built in balanced cantilever with 10 ft (3.0 m) long cast-in-place segments using one pair of travelers on a typical one-week cycle, Figures 10.6 and 10.7. Typical quantities of materials are as follows for the deck alone:

Mot-and Bridge in Lyons, France

FIGURE

ON PIER

FIGURE

10.5.

10.4.

Morand Bridge, perspective view of the structure.

AT MID-SPAN

Mm-and Bridge, typical cross section.

Deck area

Concrete Reinforcing steel Prestressing steel (longitudinal and transverse)

443

31,200 ft*

3,100 yd” 618,000 lb 256,000 lb

2,900 2,400 280,000 116,000

m2 m3 kg kg

FIGURE 10.6. Mm-and Bridge, construction of the superstructure.

Both concrete and reinforcing steel quantities far

exceed those required for a typical highway because of the very important increase of loads due to the railway lines in the box girders. The structure was completed and opened to traffic in 1977.

Segmental Railway Bridges single box carries the twin tracks, with the depth varying between 13.6 ft (4.15 m) and 17.9 ft (5.45 m) for the maximum span length of 280 ft (85 m) as shown in Figures 10.9 and 10.10. The segmental deck was cast in place, with travelers working in the conventional balanced cantilever fashion.

10.5

The extension of the Paris mass-transit system in the highly populated southeastern suburbs was the occasion for building a long elevated segmental prestressed concrete railway structure in a sensitive urban environment, Figure 10.11. This structure, located in the city of Marne la Vallee, includes a bridge over the Marne River and a long viaduct carrying two parallel railway tracks. Near the transition between the river bridge and the viaduct a passenger station is carried by the bridge structure. Three major considerations guided the choice of the structure:

FIGURE 10.7. MOI and Bridge, (onstl uction o f superstructure. Note pier segment for second parallel box girder. 10.4

Mane la Vallee and Torcy Bridges for the New Express Line near Paris, France

Cergy-Pontoise Bridge near Paris, France

A new railway line was completed in 1977 between Paris and the new satellite town of Cergy-Pontoise. A major prestressed concrete structure carries this line over several obstacles, including an interchange between two expressways (A-86 and A-14) and two branches of the Seine River. The trestle structures have a solid slab deck with spans varying between 65 ft (20 m) and 117 ft (35.60 m). Typical dimensions of the two main bridges over the Seine are shown in Figure 10.8. A

Maintain maximum clearance at ground level, not only to reduce the visual disturbance to the neighboring population, but also to allow all piers of the new structure to be fully compatible with the layout of all existing and future roads.

Elevation (a)

Typical cross section (b)

FIGURE 10.8. Cergy-Pontoise Bridge, dimensions. (a) Elevation. (b) Typical cross section.

Marne la Vallee and Torcy Bridges near Paris, France

FIGURE 10.9. Ccrg!-Pontoise str uctio n .

Bridge, cantilever con-

445

FIGURE 10.11. Marne la Vallee Bridge, ae~i,~l Liew of the completed structure.

Bridge, main span

FIGURE 10.12. Mane la Vailee Bridge, view of finished structure from ground level.

Produce a structure that is aesthetically pleasing when seen constantly from nearby.

In plan, the structure is laid out on a curve with a minimum radius of curvature of 1640 ft (500 m), Figure 10.11. Characteristic dimensions of the Marne la Vallee Viaduct are shown in Figures 10.13 and 10.14 and are summarized as follows:

FIGURE 10.10. Ckrgy-Ponroise closure.

Protect the neighboring population from unacceptable noise aggression. Basically, the structure is a single box of constant depth built of precast segments assembled by prestress into a continuous beam; the beam rests upon vertical piers provided with an architectural shape and regularly distributed at distances of 90 ft (27 m) to 120 ft (36 m), Figure 10.12. Both parallel tracks are laid on the transversely prestressed deck slab of the box girder and on a crushed-stone bed retained sideways by three continuous reinforced concrete walls. A central noise barrier separates the two opposite tracks and prevents the noise of a train riding one track to travel across to the other. At the edge of the concrete box girder, precast concrete panels manufactured with special white cement improve the appearance of the structure while providing the outside sound barriers.

1.

Bridge over the Marne River: a. Total length, 528 ft (161 m). b. Three-span continuous bridge with spans of 157,246, and 125 ft (48,75, and 38 m). c. Cross section: constant-depth box section with depth of 12.8 ft (3.90 m), web thickness varying from 20 to 35 in. (0.50 to 0.90 m) and bottom flange thickness from 7 in. (0.18 m) at midspan to 5 1 in. (1.30 m) over the river piers. Length of precast segments 5.6 ft (1.71 m). d. Two river piers are founded on largediameter bored piles and support the superstructure through special teflon bearings.

n I

t

ALLUVIUM

DEPOSITS

LIMESTONE

SAND FIGURE

446

10.13.

Mar-ne la Vallee Bridge, typical sections of deck and piers.

Marne la Vallee and Torcy Bridges near Paris, France

447

f. All bearings in the viaduct are standard laminated elastomeric pads.

ckm--

-

4~70 I I 00

12 CROSS

-

250 ~~~

12

MID SPAN CROSS SECTION

SECTION ON PIER

g. Piers are made of twin columns located under the webs of the box girder and connected at ground level by a common footing, which transfers the loads to deep slurry trenched walls anchored in limestone. The number and position of these bearing walls under each pier has been determined in relation to the magnitude of the transverse and longitudinal horizontal loads transferred by the superstructure, particularly in the curved portion of the viaduct.

(0)

(b) FIGURE 10.14. Xlarne la Vallee and .Torcy Viaduct, topical deck sections. (0) I\larne la Vallee trestle and I‘orq Viaduct cross section. (h) klarne la Vallee Bridge over the Llarne River. 2. Elevated viaduct: a . Total length, 4482 ft (1367 m). b. The viaduct is divided into 11 sections separated by expansion joints, allowing compatibility of thermal stress between the continuous welded rails and the concrete superstructure. The typical section is 412 ft (126 m) long with four spans of 88, 118, 118, and 88 ft (27, 36, 36, and 27 m). c. The two south viaduct sections adjacent to the main river crossing carry the passenger station and have shorter spans 69 and 92 ft (21 and 28 m). d.

Typical cross section is a single box carrying the two tracks with two main vertical webs 35 in. (0.90 m) thick and two sharply inclined facia webs used essentially for architectural purposes to reduce the apparent structural depth of the box and focus the eye on the high parapet wall.

e.

Average length of precast segments 7.5 ft (2.30 m).

The entire project was predicated on the use of precast segments with match casting and epoxy joints. A precasting yard on the south bank of the Marne, using four casting machines, produced the 690 segments with a maximum weight of 60 tons (55 mt) in eleven months. Segments were transported with a tire-mounted self-propelled carrier over the finished portion of the deck and placed in the structure with a launching gantry, Figure 10.15, in balanced cantilever. The gantry used on that project was that designed and built earlier for the B-3 Viaducts project. The gantry allowed all operations to be performed from the top in complete independence from the ground and all its related constraints. Placing of all segments was performed in a period of nine months between March and December of 1976, including the three spans of the main bridge and the fortv-four spans of the viaduct. The entire project was’completed in 24 months (including preparation of the final design), for a total deck area of 190,000 ft* (17,600 m’). Figure 10.16 shows

FIGURE 10.15. Marne la Vallee Bridge, precast segments placed with the launching gantry.

FIGURE 10.16. l\larne la \‘allee Bridge, crossing the Marne River and elevated passenger station.

FIGURE 10.18. Torcy Viaduct, segment transportation from Marne la Vallee to Torcy.

The total length of 1870 ft (570 m) is divided into three separate sections: one four-span unit, one nine-span unit, and one four-span unit.

FIGURE 10.17. Marne la Vallee Bridge, aerial view of the river crossing, a passenger station, and the elevated viaduct.

Precast segments were placed in the structure with an overhead launching gantry of a type slightly different from the one used previously, although calling on the same sequence of movements. Two parallel longitudinal trusses make the track for a transverse overhead portal crane carrying and placing the segments between the trusses. Figures 10.19 and 10.20 show the general view of the gantry and the detail of one segment placing. The overall view of the finished bridge appears in Figure 10.21.

the northern span of the river crossing and the elevated passenger station. Figure 10.17 is an aerial view of the overall project. In view of the success of this first application of segmental construction in urban railway elevated structures, the Paris Mass-Transit Authority decided to extend the same concept to construct another large structure a few miles eastward: the Torcy Viaduct. Fortunately, the precasting yard for the first bridge was still available and all segments could be manufactured there and trucked to the second bridge site, Figure 10.18. Dimensions of this new bridge are as follows: Cross section: exactly the same as for the Marne la Vallee elevated viaduct. Distribution of spans: 17 spans with typical span length of 115 ft (35 m).

FIGURE 10.19. ~l‘orcy Viaduct, precast segment placing with launching gantry.

Clichy

Railway

Bridge near Paris, France

FIGURE 10.20. I’orcy V i a d u c t , d e t a i l o f s e g m e n t handling between twin trusses of launching gantry.

FIGURE 10.21. structure.

10.6

449

vated metro. It crosses the Seine River adjacent to a new highway bridge between the cities of Clichy and Asnieres, as shown in Figure 10.22. Layout and principal dimensions appear in Figures 10.23 and 10.24. The prestressed concrete segmental structure is 1350 ft (412 m) long with a 280 ft (85 m) main span over the river with a deck of variable depth. The river piers of the two railway and highway bridges match exactly to minimize water flow and barge traffic disturbance. A provision is made for a second future highway bridge at the other side of the railway bridge, as seen clearly in Figure 10.25~. The restricted transverse clearances between the three structures and their corresponding traffic explains the special shape of the piers for the center railway bridge, which was carefully studied architecturally to enhance the appearance of the project. Foundations were very close to one another but could be maintained structurally independent to better control settlement and avoid vibration interference between bridges and in the ground. To carry the two railway tracks, the deck has a typical cross section consisting of three precast webs connected by a bottom slab, which forms essentially the compression flange over the piers, and an intermediate slab, which receives the ballast, Figures 10.25b and 10.26. The depression thus realized between the web top flange and the tracks has several advantages, including providing full safety against derailing on one track and reducing the noise level. Construction of the superstructure included match casting of all webs in a yard near the project site. The webs were placed in balanced cantilever with a light portal crane carried by the finished

Torcy Viaduct, view of the completed

Clichy Railway Bridge near Paris, France

At about the same time the two structures described above were built, a large and innovative railway bridge was constructed in the northeastern suburb of Paris for another extension of the reno-

FIGURE 10.22. Clichy Railway Bridge, view of the completed structure.

Segmental Railway Bridges

450

1 +o L

t

3

9 !i

sur

4

1

1

.

7

0

Lot 4 SW _ 76s~ .I

m

32

?i3

j+ rl,

FIGURE 10.23. Clichy Railway Bridge, layout and elevation of the structure. (n) Plan view. (h) Elevation.

cf!mfy 4 .

~-~ (kg

FIGURE

C-7)

10.24.

~_~~~ ~~-~~ ~ (rps) cps) 5ElNE

RIVER

P/a, (?&

6%fJ

m-g

Clichy Railway Bridge, main dimensions of segmental structure.

portion of the deck, Figure 10.27. Maximum weight of precast webs was 19 tons (17 mt), whereas segments that included the full three-web box (or even a more conventional single box for the equivalent span length) would have weighed in excess of 66 tons (60 mt). After assembly of precast webs with longitudinal post-tensioning, the two

twin slab sections were poured in place between the webs in balanced cantilever on very simple travelers. Web segments were 7.3 ft (2.22 m) long for the constant-depth part of the deck and 4.8 ft (1.48 m) for the variable-depth part. In fact, the slabs were cast in place between the three webs in two or three increments of that length respectively (a length of

Clichy

Railway Bridge near Paris, France

451

LEGENDE

(a)

FIGURE 10.25. Clichy Railway Bridge, typical sections of piers and deck. (a) Elevation of land and river

piers. (b) Dimensions of the deck cross section.

14.6 or 4.44 m) to reduce the number of site operations. A three-day cycle of operations could be constantly maintained, including some overtime work for the larger segments near the river piers. Overall, construction in cantilever of the total superstructure took one year between September 1977 and September 1978. A special design aspect, specific to railway bridges, was the transfer of horizontal loads (in-

FIGURE 10.26. Clichv Railway Bridge, pier segnrent

and cantilever constr&ion.

duced through braking or starting of the trains over the bridge), to the piers and foundations. A single fixed bearing was provided over pier P6, the foundation of which was designed to transfer to


452

the need arise in the future. Two families of tendons could be added: Above the lower slab in the two voids of the box section, anchors being provided in blisters already built in the structure. Atop the center precast web and on the outside face of the two facia webs, anchor blocks and deviation saddles being prestressed by high-strength bolts to the precast webs.

FIGURE 10.27. Clichy Kailway Hrtdge, cast webs for cantilever construction.

plar~ng

pre-

the limestone stratum the total maximum horizontal load of 660 tons (600 mt) applied to the bridge. There are three pot bearings between the deck and the pier shaft, each capable of safely transferring half of the maximum horizontal load. Each bearing can thus be changed under traffic without reducing the capacity of the structure. Special provisions were also included at the design stage, Figure 10.28, to allow additional prestressing to be incorporated in the structure should

The large precast architectural panels on both sides of the deck could be temporarily removed to allow this work of additional prestresstng to be performed. Upon completion, all additional tendons would be fully protected and concealed behind the panels. The new line has been open to traffic since May 1980, and the first months of operation confirm that the precautions taken to reduce noise and vibration disturbance through welded continuous rails and sound-barrier panels make such elevated structures an acceptable solution for mass-transit lines in urbanized areas. 10.7

fb)

FIGURE 10.28. Clichy

Railway Bridge. (a) View of ad.jacent highway and railway bridges crossing the Seine River. (b) Provisions for future additional p&tress.

Olifant’s River Bridge, South

Africa

This structure is part of a line carrying iron ore on special heavy trains 7500 ft (2300 m) long made up of 200 cars with a total weight of 19,000 tons (17,000 mt) to connect the Sishen mines to the harbour of Saldanha 110 miles north of Capetown. Olifant’s River Viaduct is today the world’s longest incrementally launched prestressed concrete structure (refer to Chapter 7) with a total length of 3400 ft (1035 m) and 23 spans of 148 ft (45 m), Figure 10.29. Shown in cross section in Figure 10.29, the single box girder deck accommodates only one track on ballast. The equivalent uniform live load of the 33 ton (30 mt) axles is 7.1 kips/lineal ft (10.5 mt/lm), which is increased by an impact factor of 1.29. The 23 spans are divided into two 1 l-span sections, each anchored to the end abutment, and one single transition span at the center. This scheme allows all horizontal loads to be transferred to the abutments. The maximum horizontal reaction including all thermal effects is in excess of 1200 tons (1100 mt). The piers, which vary between 80 ft (25 m) and 150 ft (55 m) in height, are extremely flexible and do not, therefore, have an important effect on the horizontal restraint of the structure, except during construction. The pier shafts have an I-

Incrementally

I

L

3.10 5 50m

Launched

Railway

Bridges

453

51nl --L

;I c -I

(b)

FIGURE 10.29. Olifant’s River Viaduct. (a) General

view of the structure. (b) ‘T‘ypical

cross section.

shaped cross section with longitudinally tapered faces. Neoprene bearings are used for the piers near the abutments and teflon sliding bearings in the center of the structure. The deck was entirely constructed behind one abutment (see schematic view in Figure 7.29) and incrementally launched in one direction. Construction time for the superstructure was nine months, with a theoretical cycle of 10 working days for a typical 148 ft (45 m) span realized after 10 spans; it was further reduced to seven days with two shifts toward the end of the project. The total weight of the superstructure of 14,500 tons (13,000 mt) called for two 200 ton jacks for the push-out operations in increments of 3.5 ft (1 .OO m). A 60 ft (18 m) long launching nose was used in front of the first span to reduce the variation of bending stresses in the superstructure during the successive stages of construction, Figure 10.30. The bridge nearing completion is shown in Figure 10.3 1; it was opened to iron ore trains in 1976.

10.8 Incrementally Launched Railway Bridges for the High-Speed Line, Paris to Lyons, France

FIGURE 10.30. Olifant’s River Viaduct, launching nose reaching beyond a high pier.

build some new very-high-speed train lines (safe maximum speed of 200 mph or 320 km/hr) and started the construction of the first such line between Paris and Lyons, which included an entirely new structure over a distance of 250 miles (400 km) with proper connections to the existing metropolitan track and station system. The new project required 400 bridges including nine large viaducts, such as the structure shown in Figure 10.32. A very comprehensive optimization study followed, and a set of guidelines and structural standards were prepared for the French National Railways by a team of engineers headed by one of the authors. Results of the preliminary investigations and of this optimization study can be summarized as follows: 1.

To meet increased competition by domestic airlines, the French National Railways decided to

Track alignment is chosen to keep the curvature in plan more than 10,500 ft (3200 m) and preferably more than 13,000 ft (4000 m). The

4.

5.

6.

FIGURE 10.31. Olifant’s River Viaduct, view of the structure nearing completion.

FIGURE 10.32. Railwq

Viaducts for Paris-Lyons high-speed line, view of the viaduct over the Saone River.

corresponding cross fall between rails is 7 in. (180 mm). 2. All rails are to be continuously welded and placed on a ballast bed with a minimum thickness of 14 in. (0.35 m). 3. Maximum rigidity of the structures is obtained by using a continuous box section with slenderness ratio of l/14. The corresponding

maximum deflection under design load is therefore l/2000 of the span, whereas conventional specifications for normal-speed lines allow up to l/800. Adopt as much as possible single box girder decks for the two parallel tracks with minimum web thickness of 14 in. (0.35 m) and a minimum top slab thickness of 10 in. (0.25 m). The optimum span length is between 150 and 170 ft (45 to 50 m), which leaves the construction method open to various solutions (cantilever, span-by-span or incremental launching). The horizontal loads should be transferred to one abutment equipped with a special fixed bearing, allowing all piers to be relieved of any appreciable longitudinal bending. A typical H-section was adopted as the most appropriate except for certain specific locations where a box section might be required.

Because many of the viaducts were located in environmentally sensitive areas, an overall architectural study was also conducted to establish a unity of appearance for all bridges in terms of the shapes of deck and piers, parapet or guard rails, abutments and approach fills. Of the nine viaducts, two were finally constructed with conventional methods and the remaining seven were incrementally launched. This method proved economical in view of the moderate span lengths, the depth of the box section available, and especially because the superimposed dead and live loads were so much more important than for a highway bridge that the increased dead-load moments during construction were in proportion of much less significance. Table 10.1 gives the essential characteristics of the seven segmental viaducts, including principal quantities of materials for the superstructure. Elevations of five bridges appear in Figure 10.33. As an example of the construction method, some details are given for the bridge over the Saone River, where a launching nose 93 ft (28.50 m) long and weighing 71 tons (65 mt) was used in front of the first span to reduce the stress variations in the superstructure during launching, Figure 10.34. The bridge superstructure was cast in successive increments in a fixed location behind the right bank abutment in the length of a half-span, Figure 10.35. A typical sequence of operations is shown schematically in Figure 10.36. Each superstructure section is in fact cast in two stages

Table 10.1.

Characteristics of Segmental Viaducts on the Paris-Lyons High-Speed

Line

Quantities of Deck

Dimensions of Deck [ft or in. (m))

Bridge Location and Span Lengths

[ft @)I" @J Saulieu 115-3@ 144- 115 (35 - 3 @ 44 - 35) Q Serein 115-3@ 1 4 4 - 1 1 5 (35 - 3 @ 44 - 35) 0 SamRium 155 ~ 5 (a: 164 - 137 (47.2 - 5 6 50 - 41.8) @ figtine 109-8@ 144- 1 0 9 (33.4 - 8 @ 44 - 33.4) @I Rochc 108-7@149-108 (33.1 - 7 I@ 45.5 - 33.1) 0 Seine River 114-ZOl- 114 (34.8 - 61.4 - 34.8) Q cmler Cad (length) 85 - 105 - 85 (26 - 32 - 27)

Flange Thick. Bridge Length

Total Width (ft)

Height

BOX Width

(W

(f0

Web Thick. (in.)

662 (202)

41.0 (12.50)

10.8 (3.30)

18.0 (5.50)

662

w-w

41.0 (12.50)

10.8 (3.30)

1112 (339)

40.3 (12.30)

1370

C0ncr.

H.T. Steel

Reinf. Steel [Ib/yd3 (kg/&)1

Year Cornpleted

240 (140)

1978

Pier Height lft WI

[ftvf?

[Ib/yd3

(in.)

Bottom (in.)

WWI

(kg/m?1

18/49 (0.45/1.25)

11 (0.275)

IO (0.25)

461121 (14137)

2.52 (0.77)

(46)

18.0 (5.50)

18/49 (0.45/1.25)

11 (0.275)

IO (0.25)

661148 (20/45)

2.52 (0.77)

78

240

(46)

(140)

1979

11.5 (3.51)

18.0 (5.50)

20 (0.50)

12.5 (0.32)

12.5 (0.32)

46 (14)

2.46 (0.75)

84 (50)

210 (125)

1979

39.0 (11.90)

10.8 (3.30)

18.0 (5.50)

24 (0.60)

12.5 (0.32)

14 (0.35)

43/105 ( 13/32)

2.30 (0.70)

84 (50)

250 (150)

1978

39.0 (11.90)

10.8 (3.30)

18.0 (5.50)

24 (0.60)

12.5 (0.32)

14 (0.35)

43/l 15 (I 3135)

2.30 (0.70)

84

(385)

(50)

250 (150)

1978

Long. grade: 0.55% Straight in plan

429 (131)

41.0 (12.50)

13.1 (4.00)

19.0 (5.80)

24135 (0.60/0.90)

11 (0.28)

12120 (0.30/0.50)

36 (11)

84 (50)

190 (110)

1980

Long. grade: 0.2% Straight in plan

279

(85)

40.0 (12.10)

7.8 (2.37)

19.0 (5.80)

12 (0.30)

10 (0.25)

8 (0.20)

43 (13)

-

Bridge Layout hong. grade: 1.3% Radius in plan: 20,000 ft (6000 m) Lrmg. grade: 0.95% Radius in plan: 26,000 ft (8000 m) Circular profile in elevation: R = 130,000 ft (40,000 m) Long. grade: 2.5% Straight in plan Long. grade: 3.5% Radius in plan: 10,600 ft (3250 m)

(419) 1260

“Structures are numbered with increasing numbers from Paris to Lyons.

TOP

78

1978

Viaduc du Serrein

FIGURE

456

10.33.

Elevation of five segmental bridges for Paris-Lyons line.

Segmental Railway Bridges in Japan

457

(bottom slab during the first stage, webs and top slab during the second stage). The typical construction cycle allowed casting a half-span every week-that is, constructing two spans per month. The launching operation proper called for a very efficient system, developed and perfected previously in Germany, including under each web of the box section: One vertical jack with sliding plate Two coupled horizontal jacks for actual launching, allowing movements in 3 ft increments FIGURE 10.34. S;IO~C Kivcl- Bridge, approachtng pier.

launching nose

Typically, launching of an 80 ft section took three to three and a half hours, despite the large weight of the concrete superstructure, reaching 9000 tons (8000 mt) at the end of construction. Figure 10.37 shows a completed structure, and Figure 10.38 shows another aspect of the construction of these seven viaducts. 10.9

Segmental Railway Bridges in Japan

Many railroad bridges have been built in Japan using the segmental construction technique. The sketches shown in Figures 10.39 through 10.42 depict the elevation and the cross section of the following cast-in-place segmental bridges: FIGURE 10.35. Saone Kivcr Bridge, xrial 1 ie\\ with casting vard in behind abutment in foreground.

SLIDING

TEMPORARY SUPPORT F O R PUSHN;

PACE

*

.~~

-

-

-

-

Kyobashigawa Bridge Natorigawa Bridge

APPROACH _^ -wAN~PRECAST Y A R D

e

-~

SITUATION DURING FABRICATION OF SEGMENT 7 ,2r550~25~0

7Q5 0 0-%!!L

SITUATION AFTER PUSHING OF SEGMENT 7 GENERAL PRINCIPLE OF THE CONSTRUCTION METHOD BY PUSHING FIGURE 16.36. Saone River Bridge, typical construction stages of incremental launching.

FIGURE

FIGURE 10.38. Digoine ing over high piers.

Kisogawa Ashidagawa

10.37.

Saone River Bridge, view of‘ the completed structure.

Bridge, incremental launch-

Bridge Bridge

Figure 10.43 shows the Kakogawa Bridge during construction. The superstructure is made of twin constant-depth box girders, one box girder carrying one railway track. The total length of the bridge is 1640 ft (500 m), with typical span length of 180 ft (55 m). Each box is 13 ft (4 m) wide and 11.5 ft (.3.50 m) deep. The precast segments were handled by a launching gantry and assembled by longitudinal post-tensioning tendons. The erection used the balanced cantilever system. The most outstanding prestressed concrete railway structure, however, is the Akayagawa arch

bridge shown in Figure 10.44. Total length is 980 ft (298 m) and the center arch span is 410 ft (126 m). The 13 ft (4.00 m) deep box girder carrying two railway tracks is continued throughout between abutments and rests over the center gorge on a very flat arch rib through ten spandrel columns. The respective proportions are such that the deck carries all bending moments and the arch rib carries the normal load induced by its curvature. The erection scheme was unique and called for cantilever construction starting from both sides. A very strong back stay made up of a prestressed concrete member with a prestress force of 5300 tons (4800 mt) was installed diagonally between the top of the main transition piers between the arch structure and the approaches on one hand, and the foundation of the adjacent piers in the approach structures on the other hand. While erection progressed, high-strength steel bars were placed diagonally between the vertical members, forming a temporary truss structure until the crown was reached from both ends. Control of tensioning of those steel bars was very critical and complicated. Finally, all steel bars and the two temporary back stays were removed after closure of the arch at midspan. 10.10

Special Design Aspects of Segmental Railway Bridges

10.10.1 MAGNITUDE OF VERTICAL LOADS

Most bridges carry tracks laid on ballast with a minimum thickness of 10 to 14 in. (0.25 to 0.35 m).

459

Special Design Aspects of Segmental Railway Bridges SHIN

OSAKA - L 33.90

199

8Om

HAKATA 66.00

66.00

33.90

ELEVATION

p-s- 61 51

/ j ---+I 2 so 2.70 m IO' 90592

24,

i

CROSS SECTION FIGURE 10.39. Hyobashigalva

Live loading used in design of railway bridges varies between countries-Cooper loading for Anglo-Saxon countries, new UIC loading for most European countries-and also according to the nature of the structure: mass-transit lines are usually designed for much lighter loads than normal train lines. The heaviest loadings are for ore freight trains. To exemplify the basic difference between a highway and a railway bridge, Figure 10.45 compares a typical 150 ft span and a 36 ft wide deck normally designed for three highway lanes of traffic or two railway tracks. The total superimposed dead and live load is 3.6 times greater for the railway bridge. In addition, the weight of ballast (representing 40% of the total load) must be considered as a live load to cover the cases where the ballast is removed from the deck or has not yet been placed on a new bridge.

Bridge, Japan

10.10.2 HORIZONTAL FORCES

Railway bridges have to carry very important horizontal forces, between five and ten times the horizontal forces carried by a highway bridge of similar size. The standard current practice for long viaducts is to have a fixed bearing on one abutment if the bridge length is less than 1500 ft (450 m), and on both abutments and on intermediate special bents if it is greater. The order of magnitude of this horizontal force on the abutments carrying the fixed bearings is often 1000 tons for a two-track viaduct. The various forces involved are described below: Longitudinal Forces

Braking and acceleration forces


460

/

TOKYO

MORIOKA ,A

524.90 m

ELEVATION

CROSS SECTION FIGURE 10.40. Natorigawa Bridge, Japan.

Forces due to box girder deformations: creep, shrinkage, and temperature variations Loads induced by the length variations of long welded rails under temperature variations Longitudinal component of wind forces Braking and acceleration forces are oneseventh of the total weight of live loads, with a ceiling of 285 tons for braking and 53 tons for acceleration (French regulations). Forces due to longitudinal deformations of the box girder vary because of creep, shrinkage, and temperature variations. The bearing displacements induce horizontal loads by distortion or friction. Length variations of the long welded rails due to temperature variations create a horizontal force parallel to the rail. This force can be estimated at 50 tons per rail (length of rail more than 100 meters). For a two-track bridge it is 2 x 2 x 50 = 200 tons.

Longitudinal component of wind forces are described in the AASHTO specifications for bridges. Transuerse Horizontal Forces Centrifugal horizorltal force can be very important f-or high-speed trains.

For the 200 mph train from Paris to Lyons this force is more than 400 tons for some viaducts 1200 ft (380 m) long with two tracks and radius of curvature of 10,500 ft (3200 m). The lateral acceleration is more than 20% of that of gravity. Transverse wind force is described in the AASHTO standards (50 lb/f?). 10.10.3

BEARINGS

In order to gain complete control of these ver) large horizontal forces, the bearings are specially designed to take care of the vertical loads and rotation of the box girder and simultaneously to provide all possible horizontal restraints (fixed bearing, bearing free lengthwise or crosswise, or both).

Special Design Aspects of Segmental Railway Bridges

461

NAGOYA -

ELEVATION

CROSS SECTION FIGURE 10.41. Kisogawa Bridge, Japan.

These bearings are specially manufactured for this tvpe of structure, Figures 10.46 and 10.47. The sliding parts consist of a teflon-coated plate resting on a stainless steel plate, and the restraints are provided by steel keys.

ference of potential with the ground may be measured at regular intervals, and a permanent connection with the ground may be decided on as a result. 10.10.~

10.10.4 STRAY CURRENTS

For structures carrying electrified railways there is some uncertainty about the long-term effect of stray currents generated near the power lines. In order to preclude electrolytic corrosion of reinforcing steel and prestressing steel, the following precautions are now taken in prestressed concrete structures: The deck is electrically isolated from the ground, piers, and abutments by elastomeric plates. The reinforcing and prestressing steel systems of the entire deck are interconnected by mild steel bars to equalize the electric potential. The dif-

DURABILITY OF THE STRUCTURE

Because very difficult problems of train traffic would arise during repairs to these bridges, their durability needs special attention. The following provisions were established for the high-speed bridges between Paris and Lyons: Under the worst service loads the concrete must remain under a 140 psi minimum compression. For continuous bridges, the design shall be checked by weighing the dead-load vertical force on the bearings. The stressing force of the post-tensioning tendons shall be less than 80% of the ultimate strength of the tendons.


462

317.00 m

ELEVATION

fy

FJ

L.

‘i

r’.

---e-J

I

53 2.61

‘!I

2.30

1

./.

2.70

5.90 10.92 m

/

1 L I

2.41

CROSS SECTION FIGURE 10.42. Ashidagawa Bridge, Japan.

10.10.6 CONCLUSION

FIGURE 10.43. Kakogawa Railway Bridge, placing precast segments with launching gantry.

The ultimate strength of the structure should be capable of supporting the service loads increased by 30%, if 30% of the post-tensioning steel were missing. Provisions shall be made for installing additional tendons while the structure is under traffic. The additional post-tensioning force shall be 15% of the designed force minimum. It shall be possible to replace all the bearings.

This review of specific design problems of railway bridges should raise no doubts whatsoever about the advantages of prestressed concrete and segmental construction’in this field. Prestressed concrete is the safest material known today to resist indefinitely the large variations of loads such as those applied to a railway bridge. The problem of fatigue has been covered briefly in Chapter 4, and the results mentioned there apply particularly well to railway bridges. The main objective in the design and construction of prestressed concrete bridges should be to minimize and even eliminate concrete cracking, which is always a source of weakness in a structure subject to cyclic loading. The use of the provisions laid down in Section 10.10.5 should result in practically crack-free structures with an expected life free of major maintenance.

978 I 184’

,

98’

,

, I

414’

98’_,_ I

184

I

I (b)

FIGURE 10.44. Akayagalva Rail\\av Bridge, general dimensions. ((0 Elevation. (1~) l‘vpical cross section .4-A.

55 k (25tl

5.4 k/cF (8 t/ml)

55 k 125t1

Y

5.2’

Span length Deck width Number of lanes or tracks Superimposed dead load: Ballast Curb, pavement, etc. Total S.L. Live loads: Equivalent uniform load Impact factor Total L.L. Total (S.L. + L.L.)

55 k (25d (8 t/ml1

V

Description

55 k (25d

V

5.2’

v

5.2’

2.6’ +I----

Highway Bridge

Railway Bridge

150 ft (45 m) 36ft (11 m) Three lanes

150 ft (45 m) 36ft(ll m) Two tracks

1.5 kips/ft

-

6.5 kips/ft 0.5 kips/ft 7.0 kips/ft

1.5 kips/ft 6.8 kips/ft

2.4 kips/ft

30%

18% 2.8 kips/ft

8.8 kips/ft

4.3 kips/ft

15.8 kips/ft

FIGURE 10.45. Vertical loading on railway bridges. (a) Typical UIC - track loading. (b) Comparison of superimposed dead and live loading on highway and railway bridges. 463

464

Segmental Railway Bridges 10.11

Proposed Concepts for Future Segmental Railway Bridges

We should note that many types of structures described for highway bridges are equally appropriate for railway bridges: the structures described in this chapter were essentially girder or arch bridges built in cantilever or incrementally launched. Today, many design projects are based on stayed bridges. As an example, Figure 10.48 shows a proposed crossing of the Caroni River in Venzuela for heavy iron ore freight trains.

FIGURE 10.46. Detail of pot bearing with unidirectional horizontal movement.

FIGURE 10.48. Proposed crossing of Rio Caroni for iron ore railway line.

FIGURE 10.47. Detail of fixed bearing.

11 Technology and Construction of Segmental Bridges 11.1 SCOPE AND INTRODUCTION 11.2 CONCRETE AND FORMWORK

FOR SEGMENTAL

CONSTRUCTION 11.2.1 11.2.2 11.2.3 11.2.4 11. 3

Concrete Design and Properties Concrete Heat Curing Dimensional Tolerances Formwork for Segmental Construction

POST-TENSIONING

MATERIALS

AND

11.6

11.6.1 11.6.2 11.6.3 11.6.4 11.6.5

OPERATIONS

11.3.1 General 1 1 . 3 . 2 Ducts 11.3.3 Tendon Anchors 11.3.4 Tendon Layout 11.3.5 Friction Losses in Prestressing 11.3.6 Grouting 11.4

11.3.7 Unbended Tendons SEGMENT FABRICATION FOR CANTILEVER CONSTRUCIION

Tendons

CAST-IN-PLACE

11.7 11.8

SEGMENTS

Launching Girders Launching Girders Slightly Longer Than the Span hn%h Launching Girders Slightly Longer Than Twice the Typical Span

CHARACTERISTICS OF PRECAST SEGMENTS AND MATCH-CAST EPOXY JOINTS

Certain problems are common to all types of segmental construction-for example, the selection and control of materials, prestressing operations, and choice of bearings, joints, and wearing surface. Other- problems are specific to a particular construction method. The use of form travelers in cast-in-place cantilever construction and the casting and handling of segments in precast cantilever construction are two such examples. This chapter covers these various topics in the following order:

CAST SEGMENTS PLACING PRECAST

11.8.3

Flanges 11.4.5 Practical Problems in Cast-in-Place Construction Camber Control

Scope and Introduction

Survey Control During Construction Conclusion 11.6.6 Precasting Yard and Factories HANDLING AND TEMPORARY ASSEMBLY OF PRE-

11.8.1 Independent Lifting Equipment 11.8.2 The Beam-and-Winch Method

11.4.2 Self-Supporting Mobile Formwork 11.4.3 Tw&tage Casting 1 1 . 4 . 4 Combination of Precast Webs with Cast-in-Place

1 I .l

Introduction Long-Line Casting Short-Line Horizontal Casting Short-Line Vertical Casting Geometry and Survey Control Segment Precasting in a Casting Machine Segment Casting Parameters Survey Control During Precasting Operations

11.4.1 Conventional Travelers

11. 5

11.5.1 First-Generation Segments 11.5.2 Second-Generation Segments 11.5.3 Epoxy for Joints MANUFACTURE OF PRECAST SEGMENTS

REFERENCES

1.

Problems common to all segmental bridges

2. Problems specific to cast-in-place cantilever construction 3. Problems specific to match-cast segmental bridges with particular emphasis on cantilever construction, which is the most widely used method. In designing segmental bridges, it is important to pay attention to certain details at the time of conception, in order to keep the project as simple as possible and thereby achieve economy and effi465

466


ciency during construction. The following guidelines apply to both cast-in-place and precast construction: 1.

Keep the length of the segments equal, and keep the segments straight even for curved structures (chord elements).

2.

Maintain constant cross-section dimensions as much as possible. Variations of cross-section dimensions should be limited to change of depth of webs and thickness of bottom slab.

3. 4.

Corners should be beveled to facilitate casting. Segment proportions (shear keys, for example) should be such as to allow easy form stripping.

5.

Avoid as much as possible surface discontinuities on webs and flanges caused by anchor blocks, inserts, and so on. Use a repetitive layout for tendons and anchors, if possible.

6. 7. 8.

Minimize the number of diaphragms and stiffeners. Avoid dowels passing through formwork, if possible.

11.2

Concrete and Formwork for Segmental Construction

11.2.1

CONCRETE DESIGLV

A,VD PROPERTIES

Uniform quality of concrete is essential for segmental construction. Procedures for obtaining high-quality concrete are covered in PC1 and AC1 publications.‘** Both normal weight and lightweight concrete can be made consistent and uniform by means of proper mix proportioning and production controls. Ideal concrete will have a slump as low as practicable, notwithstanding the possible use of special placing equipment such as pumps, and a 28-day strength greater than the minimum specified by structural design. It is recommended that statistical methods be used to evaluate uniformity of concrete mixes. The methods and procedures used to obtain the concrete characteristics required by the design may vary somewhat, depending on whether the segments are cast in the field or in a plant. The results will be affected by curing temperature and type of curing. Liquid or steam curing or electric heat curing may be used. In temperate climates and where curing is carried out in an isothermal enclosure, only small ad-

ditions of heat are required to maintain the curing temperature, full advantage being taken of the heat of hydration generated by the fresh concrete. In this case heat demand will be a function of the ambient temperature, more heat being required in winter and little or no additional heat during hot summer weather. Where segment production rate is not critical, it may be possible to do without accelerated curing and simply use a normal curing period of a few days, during which the concrete is well protected against excessive temperature variations and all exposed surfaces are kept moist. A sufficient number of trial mixes must be made to assure uniformity of strength and modulus of elasticity at all important phases of construction. Careful selection of aggregates, cement, admixtures, and water will improve strength and modulus of elasticity and will also reduce shrinkage and creep. Soft aggregates and poor sands must be avoided. Creep and shrinkage data for the concrete mixes should be determined bv tests. Corrosive admixtures such as calcium chloride should never be used, since they can have a detrimental effect on hardened concrete and can cause corrosion of reinforcement and prestressing steel. Water-reducing admixtures and also airentraining admixtures that improve concrete resistance to environmental effects, such as de-icing salts and freeze and thaw actions, are highly desirable. Very careful control at the batching stage is required, however, since the advantages of airentrained concrete cannot be relied upon unless the quantity of entrained air is within specified limits. The cement, fine aggregate, coarse aggregate, water, and admixture should be combined to produce a homogeneous concrete mixture of a qualitv that will conform to the minimum held-test and structural design requirements. Care is necessary in proportioning concrete mixes to insure that the\ meet specified criteria. Reliable data on the potential of the mix in terms of strength gain, creep, and shrinkage performance should be developed to serve as the basis for improved design parameters. Proper vibration should be used to permit the use of low-slump concrete and to allow for the optimum consolidation of the concrete.

11.2.2 CONCRETE HEAT ClJRI,\‘G An early concrete strength usually is required to reduce the cycle of operations and to maintain the

Concrete and Formwork for Segmental Construction

efficiency of the special equipment used either in cast-in-place or in precast construction. Two methods may be used for this purpose, either separately or together: (a) preheating the fresh concrete, before placing it in the forms or in the casting machines, (b) heat curing the concrete after consolidation in the forms. In the first case the concrete is preheated to about 85 to 90°F (30 to 35°C). This operation is achieved in several ways: 1.

2.

3.

Steam heating the aggregates-a simple solution that presents the disadvantage of changing the aggregate water content Heating the water-a solution that has limited efficiency, owing to the small proportion of water in comparison with the other components (water at 140°F raises the concrete temperature by only 20°F). Direct heating of the concrete mix by injecting steam into the mixer itself-the best solution and the one most easily controlled.

To avoid heat loss, the forms are generally insulated and some source of radiant heat is installed inside the segment (radiators or infrared elements). In the second case, the concrete is heated in its mold inside a container in which low-pressure steam is circulated. In this way it is relatively easy to obtain the strength required for prestressing operations [3500 to 4000 psi (25 to 28 MPa)] after one or two days, even in winter. If however, tensioning operations are to be performed earlier, after 24 hours for example, modifications must be made to the concrete in the anchorage zone. Electrical resistances may be embedded in the concrete, or precast end-blocks may be used. Precast end-blocks were used notably for the Issyles-Moulineaux, Clichy, and Gennevilliers Bridges. For the Gennevilliers Bridge, despite the exceptional dimensions of the box girder deck, two segments were cast each week through an early stressing of the prestress tendons. In the case of precast segments, the accelerated curing of the concrete must attain two apparently contradictory objectives: 1. 2.

Accelerated curing to permit rapid stripping. Final compressive strength as near as possible to that of the design concrete.

Several curing systems may be considered:

467

1. Conventional kilns. 2. Direct heating of forms with electric resistances. 3. Direct heating of forms with low-pressure steam. The use of a conventional kiln entails several precautions. First, a constant temperature must be maintained in the kiln. Second, the segment sections of varying thickness are all heated to the same temperature, which may produce unacceptable local thermal gradients and cracking if heat curing is excessive. Finally, the heated segment may be subjected to a thermal shock when removed from the kiln, if the difference between the ambient temperature and the kiln temperature is greater than 60°F. However, kiln curing is a simple solution and is acceptable for long curing cycles-for example, of 10 to 14 hours. Form heating by means of electrical resistances is perfectly adapted to long curing cycles. This system permits a wide range of adjustment per zone, varying the temperature between the thick and thin sections of the segment and thereby minimizing thermal gradients and eliminating the risk of permanent damage to the concrete at the beginning of its solidifying phase. The heating of forms with low-pressure steam is preferable for short curing cycles lasting less than five hours, as it permits the distribution of a large quantity of calories over a short period, causing a rise in the internal temperature of the concrete of the order of 20 to 30°F (10 to 15°C) per hour. This system, however, requires a complex regulator to ensure an equal temperature in all the form panel enclosures, at all times during the treatment, whatever their thermal inertia and the external influences to which they are subjected, Figure 11.1.

FIGURE 11.1.

Viaducts).

Heat-curing control system (B-3 South


468

The different systems (kiln, electrical resistances, and low-pressure steam) have all been applied successfully to segmental bridges. The segments for the Choisy-le-Roi and Courbevoie bridges were kiln cured. Electric heating was adopted for the construction of the upstream and downstream bridges on the Paris Ring Road and the Blois Bridge, among others. Form heating using lowpressure steam was used for the Pierre Benite Bridges, the Oleron Viaduct, and the B-3 South Viaducts. Whether forms are heated by electricity or b) steam, it is relatively easy to produce a long curing cycle, and the desired final concrete strength is easily obtained. A short curing cycle, on the other hand, requires a great deal of caution and meticulous preliminary calculations. Particular attention must be given to:

In order to avoid a drop in the long-term mechanical properties of the concrete, the temperature curve during the heat curing must necessaril\ include, see Figure 11.2:

1.

If the initial curing period is short

An initial curing period of two to three hours, during which the concrete is kept at the ambient temperature ,411 increase in temperature at a low rate of less than 36°F (20°C) per hour A period (depending upon the concrete strength to be attained) during which the temperature is held constant and below 150°F (65T) A period during which the concrete is cooled at a rate similar to that used for the temperature increase ‘The loss of strength in the long term will be greater:

Choosing a cement, the performance of which is adapted to the accelerated curing of concrete (preferred is artificial Portland cement with: C,,A s 11% and C,,SIC,S 2 3). 2. Consistently manufacturing concrete with a minimum water content and a maximum temperature of 95°F (35°C) at the time of pouring.

As an example, the short-cycle treatment used for the B-3 Viaduct segments was the follo\~ing, see Figure 11.3:

3. Using sufficiently rigid forms to resist the thermal expansion of the concrete in its plastic state while heating.

Initial period of 14 hour at 95°F (35°C) (mixing temperature)

PREHEATING

212 i

If the temperature increase is rapid If the maximum temperature is high

ALTERNATIVE

I

FORM I

I

I

I

I

2 to 3H -Y-Jw\COOLING piNI;; TEMPERATURE CONSTANT TEMPERATURE INCREASE PERIOD

FIGURE 11.2. Heat-treatlnent c-~c-le.

ti

STRIPPING

Concrete and Formwork for Segmental Construction 1 I .2.4 SHORT

LONG

CYCLE

CYCLE

Temperature increase of 27°F (15°C) per hour for 2 hours .-\ constant temperature of 150°F (65°C) for lf hours Figure 11.3 shows an example of- tong-cycle heat treatment, the Conflans Bridge, which had a total heat-treatment duration of 19 hours.

11.2.3 DI,~lESSIO,\-‘4L TOLERrllVCES Formwork that produces tvpical bridge box girder segments within the following tolerances is considered to be of good quality3.4:

FORMWORK FOR SEGMENTAL CO,\5 TR UCTIO,V

Formwork along with its supports and foundations must be designed to safely support all loads that might be applied without undesired deformations or settlements. Soil stabilization of the foundation mav be required. Since economical production of cast-in-place or precast segments is based on repetitive use of the same forms as much as possible, the formwork must be sturdy and special attention must be given to construction details. Where formwork is to be assembled by persons other than the manufacturer or his representatives, particular care must be taken with erection details and assembly instructions. All elements of the formwork must be easy to handle. 1-Z Formwork for structures of variable geometry will need to be relatively flexible in order to allow adaptation at the various joints. Both external and internal forms are usually retractable in order to leave a free working space for placing reinforcing steel and prestressing ducts3 Special consideration must be given to those parts of the forms that have variable dimensions. To facilitate alignment or adjustment, special equipment such as turnbuckles, prefitted wedges, screws, or hydraulic jacks should be provided. Tendon anchors and inserts must be designed in such a way that they remain rigidly in position during casting. Projecting anchorage blocks or other such irregularities should be detailed to permit easy form stripping.3 If accelerated steam curing with temperatures of the order of 130°F (55°C) is to be used, then the deformations of the forms caused by heating and cooling must be considered in order to prevent cracking of the young concrete. In general, internal vibration using needle vibrators should always be applied. External vibrators, if used, must be attached at locations that will

2; in. +f in. to 0 in.

(+ 10 mm)

Depth of top slab Overall depth of segment

k+ in.

(55 mm)

?& of depth with f in. min.

(5 mm min.)

Overall width of segment

?h of width

(5 mm min.)

Length of match-cast segment

with f in. min. *a in.

(25 mm)

Diaphragm

ki in.

(210 mm)

Width of web Depth of bottom slab

dimensions

469

(+lO mm to 0 mm)

470


achieve maximum efficiency of consolidation and permit easy replacement in the case of a breakdown during casting operations. External vibration may lead to fatigue failure in welded joints, and regular inspection should be made to help prevent any sudden failure of this kind.3 Paste leakage through formwork joints must be prevented by suitable design of joint seals. Normally this can be achieved by using a flexible sealing material. This is particularly important at the joint face with the matching segment, where loss of cement paste can lead to poorly formed joint surfaces and subsequent spalling and loss of matching, requiring repair. Special attention must be given to the junction of tendon sheathing with the forms.3l4 All form surfaces, especially welded joints in contact with the concrete, must be perfectly smooth and free from reentrant areas, pitting, or other discontinuities, which could entrap small volumes of concrete and lead to spalling during form stripping.3

I I .3

Post-Tensioning

Materials

and

Ducts must have sufficient grouting inlets, shutoff valves, and drains to allow proper grouting and to avoid accumulation of water during storage. Vent pipes should not be spaced more than approximately 400 ft (120 m) apart.’ This spacing may have to be reduced, depending upon the expertise of the personnel performing the grouting. Particular attention must be paid to the qualit! of duct connections at the joints between segments. At the joints, accurate placing is mandator\. ~I‘he method of duct connection depends on the type of joint3: Telescopic sleeves -wide joints Screw-on

type

pushed

over

sleeves-wide

projecting

ducts

joints

Internal rubber or plastic sleeves-match-cast joints Gaskets

or

other

special

seals-match-cast

joints

No special provisions: clean ducts with a torpedo after jointing to remove penetrated epoxv if anv-match-cast joints

Operations

1 1 3 . 1 GE,\‘ERAL Technical details relating to the different methods available are described in the various posttensioning manuals5g6 and in the specific documents issued by suppliers. 113.2 DUCTS Ducts are used to form the holes or enclose the space in which the prestressing tendons are located. The ducts may be located inside or outside the concrete section. Although in some instances the tendons are placed in the ducts before concreting (cast-in-place and span-by-span construction), post-tensioning tendons will normally be threaded into the ducts af-ter erection of the segments. The duct cross section must, therefore, be adequate to allow proper threading; and in general it will be about f in. (5 mm) larger in any direction than for ducts in which the tendons are placed before concreting. The duct dimension must allow not only the installation of the tendons but also free passage of grout materials after stressing. The ratio or proportion of cross-sectional area of the duct with respect to the net area of prestressing steel should conform to appropriate specifications or codes.4 A minimum value of 2 usually leads to satisfactory results.

Connection tightness is essential in order to prevent penetration of joint material, water, or other liquids or solids into the ducts, which would introduce a risk of blockage, and also to prevent leakage at the joint during tendon grouting operations.3 1 1 . 3 3 TE,VDO.V

ASCHORS

Tendon anchors usually consist of a bearing plate and an anchorage device either in combination or as separate units. Shape and dimensions of the anchors must conform with the applicable specifications, particularly insofar as bearing stresses are concerned. Choice of anchor positions in the segments should take into account the following considerations3: Tendon layout quences.

requirements

and

installation

se-

Stresses generated around the anchors. Ease of tendon threading and stressing. Ease of formwork crete placing.

preparation, stripping and con-

Certain anchorage positions, such as the anchorage block on a thin slab shown in Figure 11.4, should be avoided. If this type of detail cannot be avoided, then particular care must be taken in design and construction of the zone concerned.3

471

Post-Tensioning Materials and Operations

Tendonanchorageblocks

FIGURE 11.4. .Anchor-qr

block position to he

;I\ aided.

Bearing plates are usually embedded in the segment at the time of casting. In certain cases they are installed against the hardened surface of the concrete with a dry mortar bed or a suitable cushioning material such as asbestos cement or synthetic resin.

struction under ordinary working conditions and supervision. The actual results obtained in a segmental bridge built in Europe are given below by way of example for the benefit of future project designers. Cantilever tendons were placed along a straight profile in the roadway slab and anchored either on the segment face or tn a block-out inside the box girder. Continuity tendons were either anchored in a block-out at the bottom slab level or draped upward in the webs and anchored in the same block-out of the cantilever tendons. All tendons were made up of twelve 0.6 in. diameter strands. Soluble oil for reducing friction in the ducts was not allowed by the consultant. The calculations were carried out using the following values for curvature and wobble friction coefficients: /..l = 0.20,

This subject has been covered in Chapter 4 relating to design. The choice of tendon layout must be treated carefully, with special attention paid to the following factors: Construction sequence with respect to tendon placing, segment casting (or erection), and other construction imperatives Standardization and repetition of essential features, especialla duct and anchor positions at joints (in order to facilitate formwork design) Various loading conditions throughout struction period and in service

the

con-

When using large tendons, it is not advisable to use couplers or crossed splices, for reasons of congestion and formwork complication. Also, couplers and splices should not be located in areas where vielding mav occur under ultimate load conditions.3 In order to limit friction losses, and to facilitate tendon threading, excessively curved tendons should be avoided if possible. 11.3.5 FRICTIOS LOSSES IS PRESTRESSISG TESDO,VS Segmental construction usually calls for prestressing tendons to be installed through a succession of short duct lengths coupled to one another at the joints between segments, these being at approximately 8 to 30 ft (2.5 to 10 m) intervals. The friction factors (for curvature and wobble) usually accepted for long tendons in cast-in-place structures may not be realistic for this type of con-

K = O.OOf/ft

= 0.0021/m

The Young’s modulus of the tendon samples tested in the factory or in the laboratory varied between 28,000 and 29,000 ksi, and the variation between various heats over the whole structure was very low. According to direct tests carried out on site, and a systematic analysis of all results of tendon elongations recorded during the stressing operation, the actual Young’s modulus of a (twelve 0.6 in. diameter strand) tendon at first tensioning varied between 25,000 and 26,000 ksi, which is only 90% of the value recorded during factory and laboratory tests. Figures 11.5 and 11.6 show values of the wobble friction coefficient K measured for all the tendons in the structure’s 18 cantilevers. All the tendons are shown in Figure 11.5, while Figure 11.6 shows only those tendons in the spans without hinges, and separates the tendons anchored on the segment face from those anchored in block-outs (the tendons had the same layout except a rather severe curvature at the end). It is obvious that: As construction proceeded and the quality of manufacture and supervision improved, the results got better. At the beginning of the job, the effect of the curved ends of certain tendons was lost in the generally mediocre results. As these results got better (value of K equal to that used in calculation from cantilever 11 on), this effect became preponderant, counteracting that of the improved standard of work. As the site staff became accustomed to the work and the effort and supervision dropped, the results became gradually worse (compare cantilevers 13 and 17, for example).


472

dK,

M 2

3

k

5

6

7

8

9

10

11

12

13

lk

15

16

17

18

CANTILEVER

FIGURE 11.5. Prestressing in a cantilever bridge. Variation of uobble 1.1.ic coefficient for cantilever tendons in each of the structure’s 18 spans.

n ”

tion

1O’K

ANCHORED

IN

BLOCK - OUTS

TENDONS THE 2

ANCHORED

AT

SEGMENT 3

5

7

9

11

13

is

17

18

w CANTILEVER

FIGURE 11.6. Prestressing in a cantilever bridge. Wobble friction coefficient for C;LIItilever tendons in the 10 spans without hinges. As an example, a straight tendon in the top slab fillet between slab and web was isolated. The wob-

ble friction coefficient depends on the care exercised in fastening the duct to the reinforcing steel

cage as the concrete is poured (when the tendon is in the slab rather than in the fillet, the accidental

deviations are much smaller). For the first seven cantilevers (see Figure 11.7) the wobble coefficient

473

Post-Tensioning Materials and Operations

F L E X I B L E

1

2

3

4

5

6

+

8

L

9

L::!!::!: :-

10

11

12

0

U

15

16

17

M

FIGURE 11.7. Prestressing in a cantilever bridge. Wobble friction coefficient for a straight tendon located in the upper fillet.

reached up to six times the assumed value used in

the calculations, and yet very careful construction will enable this assumed value to be reached or at least approached closely to obtain the desired prestress with little room for uncertainty. The presence of hinged segments not only complicates the tendon profile and the construction phases, but introduces uncertainty about obtaining

the required prestress force. Owing to the technical restrictions imposed by the consultant, the traditional prestress layout employed in earlier bridges could not be used. Consequently, long tendons stressed only at the opposite end had to be accepted. It was thought that a realistic value of the final force for each of the tendons (twelve 0.6 in. diameter strand) would be 350 kips (160 mt). It is


474

fortunate that a direct check was made at the site, which revealed the actual initial load at transfer to be the following for the four tendons under consideration: 130 kips (60 mt), 210 kips (96 mt), 130 kips (60 mt), and 200 kips (90 mt). The average initial prestress load per tendon was therefore 170 kips (78 mt), and the probable final force would have been 150 kips (70 mt) as compared to the assumed value of 350 kips (160 mt). Fortunately, the situation could be easily corrected and remedial measures put into effect as follows: 1. The reinforcing steel and local prestressing tendons allowed for a certain margin of safety. 2.

It was possible to restress two of the four cables in the first cantilever and then to change the profile and method of placing segments in order to stress all the tendons at both ends for the rest of- the cantilevers.

The above results, quoted rigorously so as to illustrate several important aspects of friction losses, must not lead the reader to suppose that the safetv of the structure was at any time compromised. Thk force in a tendon varies much more slowly than anv changes in the friction coefficients for ordinar) tendon profiles. For example, in a 270 ft (80 m)

tendon stressed at both ends, if the friction coefficients are multiplied by 4, the minimum force in the tendon is reduced bv only 16%. It is interesting to examine the results for the actual prestress obtained in cantilevers 2 and 3 (the ones having the worst results) shown in Figure 11.8 for each section, compared with the prestress used in the calculations. The lack of prestress, most marked at midspan, was compensated by additional tendons to bring the force back up to that required by the calculations in the first two spans. Afterward, the originally calculated prestress was alwavs sufficient. To summarize, the authors wish to underline the following points: 1. Bench tests should be performed on site to determine a realistic value of the modulus of elasticity of the tendons to be used to compute the theoretical tendon elongations. 2.

Realistic values of curvature and wobble friction coefficients should be used in the design and further controlled on site. Direct friction tests should be made together with a statistical analvsis of the measured elongations for all tendons.

SUPPORT

5UPPORT MID - 5PAN %f-

l.ax,

-

EFECTIVE

FIGURE

PRESTRESS IN

11.8.

5PAN5

2AND3

h

-

Prestressing in a cantilever bridge. Effective prestress in spans 2 and 3

1.W

Segment Fabrication for Cast-in-Place Cantilever Construction

475

Provisions should be made at the design stage for additional prestress to compensate for any unexpected reduction in the design prestress force due to excessive friction. This may be done as follows: a. By adding additional ducts over and above the number required by design calculations; if this method is used, the unused ducts at the end of construction must be grouted to prevent water from seeping inside and subsequently freezing with disastrous effects on the structure. b . By using larger than required sizes for some of the ducts, so as to allow the use of larger-capacity tendons if required. By providing anchor blocks and possible C. deviation saddles so as to allow the installation of external tendons located inside the box girder but outside the concrete section.

used because it increases the moisture content of the air and reduces the natural corrosion protection. Another important and sometimes acute problem relates to potential grout leakage at segment joints, which can lead to the passing of grout from one duct to another. For this reason ducts must be well connected and sealed at joints. To check the grout tightness of the joints and to avoid blockages, it is advisable to flush the ducts with water under pressure before grouting. Any leakage points thus detected may then be sealed. If communication is discovered between tendon ducts, the tendon groups affected should be grouted in one operation after threading and stressing of all the tendons involved. 3 If couplers are being used (notably for single-bar tendons), precautions tnust be taken to limit the risk of grout blockage at the coupling points. Couplers must be housed in special enlarged enclosures with two essential features3:

If the correct approach is taken at the conception stage, perf-ect control of this aspect of prestress mav be obtained and verv satisfactorv structures can be built that give maintenance-free long-term performance.

I. Clear cross-sectional area for the passage of grout equal to or greater than that for the rest of the tendon.

3.

I I .3.h

2.

Independent grout inlets and vent pipes.

GROC’TI,\‘G

As in conventional post-tensioned structures, segmental construction requires the grouting of prestressing tendons after tensioning to provide corrosion protection and to develop bond between the tendon and the surrounding concrete. Current recommendations and provisions of good practice are therefore applicable to segmental bridges. However, several important points need to be examined. Grouting must not be carried out if the temperature in the ducts is less than 35°F (2°C) or if the surrounding concrete temperature is less than 32°F (2°C). This requirement virtually precludes grouting operations during the winter months in the northern and middle western United States, unless very special winter precautions are used. It is preferable to postpone all grouting operations until the following spring, even though some tendons may be left tensioned and ungrouted for a long period. Attention must then be given to corrosion protection of the high-tensile steel bars or strands. Satisfactory protection is obtained by sealing all tendon ducts at both ends after blowing out with cool compressed air. Hot air should not be

I1 3.7 UNBONDED TENDONS Unbonded tendons may be used in segmental construction provided that the performance requirements of the post-tensioning steel are also met by the tendon anchorage, notably with respect to fatigue characteristics. In unbonded post-tensioning a corrosion protection system must be provided to guarantee at least the same degree of corrosion protection as for bonded tensioning. This may be achieved by enclosing the tendons in flexible ducts (such as polyethylene pipes) and by cement grouting after tensioning.

I I .4


11.4.1

CONVENTIONAL

TRAVELERS

The conventional form traveler supports the weight of fresh concrete of the new segment by means of longitudinal beams extending out in cantilever from the last segment in order to support the forms and service walkways.

Form Travelers with Top Main Beam (Figure 2.83) The longitudinal main beams or girders are usually located above the segment to be concreted, in line with the webs. The outside forms, the bottom forms, the work floor, and the service walkways are hung from the main beams with the help of cross beams. The inside forms are supported on a trolley, which travels inside the deck. The main beams are anchored to the previous segment. In order to maintain stability during the pouring operation a counterweight is sometimes used to reduce the uplift forces applied to the concrete section. When the traveler is transported to its new position ready for the next segment, the counterweight keeps it in balance between two successive anchoring positions. The main beams that support the load due to concrete, forms, walkways, and so on are often subject to large deflections, which can give rise to transverse cracking along the joints between segments. These cracks appear at the upper face of the bottom slab and at the connection between web and top slab. This undesirable condition can be avoided by using a rigid structure; the weight of the traveler is increased together with the prestress required in the cantilevers. The form traveler used for the Oissel Bridge weighed 120 tons (110 mt) and may be considered as a heavy form traveler. If the travelers are light, care must be taken to compensate deflections during concreting by adjusting jacks. This type of traveler weighs (excluding counterweight) a little less than half the maximum concrete segment weight. An example of a light form traveler is shown in Figure 11.9 for the Tourville Bridge. Each traveler weighs 33 tons (30 mt). Form Travelers with Lateral Main Beams (Figure 11.10) Travelers with their main beams above the bridge deck present the disadvantage of hindering the construction operation concerning the upper part of the segment. For this reason certain form travelers have their main beams disposed laterally parallel to the outside webs, underneath the bridge deck. This solution leaves a clear working surface and allows easy access to all surfaces to be formed, reinforced, and concreted. In this way, the technology originally developed for precast segmental construction can be applied to cast-in-place cantilever methods, resulting in shorter construction cycles. The Moulin-les-Metz Bridge in eastern France, Figure 11.11, was constructed using this type of form traveler.

FIGURE 11.9. eler. 1 I .4.2

I‘ourville-la-Kwiere

Bridge form trav-

SELF-SUPPORTING MOBILE FORMWORK

In the case of traditional form travelers, the resulting deflections seen during construction are almost entirely due to the main beams. The formwork as such usually acts only as a mold and does not support any part of the total load, even though it comprises very stiff walls. In several recent bridges the traveler concept has been modified so as to use the rigid formwork as the weight-carrying member, thus producing a self-supporting rigid mold. Several advantages are gained with this concept: Surveying control and correction of bridge deck geometry are easily obtained. Cracking near the joints caused by the deflection of conventional travelers is completely eliminated. The work area is maintained completely free and allows prefabricated reinforcing steel cages to be used as in precast segmental construction. This type of mobile formwork was first used for constant-inertia bridge decks such as the Kennedy Bridge, Dijon, and the Canadians Interchange in Paris, Figure 11.12. During the concreting operations, the mobile formwork is prestressed to the existing deck. The exact positioning of the formwork is obtained by

Segment

Fabrication

for

Cast-in-Place

Cantilever

Construction

1 a CONCRETING PHASE

2, LAUNCHING PtlASE

FIGURE 11.10. Typical for.rn traveler with later .a1 main beams.

FIGURE 11.11. Moulins-les-Metz

form traveler.

means of adjusting pins located at the rear in reservations provided in the previously poured segments. The formwork is transported to its new position, ready for the next segment, on an overhead trolley, which travels along short steel girders cantilevered out from the existing hardened concrete in line with the webs. A further refinement was to use pretensioned reinforcing to add to the stability of the traveler

FIGURE 11.12. Canadians Viaduct (Paris), view of form traveler in operation.

while pouring the segment. Figure I 1.13 shows the arrangement for the Canadians Viaduct in Paris, France. Monostrands located in the webs are provisionally anchored to the front of the traveler and embedded in the webs of the concrete segments to be incorporated in the reinforcement of the permanent structure. The use of the self-supporting mobile formwork was later extended to variable-depth bridge decks

478


S&le S t r a n d s 1

(4x0 6”)

tic R e a r Fixation ’

iA

i.E

FIGURE 11.13.

U er Fixations /-=-7

Sinqle S t r a n d s Canadians

2.

\

’

Lower Fixations

Viaduct (Paris), details of the self-supporting form traveler.

as well as three-web cross sections, as in the Clichy, Orleans, and Gennevilliers Bridges. The structural members of the mobile formwork are therefore the side forms of the exterior face of the outside webs and the bottom forms of the underside of the bottom slab, both of which are stiffened transversely by front and rear frames braced together for additional rigidity, Figure 11.14. In this manner a rigid box is formed, which is prestressed to the existing deck. The change of section height is achieved by vertical displacement of the bottom forms, which are fastened to the front stiffening framework and bottom slab of the last segment. The stability of the self-supporting mobile forms of the Gennevilliers Bridge was ensured by (Figure 11.15): 1.

I SECTION B.B

SECTION A.A

\

’

Two steel pins fixed to the top of the outside forms and matching imprints provided on the outside face of the previous segment, the connection being assured by high-strength bars going through each web. Two steel pins fixed to the upper surface of the bottom forms and matching the corresponding imprints provided in the last segment bottom slab, again held by prestress bars.

The self-weight of the mobile forms and the fresh concrete creates an overturning moment, which is

balanced by two forces F sustained by the previously described locating pins. Practically all the shear force is taken by the upper pins. Because of the large forces transmitted through the top pins to the concrete, precast concrete elements are used to avoid the transmission of high stresses to young concrete, Figure 11.16. These forces are transmitted by friction between pin and concrete, and this determines the necessary prestress force. 11.4.3 TWO-STAGE CASTING

The method of two-stage casting involves, first, the fabrication of the bottom slab and the webs together with a small part of the top slab in order to create a flange in which all or some of the cantilever tendons can be located. This operation, carried out using a conventional form traveler, produces either a U-shaped or a W-shaped section, depending on the number of webs, Figure 11.17. After the cantilever tendons are stressed the form traveler is moved to the next position, the top is poured using a mobile formwork of relatively simple design. This second stage usually follows the first with a minimum interval of two or three segments, and concrete can be placed in a simple pour over the length of several segments. This method has the advantage of reducing the concrete volume to be supported by the form traveler, thus reducing the weight of the traveler.


bunt a

n

d

479

rrar stiffrncrs !

‘B0tt0m

f0rms

’

Mob/h t r u s s

\Bottom

forms

FIGURE 11.14. Self-supporting mobile forms for variable-depth bridge decks. (a)

Concreting. (b) Moving f&ward.-

In addition, the second stage is independent with respect to the first and so is no longer on the critical path of concreting operations. The bridge decks of the Saint Isidore and Magnan Viaducts on the Nizza A-8 bypass were constructed using this method. All of the 130 ft (40 m) spans of the Saint Isidore Viaduct were completed for stage one only, including closure to the preceding span, before the second stage was completed, using mobile formwork which rolled along the bottom slab from one abutment to the other. As regards the Magnan Viaduct, the second stage followed the first with an interval of three seg-

ments, because of the long spans in this structure. The same procedure was used for the Clichy, Joinville, and Woippy Bridges, Figure 11.18.

11.4.4

COMBINATION OF PRECAST WEBS WITH CAST-IN-PLACE FLANGES

The preceding methods allowed a considerable reduction in the construction cycle. Two pairs of segments could thus be completed every week, corresponding to an average rate of construction of 7 to 10 ft (2 to 3 m) per working day.

MOBILE

FORM

STABILITY

Prestressina

FIGURE 11.15.

bars

Stability of the Genne~illiers Bridge self-supporting mobile forms.

_

PRECAST

JOINT

_

Pin

FIGURE 11.16. Precast

FIGURE 11.17.

gusset

for Genne\illiers Bridge

Two-stage construction of a two-web bridge deck.


y’

. I. .1. .(r+.“.r.,

--,.. . ” ” ., x -,___-_ i. j ‘1. .‘.W *&e “,~..,‘?C~~:r^~.y’~.~~ **v.,; ;a -3 - _, _ ,

FIGURE

11.18. \Voippy Viaduct, France. Detail of the self supporting form traveler and two-stage casting.

The main obstacle preventing further reduction in the construction cycle and therefore a closer approach to the speed of precast segmental construction is the lack of strength of young concrete and the consequent interference with stressing operations. Apart from several other methods already discussed, the problem can be partially overcome by using precast end blocks or precast webs or both. This was first tried for the construction of the Brotonne Viaduct approach spans, Figure 11.19. The webs, which were rather thin and heavily inclined, were precast in pairs and pretensioned, Figure 11.20. The deck Was cantilevered out from the piers using 10 ft (3 m) long segments assembled in two phases. In the first phase, the precast webs weighing up to 18 tons ( 16 mt) were placed inside the form traveler, previously adjusted to the bridge profile including the desired camber. The webs were then prestressed to the preceding segment with provisional prestress bars, the joint being TENSION

FIGURE 11.19. Brotonne Bridge, mobile form car-

rier.

match-cast or cast in place. The second phase consisted of casting the rest of the segment inside the form traveler, which was now suspended from the newly stressed webs. This procedure, which requires partial prefabrication of the segments using light casting equipment, enables a considerable simplification of the form traveling equipment, the limitation of total weight to 39 tons (35 mt), and a reduction in the construction cycle such as to produce, even for a cable-stayed bridge, as many as four segments per week for each pair of form travelers. COUPLER TENSION

RODS STEEL

FOR RODS

TENSIONING

JACK

36 mm dia TENSION RODS

FORMS

ADJUSTABLE

ADJUSTABLE BRACKETS

11-36

mm dia TENSION RODS

TENDONS

BED FOR PRETENSIONED WEBS

FIGURE 11.20. Brotonne Bridge, precasting of webs.

CASTING

BED

482


During construction of the Brotonne cablestayed bridge, the precast webs were placed by tower crane traveling parallel to the bridge deck above the river banks and by an overhead gantry crane above the Seine River. Another example of the use of precast webs is found in the Clichy Bridge carrying the metropolitan line over the Seine in the northwest of Paris. The bridge deck with a 280 ft (85 m) maximum span consists of a three-web box girder without cantilever flanges and with the deck supporting the live loads as low as possible in order to reduce the length of the access ramps to the structure. The 8 ft (2.5 m) long segments were also constructed in two stages, Figure 11.2 1. The precast webs, with epoxy match-cast joints, are placed with the aid of a mobile handling system rolling along the webs of the previously placed segments. They are then prestressed to the existing structure before the top and bottom slabs are poured in place on the length of two segments. 11.4.5 PRACTICAL PROBLEMS IN CAST-IN-PLACE CONSTRUCTION CAMBER CONTROL

Before proceeding with the cantilever construction proper, a starting base must first be completed on the various piers. This first special segment, called a pier segment or a pier table, is generally constructed on a temporary platform anchored by

FIGURE 11.21. Precast web placing equipment for Clichy Bridge carrying the metropolitan line over the Seine River.

ELEVATION

SECTION

Rotressinq

ban/ I

I

FIGURE 11.22. Construction of the pier segment for a cast-in-place cantilever deck.

prestressing the pier top, Figure 11.22. This special segment may either be given the minimum length to insure adequate connection to the pier for the stability of the future cantilever or else be of such length as to allow both travelers to be installed simultaneously, Figure 11.23. Another important problem relates to the safety of the travelers during construction. Chapter 4 described the difficulties of ensuring pier safety in the event a form traveler fell during transfer from one position to the next. The difficulties would even be greater in the event of an accident during the casting operation. Consequently, all precautions must be taken both at the design stage and during construction to eliminate this potential hazard. The load-carrying members of the traveler must be carefully inspected and Ray even be load tested before use so as to practically eliminate the danger of structural failure. The most critical areas are in the safety of the suspension rods and the transfer of the traveler reactions to the concrete. Preferably all suspension rods and anchor bars should be doubled. Also, the prestressing tendons must have an adequate margin of safety. Use of a single strand or a single bar in each web of the box should be avoided. Rather a multistrand tendon with individual anchors for each strand or two prestress bars should be used. Worldwide use of cast-in-place cantilever construction has established an extremely good safety record, much better than that for cast-in-place construction on fabework. Accidents are very few and far between; however, designers and constructors must always be safety conscious.


483

stress forces impose upon the cantilever a new deflection curve. 3. Deflections of the various cantilever arms after construction and after removal of the travelers before continuity is achieved with the other parts of the deck. 4. Short- and long-term deflections of the continuous structure, including the effect of superimposed dead loads (curbs, railings, pavements, utilities, and so on) and live loads. 5. Short- and long-term pier shortenings and foundation settlements.

FIGURE 11.23. Stal-t of cantilever construction from the piel- segment. (n) Short pier segment - successive inst;illation of travelers. (6) Long pier segment -simultaneous installation of travelers.

The most critical practical problem of cast-inplace construction is deflection control, particularly for long-span structures. There are five categories of deflections (or space geometrical movements of the structure) during construction and after completion: 1.

Deflection of the travelers under the weight of the concrete segment. This value is given by the manufacturer or may be computed and checked at the site during the first operations. 2. Deflection of the concrete cantilever arms during construction. For each casting of a pair of segments, the weight of the concrete segments and the corresponding cantilever pre-

Using the data available on concrete properties and foundation conditions, the designer should compute the various deflections mentioned under items 3, 4, and 5 above, assuming the bridge unloaded for foundation settlements and long-term concrete deflections and half the design live load for computation of the short-term concrete deflections. The sum of the various deflection values obtained in the successive sections of the deck allows the construction of a camber diagram, which should be added to the theoretical longitudinal profile of the bridge to determine for each cantilever arm an adequate casting curve. This casting curve is the goal toward which construction proceeds during cantilever casting. The essential difficulty is that no absolute coordinates are available in a system where everything changes at each construction stage (transfer of traveler, concrete casting, or cantilever prestressing). A very simple example may illustrate the solution of the problem of accommodating the deflections described under item 2 above. For simplicity, assume only a four-segment cantilever arm, for which a horizontal longitudinal profile is required, Figure 11.24. As outlined in Chapter 4 and summarized briefly above, the designer analyzes the various deflection curves for each construction step (casting segment and precasting). The typical results are shown in Figure 11.24. The cumulative deflection curve is immediately obtained together with the camber diagram, Figure 11.25. The use of the camber diagram for determining the adequate deflection at each construction stage is simple; however, it is much less simple to use in a proper manner in the field, and experienced surveyors have often made mistakes. When properly used, the camber diagram allows the determination at each joint, of offset values such as yle2, yzm3, and y3.4 at each point, which will

484


I

ELEVATION OF TYPICAL CANTILEVER

Downward deflection IS posltlve I CASTING AND PRESTRESSING SEGMENT

0 0 0 0

I

VERTICAL DEFLECTIONS (in mm) &

&J

@

(-11)

(-17)

(-23)

1

-5

2

1

5

(9)

1131

3

5

10

20

(30)

4

8

16

29

49

+9

+22

41

69

TOTAL DEFLECTION

FIGURE 11.24. struction stage.

Partial deflections due to girder weight and prestressing

bring the traveler in the proper position to realize the desired final geometry. The sketch and table in Figure 11.26 show how to use the camber diagram properly. It is very important to realize that at no construction stage does the profile of the cantilever coincide with either the final deflection curve or the camber diagram. The natural tendency would be to build up the traveler to the required offset to make its nose fall

at each ~011.

exactly on the camber diagram. The results of this improper procedure are shown in detail in Figure 11.27. The bridge is built with an undesired double curvature, particularly undesirable toward the end of the cantilever. When the mistake is discovered, it is usually too late to put into effect any remedial measures, because the final shape of a cantilever depends essentially upon the accuracy of the geometry near the piers, where the deck is sub-

Characteristics of Precast Segments and Match-Cast Epoxy Joints

-69

/

Camber

curve

41

/’

,’

.' /‘

485

Yr 9 6 0

.LY,-;-~

( a s s u m e polygonai lme 1

FIGURE 11.25. Cumulati\-e

camber.

deflection curve and choice of

b.

jetted to the highest moments and where its deflections have the greatest effect at midspan. 2. I I .5

Characteristics of Precast Segments Match-Cast Epoxy Joints

and

Developed originally to allow a rapid and safe assembly of precast segments at the construction site, the technique of match casting was progressively refined as experience was gained. We shall describe the characteristics of segments in the early structures to further highlight the latest improvements and variations of the original concept. 11.5.1 FIRST-GE~VERATIOS

SEGME.VTS

In those early structures the epoxy resin played several important roles: 1. During assembly before hardening: a. To lubricate the mating surfaces while final positioning took place.

To compensate for minor imperfections in the match-cast surfaces.

In the finished structure after hardening: a. To ensure the watertightness of the joints, especially in the top slab. b. To participate in the structural resistance by transmitting compression and shear forces. However, before hardening of the epoxy resin, the joints present no shear resistance whatsoever, because the epoxy behaves like a perfect lubricant. It was therefore necessary to provide shear keys in each web in order to ensure the shearforce transfer between segments. These keys, as well as those situated in the top slab, also allowed a very accurate assembly of one segment with respect to another.

During assembly of the deck, some sort of temporary fixation, either mechanical or by means of prestress bars, allowed the placing equipment (launching girder, crane, and so on) to be quickly

486


1

I

FIGURE 11.26. Follow-up of deflections with proper

camber

diagram.

unloaded without waiting for the cantilever tendons to be stressed. Figure 11.28 shows how a typical first-generation segment can be assembled to the existing structure using a temporary apparatus located on the top and bottom slabs, which is used to create forces F, and F, which ensure the equilibrium of the new segment at the joint. These two forces, combined with the weight W of the segment, give the resultant force R, which is inclined with respect to the joint. Because of the very small coefficient of friction of the epoxy, the shearing component of R produced by W can be balanced only by the vertical component of the reaction C, which exists normal to the bottom face of the web shear keys, Figure 11.28. The resultant R is composed, therefore, of the oblique reaction C supported by the shear keys and a horizontal reaction N, which is responsible for securing the joint. The axial stress distribution at the joint cross section differs in this case from what would be ob-

use

of

tained by ordinary calculations. It is obvious that N is smaller than F (the sum of forces F, and FJ. Let (Y be the angle of the key support faces with respect to the horizontal; then F - N = W tan (Y, and for a typical case of tan (Y = 0.50, F - N = W/2. Consider a segment weighing 50 tons (45 mt), temporarily assembled by a prestress force of 100 tons (90 mt) located in the top slab; the axial force reduction is 25 tons (23 mt)-that is, 25% of the total applied prestress force. If the rate of erection of the precast segments is sufficient to ensure the positioning of four segments before the resin in the first joint has set, then the reduction in the effective axial force in this joint will be 100 tons (90 mt), which more or less corresponds to one tendon of twelve f in. diameter strands. The same conclusion would be valid when the permanent prestressing was used to ensure the temporary stability of the cantilever. In conclusion, it is recommended that this reduction of the effective prestress force be taken


.-~*-~

i-21)

487

.~~.-~~( 13) C-25)

FIGURE 11.27. Follow-up of deflections with improper use of camber diagram.

TEMPORARY

SEGMENT

ASSEMBLY

JOINT EPUILIBRIUM Fl

(J) lb)

(a)

FIGURE 11.28. equilibrium.

Temporary assembly. (a) Elevation of temporary assembly. (6) Joint

into account while verifying the cantilever resistance and stability. Failure to do so may result in temporary joint opening, which is undesirable although not dangerous for stability.

It is also preferable to choose the intensity and the point of application of the forces F, and F, such as to allow the axial force N to be as close as possible to the section centroidal axis, thus ensuring a

488


nearly uniform axial stress distribution over the total height and hence a resin film of constant thickness.

Permanent Assembly: Structural Importance of Epoxy Resins As regards the final prestress tendon profiles, it was shown in Chapter 4 how the resistance of the different cantilevers is ensured by a first group of tendons, known as cantilever tendons, which may be straight or curved in profile and anchored on the various segment faces. The stressing operations remain in the critical path of construction because a new pair of segments cannot be placed before the last pair has been stressed to the existing cantilever, Figure 11.29. The second group of tendons joins the different cantilevers together and makes the structure COIItinuous. They are anchored either in block-outs in the bottom slab or in the fillets at the junction between the top slab and the webs after upward deviation to top slab level. The service shear forces that act upon the joints vary according to the type and characteristics of the structure. In variable-depth bridge decks with draped prestressing tendons the shear stress across the joints is usually low. In a long-span, constantdepth bridge deck with straight tendons, however, the shear stresses at the joints can exceed 600 psi (4 MPa), as was the case in several structures mentioned in Chapter 4. A bad choice, or improper use, of the epoxy resin can be a critical factor concerning the shear resistance of the joints, and for this reason joints of this type require strict quality control. In general, the different types of epoxy resins available have final strengths substantially exceeding that of concrete, so they do not constitute a weak point in themselves. Several conditions must

be satisfied, however, in order that the resin cure properly. 1. Mixing the constituents in their correct proportions. 2. Eliminating any solvents that have a fatal effect on the propertles of the resin. 3. Avoiding any flexible additives, such as thiokol, that greatly increase the deformabilit) of the epoxy. 4. Mixing and applying carefully. With respect to the last point, the surfaces to be joined must be specially treated if the best results are to be obtained. Comparative tests have shown that sand blasting gives the most satisfactorv I-esuits, the surfaces being kept clean. dry, and free from grease during placing. I n damp or rain! weather alcohol is burnt on the joint surfaces to eliminate surface moisture. The water present in the concrete itself has no detrimental effect on the performance of the resin. It has also been established that rapid placing ot successive segments has a favorable effect on the properties of the resin. The additional compressive stress applied to an epoxy joint under polymeriLation when the next segment is prestressed improves the resin’s ultimate mechanical properties. Finally, note that in variable-height structures the joint detailing is such that the joint plane is not normal to the principal stress, especially at the bottom slab level. The epoxy joint is then subjected to shear forces that may be quite large and that can cause failure of the bottom slab in the event of nonpolymerization of the epoxy resin. In addition to the precautions taken to ensure correct curing, one may provide against the risk of bad results by including shear keys in the bottom slab.

FINAL SEGMENT ASSEMBLY 1 I S.2

DETAIL

A

FIGURE 11.29. Final segment assembly.

SECOND-GE,\‘EKA

TI0.V SEC;.LlE.VTS

Although the characteristics and performance of the first structures built with match-cast joints are not in doubt, it seems a good idea to investigate new types of joints allowing the transmission of shear forces without relying on the strength of epoxy resins. Second-generation segments do just this, being equipped with interlocking keys in the top and bottom slabs and in most of the height of the webs. This configuration of shear keys at regular intervals, which improves the behavior of joints bv relieving the epoxy of its structural role, has the


advantages of simplicity and safety. This type of segment has been used with success in several bridges, notably the Alpine Motorways, the Saint Andre de Cubzac Bridge, and the Sallingsund Bridge, and more recently in several structures in the United States such as the Long Key and Seven Mile bridges in Florida. Ribs and Interior Anchorage B1ock.s Anchorage blocks (blisters) or stiffening ribs are currently used inside the segments for the final longitudinal prestress anchors. The tendons, ensuring the stabilit\, and resistance of the cantilever and placed progressively as construction proceeds, can be anchored away from the joint faces, thereby rendering the stressing operations and the segmentplacing operations independent of one another. The ribs and anchorage blocks are generally used to house the temporary prestress that ensures the provisional stability of the cantilever, thus leaving the top slab completely free. Bolted Ribs Despite the tensile strength of the epoxy resin at a glued joint, no tensile resistance is usually considered, as precast segmental structures are nearly always totally prestressed and so no tensile stresses can develop across the joint. However, we can further improve epoxied match-cast joints by giving them a certain resistance to tension by using bolted ribs, which ensure the continuity of the longitudinal reinforcing steel, Figure 11.30. 11.5.3

EPOXY FOR JOAVTS

The structural importance of the thin layer of epoxy resin forming the joint between two adjacent precast segments was discussed in Section 11.5.1. We now take a closer look at the physical and mechanical properties of these resins and the various precautions to be taken to ensure satisfactory and consistent results. -. Epoxy Types Epoxy resin glues are made up from two components: the epoxy resin and the hardener. Mixing these two components in the correct proportions gives a thermostable product

489

with properties that depend upon the type of resin and hardener used. Three grades of epoxy resin are commonly used, depending upon the ambient temperature range under which the resin is to be applied: 40 to 60°F (5 to 15°C) 60 to 75°F (15 to 25°C) 75 to 105°F (25 to 40°C)

Fast-reacting epoxy Medium-fast-reacting epoxy Slow-reacting epoxy

1. Color The resin and the hardener must be of clearly contrasting colors thus avoiding any confusion. When properly mixed, the final product is to be a homogeneous gray color similar to that of concrete. 2. Shelf life of components Both components may be stored for up to one year, provided that the storage temperature is kept between 50 and 70°F (10 and 20°C). After three months’ storage it is necessary to check that the epoxy resin shows no sign of becoming crystalline. If it does, then special treatment must be given to the resin, followed by tests, before use. 3. Pot Life of the Mixed Glue The pot life of an epoxy resin is a measure of the time interval between the mixing of the components together and the moment when the glue becomes no longer workable. The workability of the glue is determined by its internal temperature, depending upon the grade of epoxy resin employed. For a 10 lb (5 kg) mix used on site, mixed under isothermic conditions until an even color of mix is obtained, the following results are required:

Epoxy

Workability Limit Temperature

Grade

5 to 15°C 15 to 25°C 25 to 40°C

40°C ( 104°F) 40°C ( 104°F) 55 to 60°C

(131 to 140°F) The pot life must be approximately: Ambient Temperature

Epoxy

Grade

5 to 15°C

1 5 t o 25°C 25 t o 40°C

41°F (5°C)

( 10%)

50°F

40 min.

15 min.

59°F ( 15°C)

68°F (20°C)

2 0 min.

1 5 min.

95°F

86°F (30°C)

(35°C)

2 5 min.

1 8 min.

490


BOLTED RIB JOINTS

I

FIGURE 11.30.

On site, each 10 lb (5 kg) mix of epoxy resin must be applied to the concrete surface within the potlife period as specified above. 4. Open Time of the Applied Epoxy Glue The

open time of the glue is defined as the period between its application to the concrete surface and the moment when it reaches its workability limit temperature. Because of the much greater heat dissipation from the thin layer [& to a in. (1 to 3 mm)] on the concrete surface, the applied glue takes much longer to reach the workability limit temperature than the mix in the pot. The open time must never be less than one hour, regardless of the grade used. One measuring device used to determine open time is the Vicat’s needle shown in Figure 11.3 1. A 1 mm layer of epoxy glue is spread onto a steel plate, and the stopwatch is started. The time lapsed before the needle will penetrate only 0.5 trim into the glue layer is defined as the open time. 5. Thixotropy This characteristic gives an indication of the epoxy resin’s ability to be applied to vertical surfaces with relative ease and yet with subsequent running. Thixotropy may be measured using Daniel’s gauge, Figure 11.32. The gauge is placed on a level surface with the gutter section horizontal. The gutter is then filled with freshly mixed resin and hardener and abruptly turned to the upright position, as shown in the diagram. The flow time relationship is recorded. The test should be carried out at the maximum temperature for which the resin is specified. A resin that flows less than 30 mm in 10 minutes-is suitable for application to vertical concrete surfaces. Other testing methods are available such as the sag flow apparatus according to ASTM D2730-68.

Bolted rib joints.

Other characteristics of the epoxy glue that may be tested on site are: The ease with which the excess resin may be squeezed out of the joint when subject to uniform pressure.

The angle of internalfriction:

FIGURE 11.31. Open-time testing-Vicat’s needle.

491

Characteristics of Precast Segments and Match-Cast Epoxy Joints ,4X4cm,

s’

/ 100 mm FIGURE 11.32.

.I‘hisotropy testing-Daniel’s gauge.

FIGURE 11.33.

Shear-resistance

test

rate and solubility in water:

ceptable ultimate shear stress at the interface is 1400 psi (10 MPa).

Maximum permissible true water absorption12%. Maximum permissible quantity of epoxy soluble in water at 25°C (i7”F)--4%.

2. Shear Modulus The instantaneous shear modulus (Ci) must be greater than 220,000 psi (1500 MPa) at:

Minimum required value according to Mostens (DIN 53458) on week-old 10 x 15 x 120 mm test rods is 50°C (122°F).

15°C (59°F) for grade 5 to 15°C 25°C (77°F) for grade 15 to 25°C

,Mechu rricnl p)-opertie.)

40°C (104°F) for grade 25 to 40°C

1. Shenr resistance The shear resistance of the mixed epoxy glue is determined on rectangular concrete test specimens with the following dimensions: 1.6 x 1.6 x 6.3 in. (4 x 4 x 16 cm) with a resin interface at 17” to the vertical, Figure 11.33. The concrete test pieces are made from a highquality concrete comparable to that used in precast segment manufacture and are cured under water seven days f-rom time of casting. After removal from the water the pieces are dabdried and the surfaces to be assembled are prepared by shot blasting, wire brushing, or other similar methods to remove laitance. The test pieces are then resubmerged in water for three hours, after which they are removed and dabbed dry with a clean cloth. The resin is then applied in a layer of & in. (2 mm) on one surface and the test beam clamped in an assembly that maintains a normal pressure on the interface of 2 1 psi (0.15 MPa). The assembly is stored for seven days at a temperature representative of the desired working conditions, and then the test is carried out. The minimum ac-

The long-term shear modulus must be greater than 14,500 psi (1000 MPa) after 28 days at the same temperatures as above. Solid cylindrical test pieces are used for measuring these values in conjunction with the easily made test apparatus shown in Figure 11.34. Certain epoxy resins show an excessive sensitivity to high temperatures that makes them unacceptable in warm climates. Figure 11.35 shows comparative results of ten different resins tested for the Rio Niteroi Bridge. It is obvious that a product that becomes practically plastic with no shear modulus at 60°C is completely unacceptable. 3. Tensile Bending Strength A three-point bending test is carried out on a pair of glued concrete cubes with a compressive strength of 5700 psi (400 kg/cm2), Figure 11.36. The faces to be glued are shot blasted, or bush hammered, so as to remove laitance. The cubes are then submerged in water for 72 hours. When taken out of the water the surfaces to be glued are dried simply by dabbing with a clean cloth. Immediately after the dab

Shrinkngp:

Must be practically nil.

Water absorption

Hen t resista rice:


492

Dial gauge,

I

I x1

View X-X

View from one side

FIGURE 11.34. Shear=nioclulus

test.

G (M.&) 2500

2000 1500

1000

500

0

20

30 FIGURE 11.35.

40

50

Variation of’ shear ~nodulus

drying the glue is applied in a layer of & in. (1.5 mm) to one of the prepared faces. The corresponding face of the other cube is placed against the glue layer, and the two cubes are clamped together with a clamping force of 300 lb (150 kg). The assembly is then wrapped in a damp cloth, which must be kept wet until the three-point bending test is carried out. 4. Compressive S t r e n g t h The compressive strength is determined according to DIN 1164 on 4 cm (14 in.) cubes of cured epoxy glue. After 24 hours (from the time of preparing the samples) at the maximum temperatures for each grade the compressive strength must be not less than 12,000 psi (80 MPa). The loading rate is to be approximately 3600 psi (25 MPa) per minute.

5. Elastic modulus in compression The instantaneous modulus (Ei) is determined on cubes of pure epoxy after curing for seven days at the

60

m

t (‘Cl

G with temperature.

maximum group temperature. These cubes are the same size as those used for the compressivestrength determinations. The modulus must not be less than 1,140,OOO psi (7850 MPa)

Practical Use of Epoxy in Match-Cast Joint.\

In

regard to the use of the resin, the two components should be mixed carefully and quickly as near as possible to the surfaces to be coated. Under no circumstances should oil or grease be allowed to come into contact with surfaces that are to be glued. Most standard demolding agents are suitable for use, but care should be taken to ensure that no oil-based demolders are used. Exposure to weather during the storage period is often sufficient to remove the demolding agent. For best results, surface laitance should be removed by shot blasting or bush hammering. This treatment is normally carried out in the storage yard. With the use of multiple keys, the structural role of the

Manufacture of Precast Segments ,L o a d

amlied

here

,

,

p--j -~ -___ FIGURE 11.36. Tensile bending-strength test.

epoxy is considerably reduced and a special preparation of the surface is not a mandatory feature. Immediately before the glue is applied, the surfaces are to be cleaned to remove traces of dirt, grease or oil, and dust. Under normal climatic conditions it will not alwavs be possible to avoid dampness on the surfaces to be glued. If the surfaces do show signs of moisture, they must be dab dried with a clean cloth, and no gluing may proceed until all free water has been eliminated. The thickness of the glue layer should be about h in. (1.5 mm). As soon as possible after the resin has been applied, the surfaces must be brought together. Pressure must be applied before the open time of the epox) resin expires. The pressure applied by either temporary or final prestress should not be less than 30 psi (0.2 MPa).

11.6

Manufacture

of

Precast

ing along the bed for the successive casting operations. 2. Short-line casting (with either horizontal or vertical casting), where segments are manufactured in a step-by-step procedure with the forms maintained at a stationary position. For match-cast joint structures, the accuracy of the segment geometry is an absolute priority. Adequate surveying methods and equipment must be used to ensure an accurate follow-up of the geometry and an independent verification of all measurements and adjustments. Immediately after the manufacture of a segment the as-cast geometry should be controlled and compared to the theoretical geometry to allow any necessary adjustment to be incorporated in subsequent casting operations. This aspect of match casting is particularly important for the short-line method and will be covered later in this chapter.

Segments 11 h.2 LONG-LINE CASTLX’G

11.6.1 1,VTRODUCTION

The various methods used until now for precasting segments fall into two basic categories: 1.

Long-line casting, where all segments to make up either half or a full cantilever are manufactured on a fixed bed with the formwork mov-

In this method all the segments are cast, in their correct relative position, on a casting bed that exactly reproduces the profile of the structure with allowance for camber. One or more formwork units travel along this line and are guided by a preadjusted soffit. With this method the joint surfaces are invariably cast in a vertical position.

494


Figure 11.37 shows the casting sequence.3 The pier segment (3) is cast first, then the segments on either side of the pier segment (1) and (2). If a pair of forms is used, then the symmetrical segments on each side of the pier segment can be cast simultaneously, thus saving casting time. As segment casting progresses, the initial segments may be removed for storage, leaving the center portion of the casting bed free. If enough forms are available, then the casting of a second pair of cantilevers may proceed even though the first pair is not completely cast. Figure 11.38 shows the typical cross section of a long-line casting bed with the formwork in operation. The method was initially developed for constant-depth box girders (Choisy-le-Roi and Courbevoie Bridges). It was later extended to the case of variable-depth decks such as the Oleron Viaduct (the two sketches of Figures 11.37 and 11.38 refer to this structure) and also adopted in other countries (Hartel Bridge in Holland). The important advantages of the long-line casting method are: It is easy to set out and control the deck geometry.

After form stripping, it is not necessary to immediately transfer the segments to the storage area in order to continue casting. The disadvantages are: Substantial space may be required. The minimum length is usually slightly more than half the length of the longest span of the structure, but it depends upon the geometry and the svmmetrv of the structure. The that the line

casting bed must be built on a firm foundation will not settle or deflect under the weight of segments. If the structure is curved, the long must accommodate this curvature.

All equipment necessary for casting, curing, and so on must be mobile. 11.6.3 SHORT-LIIVE

HORIZOh’TAL

The short-line casting method requires all segments to be cast in the same place, using stationary forms, and against the previously cast segment in order to obtain a match-cast joint. After casting and initial curing, the previously cast segment is

Segments completed.

Segments being cast

FIGURE 11.37. Typical long-line precasting bed. Travelling crane leg

Mobile outside r formwork

II

FIGURE

11.38.

CASTI.YG

Telescopic inside f-form work

Typical cross section of long-line casting bed with formwork.

495

Manufacture of Precast Segments

removed for storage and the freshly cast segment is moved into its place. The casting cycle is then repeated. This operation is illustrated in Figures 11.39 and 1 1.40.“*4

BLANK END

TO

STORAGE

/

It is important that the reader fully comprehend the principle of the method insofar as building a deck of a given geometry is concerned. When a straight box is desired, Figure 11.41, the match marking mate segment (n - 1) is moved from the casting position to the match-cast position along a straight line, and this is usually verified by taking measurements on four elevation bolts (a) embedded in the concrete roadway slab and two alignment stirrups (b) located along the box centerline. A pure translation of each segment between the cast and match-cast positions therefore results in the construction of a perfectly straight bridge (both in elevation and in plan view), within the accuracy of the measurements made at the casting site. To obtain a bridge with a vertical curve, the match-cast segment (n - 1) must first be translated from its original position and then give a small rotation in the vertical plane (angle CY shown in Figure 11.42). Usually the bulkhead is left in a fixed position, and all segments have in elevation the shape of a rectangular trapezoid with the tapered face along the match-catch segment. It is therefore only necessary to adjust the soffit of the cast segment during the adjustment operations.

FIGURE 11.39. t i on.

‘I‘ypical short-line precasting opera-

A curve in the horizontal plane is obtained in the same fashion, Figure 11.43, by first moving the match-cast segment (n - 1) to its position by a pure translation followed by a rotation of a small angle p in plan to realize the desired curvature.

TO

STORAGE

4

/,

FIGURE 11.40. Formwork used in casting segments.

ELEVATION

TRANSVERSE

STRAIGHT

SECTION

BRIDGE

PLAN VIEW

FIGURE 11.41. Straight bridge.

ELEVATION

_------e-------A-

TRANSVERSE SECTION

m-e----1-------;

BRIDGE WITH VERTICAL CURVE PLAN VIEW

FIGURE 496

11.42.

Bridge with vel-tical curve.

497


TRANSVERSE SECTION

ELEVATION

BRIDGE WITH HORIZONTAL CURVE PLAN VIEW

FIGURE 11.43. Bridge with horizontal curve.

Change in the superelevation of the bridge may also be achieved with a short-line casting; however, the principle is a little more difficult to properly grasp, Figures 11.44 and 11.45. A constant transverse fall of the bridge does not need to be repeated in the casting machine. Segments may be cast with soffit and roadwav slab both horizontal and placed at their proper attitude in the bridge by offsetting the bearing elevation under the webs to obtain the desired cross fall. Only a variable superelevation must be accounted for in the casting operation, and this is the normal case in bridges with reverse curves and in transition areas between curves and straight alignments. In such a case match-cast segment (n - 1) needs to be rotated by a small angle such as y around the bridge centerline. Because the bridge geometry is usually defined at roadway level and not at soffit level, the rotation given to the match-cast segment results in a slight horizontal displacement of the soffit in the casting machine, which must be accounted for. Also all surfaces of the box segment (top slab, soffit, and webs) are no longer true planes but are slightly warped. To allow the formwork panels to adjust to this change of shape, it is absolutely mandatory to eliminate all restraints such as closed torsionally stiff members.

The basic advantages of the short-line casting method are therefore the relatively small space required and the fact that all equipment and formwork remain at a stationary position. The mobility of equipment necessary for the long-line method is no longer needed. Also, horizontal and vertical curves as well as variable superelevation are obtained with short-line casting without the major change in soffit configuration that would be required in the long-line casting method. However, success will depend upon the accuracy of adjustment of the match-cast segments, and precise survey and control procedures must be initiated (Section 11.6.5). This last aspect represents the major potential disadvantage as a direct consequence of the intrinsic potential of the method. 11.6.4

SHORT-LINE

VERTICAL

CASTING

Normally, for both the long- and short-line methods, the segments are cast in a horizontal position. A variation in the short-line method is that used for the Alpine Motorways near Lyons, France, where the segments were cast in a vertical position (cast on end) as shown in Figure 3.100. The procedure is as follows: after the first segment is cast, the forms are removed and moved upward

498


ELEVATION

TRANSVERSE

SECTION

BRIDGE WITH VARIABLE SUPER ELEVATION

PLAN VIEW

FIGURE 11.44. Short-line casting-bridge with variable super-elevation.

\

END BULKHEAD

FIGURE 11.45. Short-line casting-isometric view of segment casting with variable superelevation.

so that each succeeding segment can be cast above the previous one. After a segment is cast and cured, the segment beneath it is transferred to storage and the one removed from the forms is moved down, to rest on the floor. The advantages

claimed for vertical match casting include easier placing and vibration of the concrete. However, special handling equipment and procedures are required to rotate the segment from the vertical to its final horizontal position.

Manufacture of Precast Segments 11.6.5

499

GEOMETRY AND SURVEY CONTROL

Segment Precasting in a Casting Machine

The principles described in this section apply to short-line horizontal casting but may be easily extended to vertical casting. The apparatus used to form the concrete segment is usually referred to as a casting machine and is made up essentially of five components: 1. 2.

3. 4.

5.

The bulkhead that forms the front section of the segment. The match-cast segment, properly coated at the front end section with a suitable demolding agent and used to form the back end section of the newly cast segment. The mold bottom (or soffit). The side forms, properly hinged for stripping and firmly sealed to the bulkhead and the match-cast segment during casting. The inside forms, which pivot and retract for stripping. The inside forms, which piv.ot and retract for stripping.

The relationship between an individual segment and the finished structure is established by means of three different systems of reference: The final system of reference, which is the reference for the finished geometry of the structure. In this system each segment is described by its basic geometry. 2. The auxilia? system of reference, which corresponds to the precasting machine and is attached thereto. 3. The elementa reference system, which is attached to each segment and would be the equivalent of intrinsic coordinates in space geometry.

1.

The principle of the precasting method is as follows. During the casting of segment A (segment B being in the match-cast position) the elementary reference system of A is identical with the auxiliary reference system, that of the casting machine. To position B with respect to A becomes simply a matter of positioning B with respect to the precasting machine. It is the task of the design office to provide the theoretical geometric information necessary for positioning. The values are computed from the basic geometry with the addition of the relevant compensatory values for deflections. The definitions of these reference systems are presented below.

FIGURE 11.46. Auxiliary reference system (castingmachine reference). The auxiliary reference system refers to the casting machine and is defined in Figure 11.46. The plane of the bulkhead is perfectly vertical. The upper edge of the bulkhead is a horizontal in this plane except when segments do not have planar top surfaces. The x, y and z axes refer to the casting-machine reference system, whereas XA, yA9 and z,., refer to the elementary system of reference. The elementary system of reference is materialized on each segment in the following manner: 1

2.

3.

This axis is represented by marks (such as saw cuts) made on two steel stirrups anchored in the top slab as near as possible to the joints. The origin 0,: The origin o, is located at the point where the x, axis intersects the plane of the joint at the bulkhead. This plane may be defined Theplunex,,o,,y,: by three fixed leveling points, the position of each point with respect to the plane x, o, y being arbitrary but invariable. For practical reasons, four leveling points are used and materialized by bolts anchored in the top surface of the segment above the webs and as close as possible to the joints.

The x, axis:

Now that the elementary system of reference has been established (all measurements and readings being made while the segment is in the casting machine before the forms are removed), the segment can be positioned with respect to the auxiliary reference system, so that it can be placed in the correct countercasting position according to the calculations supplied by the design office.

500


In order to correctly position the countercasting segment, information is needed about the final geometry of the structure. The overall geometry of a bridge structure is normally defined by the geometry of the roadway. From this roadway geometry it is necessary to determine the geometry of the concrete structure itself. The longitudinal reference line to which all the necessary parameters are related is known as the box girder line (BGL). This line may coincide with the top concrete surface of the box girder, but it may also be a fictitious line of reference if the box girder top slab shape is not regular. The box girder line is usually described using two curves, Figure 11.47:

zontal plane and follow the curvilinear abscissas. The segment lengths chosen on this basis may be retained, but in calculating the real lengths of cast-in-place closure joints and three-dimensional s curve must be used. Because of the way a casting machine works, the segment joint at the bulkhead end is invariably perpendicular to the axis of the segment. Therefore, in plan view, the segments are generally of trapezoidal shape, except for segments over the piers which are rectangular in order to provide a constant starting point for each cantilever, Figure 11.48.

One curve (a) in a horizontal plane, which gives y as a function of x for each point where the box girder line intersects a joint plane between segments and also the center points of supports (abutments or piers); this curve is simply the projection of the true space box girder line onto a horizontal plane and is sometimes referred to as the “bgl” (small letters). One curve (5) in a developed vertical plane giving z as a function of c for the same points mentioned above. Thiss curve is the real box girder line, BGL.

All measurements on a segment are made when the segment is still in the casting machine. Readings must be taken when the concrete has hardened and before formwork stripping, Figure 11.49. Horizontal alignment readings give the distance of the segment axes as marked on the stirrups from the casting-machine reference line. Longitudinal profile level readings are given by the four bolt elevations relative to the horizontal reference plane. Readings must be taken on the segment just cast and also on the match-cast segment. Corrections are applied to allow for the geometric defects in the preceding segment, Figure 11.50, and are used as “theoretical values for adjustment.”

To complete the definition of the segment position in space-at each joint line and at support centers-we must define the transverse slope of the theoretical extrados line. It is important for both the bgl and the BGL to calculate the m and s parameters, respectively, in order to obtain an accurate determination of projected and real span lengths. The calculations and structural drawings refer to nominal segment lengths and span lengths. Usually these lengths refer to the projection on a hori-

Box girder line

Horizontal pqcction of box girder line l bgl’

FIGURE 11.47. Box girder line curves.

Segment Casting Parameters

Survqr

Control During Precasting Operations

The surveyor in charge of the operations must complete a data sheet for each segment containing essentially: 1. Theoretical basic data supplied by the design office, allowing the preparation of the horizontal alignment and the two parallel bolt lines. 2 . Corrected values defined either graphically or by computer. 3. Survey control readings. 4. Linear measurements on the segments. 5. Schematic representation of the segment to rapidly verify the relative positions of the segment axes. 6. A level check to pick up any gross error in level readings on the same segment. 7 . Comments on the casting operations.

501


( Segment ax is Segment

over

/

Hj+vi*3

pier Pier f

\

- BOX girder line “bgl* or Q ( sigma ) curve

FIGURE 11.48.

Short-line casting-position of segment joints in plan view.

Ew

i’dw

L

!Esw

1dsw . A!

?-r

L3

FIGURE 11.49. Casting-machine orientation and segment

measurements.

As an example, Figure 11.5 1 shows the typical survey control made on the first four segments of a typical cantilever. Control of alignment and levels may be followed graphically or numerically by computer, using the basic geometric data obtained in the casting machine and shown in Figure 11.52. In order to avoid any significant deviation from the theoretical geometry, it is necessary to provide

for corrections when casting the next segment. Figure 11.53 shows how this would be done for the plan alignment. Similar corrections are made for the longitudinal profile on the two parallel bolt lines. It is essential not only to follow carefully the trajectory of the two bolt lines separately but also to check for each segment that the superelevation (given by the crosswise difference in level between


502

FIGURE 11.50. Plan view of’ casting operationreadings using survev instruments.

FIGURE

11.51.

Casting

operation-topical

the two bolt lines) varies regularly according to the theoretical geometry. Failure to do so has resulted in important geometric imperfections on certain projects. Suruey

Control During Construction

The nature of match-cast segmental construction is such that the structure is really “built” in the precasting yard. Although corrections can be made in the field, such corrections are undesirable and always a source of additional expense and delays. Close control of precasting is far more efficient. It is nevertheless important to check the evolution of the structural geometry during segment placing: 1.

To compare actual deflections with computed

values,

2.

survey

control

To ensure that no major errors have escaped the control in the precast yard or factorv.

Such checks at the site should include: 1 . Pier positions, height and in plan. 2 . Bearing positions, level and orientation. 3 . Pier segments, level and orientation. 4. Cantilevers proper, every third segment, including levels, superelevation, and orientation. 5. Overall geometry of the structure after continuity is achieved between the individual cantilevers. Conclukon

The principles of geometry and survey control are more complicated to explain than to use, once the

503

Manufacture of Precast Segments THEORETICAL AX15

OFO,

THEORETICALAXIS

OFO,

cAsTI Nt M~CI~NE IHG

SEGMCNTO

’

BULKHEAD

QGMENT~

t

_ _

MACHINE

TUEORETltnL

-;--- -.--~~Z?-~

LEVEL FUR 1

+L;W’: ;R yyoE;; = .

THEORETICnL, LEVEL

FOR

0 R E A L LEVELFOR SEGMENT 0

055TING M A C H I N E EUJLKHMD

sEGMENTI

FIGURE 11.52. Survey control-horizontal alignment and longitudinal profile results. ((I) Horizontal alignment. (h) Longitudinal profile.

REAL AXIS OF 1

TUE ORETICAL AXIS OF 0 (2 REAL AXIS)

FIGURE

11.53.

Typical alignment corrections during casting operations.

basic principles of a casting machine are thor-

oughly understood. The short-line method has great potential to construct segments for bridges, even those with very complicated trajectories, rapidly and economically. Outstanding examples

are the Chillon and St. Cloud Viaducts in Europe and Linn Cove Viaduct in the United States. At Saint Cloud, 120- to 140-ton segments were cast on a one-day cycle, and the final geometry of the bridge was obtained with no on-site adjustment.


504

On the other hand, a loose approach to geometry control at the casting yard may lead to serious difficulties at the project site. 11.66 PRECASTING YARD AND FACTORIES The precasting operations are usually carried out in a yard or even a factory if the size of the project allows the corresponding investment. All operations, such as: Preparation of the reinforcing steel cages and ducts for post-tensioning tendons Manufacture

of

concrete

Manufacture of segments including heat curing Storage of segments including finishing and quality control are performed in a repetitive fashion under factory conditions. As an example of typical precasting-yard layouts, Figures 11.54 and 11.55 show views of: The Saint Cloud Viaduct precasting yard with short-line casting The Oleron Viaduct precasting yard with long-line casting The typical precasting cycle (with either the long-line or the short-line method) is of one segment per formwork per day with a one-day work shift, concrete hardening taking place during the night (at least 14 hours between the completion of concrete placing in the evening and the stripping of forms the next morning). Shorter construction cycles may be obtained by reducing the time of concrete hardening, but quality may decline if all the operations are not kept under very strict control. Heat curing of the concrete to reduce the construction cycle and accelerate the rotation of the casting machines is perfectly acceptable. Its improper use, however, may alter the accuracy of joint matching between segments, as shown in Figure 11.56. This effect would be particularly significant for wide but short segments. Typical segments usually have the following dimensions: Width Length Ratio width/length

30 to 40 ft (9 to 12 m) 10 to 12 ft (3 to 3.6 m) 3 to 3.5

In the case of wide decks or long spans, where the segment length is reduced to reduce the unit weight, the usual geometric proportions may vary considerably; such is the case for two notable structures: St. Cloud St. Andre de Cubzac

width ratio width ratio

70 ft. length 7 ft, 10 58 ft, length 5.8 ft. 10

For such segments, heat curing is more likely to create small changes in the segment shape, which may build up progressively and so alter the effectiveness of joint matching. This is due to the development of a temperature gradient in the match-cast segment, which is in contact on one side with the newly cast heated segment and on the other side with the lower outside temperature. The problem may be completely eliminated by always heat curing both segments simultaneously so as to avoid any temperature gradient. Experience has proved the method totally efficient. When the project involving segment precasting is of sufficient magnitude or where climatic conditions are adverse, precasting factories are a logical extrapolation from the short-line method performed in an open precasting yard. Segment manufacture takes place in a completely enclosed building with a better use of personnel and a more consistent quality of products. An interesting example is afforded by the B-3 South Viaducts, requiring production of 2200 precast segments weighing between 28 and 58 tons (25 to 53 mt). The precasting site was installed close to the project and included four main areas: 1.

An assembly workshop, where the reinforcing steel cages were prepared and the prestressing ducts positioned. The finished cages were handled by a 5 ton tower crane. 2. A concrete mixing plant. 3.

A precasting factory where the segments were cast and cured.

4. A storage area where the finished segments were left to cure adequately. These segments were handled by a traveling portal crane. The precasting factory was equipped with four precasting machines, all of which were entirely protected from the outside environment. Two machines were reserved for the manufacture of 15 to 20 ft (4.5 to 6 m) segments and two for the 20 to

PRECASTING

YARD

S c a l e 11500

1.

Launching

track

for

tr,irder

a n d trolle!y

10

Rplnforcernent

Access ramp.

2

3. Loading

point

Launching zone

5

Segment

6.

Travelllng

I.

hlauld

8

r’rPstress1ng

/ 0% 0 J

for s e g m e n t s

g\ rdet assembly

4.

\

asspmbty

\

storage. crane t r a c k

bottom steel sturaqe

,’ ’ _/’ ,’ ,/’’ /’ ,A’

I/ _ 27 “\ ‘‘\

17 F u t u r e carriageway alignment

FIGURE 11.54. St. Cloud Viaduct, precasting yard layout. (1) Launching track for girder and trolley. (2) Access ramp. (3) Loading point for segments. (4) Launchinggirder assembly zone. (5) Segment storage. (6) Traveling crane track. (7) Mold bottom. (8) Prestressing steel storage. (9) Tower crane track. (10) Reinforcement assembly. (11) Concrete plant. (12) Precast elements. (13) Prestress tendon manufacture. (14) Offices. (15) General services. (16) Toll gate position. (17). Future carriageway alignment.

:‘i

Technology and Construction of Segmental Bedges

506

Staff

quarters

Launching

girder

assembly

Retnforcement

i Duct storage area

FIGURE 11.55. Oberon

SEGMENT

L-l LENGTH

L

FIGURE

11.56.

\‘iatiuct,

L

OffIce

precasting yard layout.

CONJUGATE

E&CT OF IMPROPER CURING OF SEGMENTS IN SHORT LINE CASTING

Effect of improper curing of segments in short-line casting.

31 ft. (6 to 9.5 m) segments, Figures 11.57 and 11.58. Each casting machine was made up of a mobile form, an end form or bulkhead, two hinged outside forms, and a telescopic inside form, Figure 11.59. Handling of concrete and reinforcing steel inside the factory was performed by two 10 ton travel cranes.

The production of the different segments involved the following operations: 1. Assembly of the steel cages in a template. 2. Steel-cage storage. 3. Final steel-cage preparation and duct installation.

Handling and Temporary Assembly of Precast Segments

FIGURE 11.57.

Casting

R-3 Sot~rh

507

Viaducts, inside view of the precasting f’acto~~~.

machine

Concrete

Plant

Control

System /

FIGURE 11.58. Inside

B-3 South Viaducts, plan view of the precasting factory. formwork

t

O u t s i d e formwork

FIGURE 11.59.

Bottom

formwork

B-3 South Viaducts, detail of a casting

machine.

4. Positioning of steel cage inside the formwork. 5. Adjustment of casting machine, including alignment of match-cast segment and sealing of all form panels. 6. Concrete casting and finishing. 7. Steam curing. 8. Formwork stripping; followed by transfer of the match-cast segment to the storage yard and

of the newly cast segment to the match-cast position by means of an independent motorized trolley.

1 I .7 Handling and Tempera y Assembly of Precast Segments In either long- or short-line casting, segments cannot be handled before the concrete has reached a sufficient strength to prevent: Spalling of edges and keys Cracking of the parts of the segment subjected to appreciable bending stresses due to self-weight Inelastic deformations that would ultimately impair proper matching of the segments Critical sections in a typical single-cell box segment are, Figure 11.60:

508


I FIGURE 11.60. Critical sections in a typical segment at time of formwork stripping.

Section A where the side cantilevers are attached to the webs Sections B and C at midspan of the top and bottom slab Section A is almost always the most critical. Section B is usually subjected to moderate tensile stress because the top slab is built-in on the web when the inner formwork is stripped. Section C is critical only on long-line casting when the casting bed does not have a continuous soffit and when the span of the bottom slab is larger than 16 to 20 ft (5 to 6 m). Experience has shown that at the time of form stripping and before any handling of the segment is allowed, the tensile cracking strength of the concrete should be at least equal to the bending stress due to the segment weight in the most critical sections (A, B, and C). Practically, the corresponding compressive strength is: f:i = 3000 to 4000 psi (21 to 28 MPa) In the casting yard, segments are usually handled by a portal crane traveling on rails or on steering wheels for added mobility. A typical portal crane in the Oleron Viaduct precasting yard is shown in Figure 11.61. Proper handling of the segment requires proper pick-up points to keep the stresses in the section within the allowable limits. A typical example of handling three different shapes of box girders is shown in Figure 11.62. For the conventional single box, inserts or through holes are provided near the web in the roadway slab, allowing lifting to be accomplished by a simple spreader beam.

FIGURE 11.61. Oleron Viaduct, portal crane in precasting yard. For the twin-box, three-web section, a four-point pick-up is usually necessary to eliminate excessive transverse bending of the top and bottom slab. A triple spreader-beam arrangement allows the load transfer from the four pick-up points to the single lifting hook. For a triple-box, four-web section (such as used in the Saint Cloud Bridge), temporary ties are provided in the outer cells to transfer the reaction of the outside webs to the center webs. A simple spreader beam is then sufficient to lift the segment. Segments must be stored in a manner designed to eliminate warping or secondary stresses. Concrete beams installed at ground level provide a good bearing for the segments, which must be supported under the web or very close thereto. If stacking is required to save storage space, precautions must be taken to transfer weight from the

Placing Precast Segments

Slmnle

509

spreader beam

Handling precast segments. (a) Two-web segment. (6) Three-web segment. (c) Four-web segment.

FIGURE 11.62.

upper to the lower layers of segments without excessive bending of the slab. 11.8 Placing Precast Segments

Transportation and placement of segments may be performed by one of several methods, depending on the site location and the general characteristics of the structure. These methods can be divided into three main categories: 1. Transportation by land or water and placetnent by an independent lifting apparatus. 2. Transportation by land or water and placement with the help of a beam and winch carried by the bridge deck itself. 3. Transportation by land, water, or along the bridge deck already constructed and placement with the help of a launching girder. There are tnethods that fall into none of these categories, such as the use of a cableway, but their use is limited. 11.8.1 INDEPENDENT LIFTING EQUIPMENT

This method, where feasible, is the simplest and least expensive. It was used for the Choisy-le-Roi, Courbevoie, Juvisy, and Conflans bridges, where the navigable stretch of water lent itself to the use

of a barge-mounted crane, ensuring the collection of segments from the precasting site and their positioning in the final structure. A terrestrial crane was employed for the Gardon, Bourg-Saint Andeal, and Bonpas Bridges. The same crane, maneuvering either on land or over water (on a barge), assured the positioning of all the segments used to construct the upstream and downstream bridges of the Paris Ring Road. When site conditions are suitable, the same lifting crane may be used both to serve the precasting yard and to transport the segments to their final position in the structure (Hartel Bridge, Holland). This principle was enlarged successfully during the construction of the bridges over the Loire River at Tours (Motorway Bridge and Mirabeau Bridge), where the segments were placed with the aid of a mobile portal frame. The portal frame is placed astride the bridge deck and moves along a track supported by two bailey bridges, one either side of the structure. The track length is approximately twice that of the typical span, and the track itself is moved forward progressively as construction proceeds. The bailey bridges are supported on temporary piers driven into the river bed. The segments are first brought to the bridge deck and then taken by the mobile portal frame, which transports them to their final position in the finished structure, Figure 1.47. Where a mobile truck or crawler crane is used for placement, there are often difficulties in the

510


positioning of the key segments at midspan, because the finished structure on either side of the key segment prevents the crane from maneuvering properly and hinders the positioning of the segment, which may be carried out only from the side of the structure. For the B-3 Motorway Bridges a special apparatus was designed to place those segments in the cantilever arm to be constructed in the direction of the completed structure, Figure 3.95. Two longitudinal girders are braced together and rested on the pier head of the cantilever to be constructed at the front, and on the existing structure at the rear. The apparatus consists of a mobile winch-trolley, ensuring the hoisting and positioning of the segments, and an advancing trolley situated at the rear and equipped with a translation motor. The front and rear supports are conceived in such a manner as to transmit the vertical loads through the segment webs. The segments on the other side of the cantilever are easily placed by the mobile crane. This beam may easily be used to ensure cantilever stability during construction when the piers are not sufficiently rigid to support unsymmetrical loading. The cantilever is rigidly fixed to the girders by clamping bars capable of resisting both tension and compression. The crane and the girders, used together, will allow a 130 ft (40 m) span to be erected in four working days. Placement of segments with a mobile crane has found another application in the construction of small-span structures such as three-span motorway overpasses (see the discussion of the Alpine Motorway, Section 3.15, and Figure 3.103). The segments are precast in a central factory, transported to the various sites by road and positioned by a mobile crane according to the erection scheme, which consists essentially of the following: Two temporary adjustable props, easily dismountable, placed at the one-fourth and three-fourths points of the central span. Temporary supports with jacks allowing cantilever construction Temporary prestress to tie the segments together before stressing the final prestress Elimination of the classic cast-in-place closure joint by direct junction of the two cantilever arms face to face. Final prestress by continuous tendons instead of cantilever-type layout.

The total construction time for such an overpass, including the piers, usually does not exceed two weeks, of which less than one week is spent on the bridge superstructure itself. This method has been used with great success for the Rhone-Alps motorway overpasses, with spans varying between 60 ft (18 m) and 100 ft (30 m).

11.8.2

THE BEL4iM-ASD-

WI,VCH .klETHOD

The beam-and-winch method of placing precast segments was conceived for the construction of the Pierre-Benite Bridges over the Rhone River. This construction method requires a fairly simple apparatus rolling along the already constructed part of the cantilever and ensuring the lifting, translation, and positioning of all the segments. The apparatus is shown diagrammatically in Figure 11.63. It consists of the lifting gear B carried by the trolle) C rolling along the bridge deck on tracks D. The segment A is brought, bv land or water, beneath the pier in question, where it is lifted bv the equipment. It is then transported to two launching beams E that cantilever out from the bridge deck, upon which it continues to advance until reaching its final position, whereupon it is lifted to its final level next to the previous segment, Figure 11.64. This system can, of course, be simplified if the segment can be brought by some independent means to a location vertically below its final position in the structure. As originally conceived, this system was not completely independent: another construction procedure was required to erect the pier segment. The pier segment was cast in place in the PierreBenite Bridges. It was precast and placed by a crane for the Ampel Bridge in Holland and by a floating barge crane for the Bayonne Bridge over the river Adour. This weakness was eliminated in the construction of the Saint-Andre-de-Cubzac Bridge. For this structure, the pier segments, which form the starting point for each cantilever, were placed by the same equipment that placed the typical span segments, Figure 3.72. The equipment was hung, with the help of cables, to an auxiliary mast fixed to a lateral pier face. The pier segment was brought in from the opposite side, lifted and placed by the mobile equipment’s winches. In the same position the following segment was located and the auxiliary mast removed, Figure 3.73. At this point it was a simple matter to reposition the mobile lifting equipment in order to place the typical span segments, Figure 3.70.

511

recast Segments

GMFNTS NFAR

PPARATUS

RIGHT RANK

TRANSFER

TROLLEY

PIER BLOCK

4ENT

PONTOON

,tream Bridge, placing apparatus.

evolved and how the original concept has been modified. Launching Girders Slightly Longer Than the Span Length We first consider the construction method of the Oleron Viaduct Bridge superstructure, Figure 3.32. The segments were brought along the top slab until they reached the launching girder, then lifted by the latter, transported to their final position, lowered so as to come into contact with the previous segment erected, and prestressed to the cantilever. The launching girder itself, slightly longer than the span length, was made up of a steel trellis beam with an entirely welded rectangular section weighing 124 tons (113 mt) and measuring 312 ft (95 m). The maximum span length of the bridge was 260 ft (79 m). The launching-girder system consists of two fixed supports, called tunnel legs, allowing the segments to pass between them, one at the rear of the girder and the other at the center. At the front end is a mobile prop enabling the girder to find support on the next pier. The bottom chords of the girder are used for the rolling track that supports the segment trolley, which can move the segment horizontally and vertically and rotate it a quar-

512


~23~~118’~141’--ll6’-7” 4 Placlng center segment

118’i141” 259’A Movmg gantry to next pier

Placing segments in doudle cantilever

FIGURE 11.65. Oleron Viaduct, launching-girder operations. (A) Rear support, (B) center support, (C) temporary front prop, (D) prop support, (E) pier segment, (F) temporary support.

ter-turn. Three phases are clearly distinguishable in the construction of a cantilever, Figure 11.65: Phase 1: Placing the pier segment The launching girder rests on three supports-the rear support, the center support near the end of the newly constructed cantilever, and the front prop, which is attached to the front of the next pier with the help of a temporary prop support.

A support adjustment was carried out with the help of hydraulic jacks when the girder was resting on the rear and central supports and the temporary front prop, before installing the pier segment. The purpose of this adjustment was to obtain the optimal distribution of the launching girder selfweight among the three supports. While the fi-ont prop is being installed, the central support rests on the end of the previous cantilever in the same position in which the rear support will be during the erection of the typical segments. In this phase the launching girder rests on two supports and is therefore statically determinate; nothing can be done to change the rear-support reaction. While the pier segment is being placed, however, the girder is resting on three supports and is statically indeterminate. It is therefore necessary to ensure that the reaction at the central support is less than or equal to that which will be produced by the rear support during the next construction stage, including the weight of the trolley and the tractor placed in the near vicinity. Several other structures have been built with launching girders of the same generation as the one used for the Oleron Viaduct. The Chillon Viaduct, Figures 3.43, 11.66, and 11.67, along the bank of Lake Leman used a 400 ft (122 m) launching girder weighing 253 tons (230 mt). The maximum span length was 34 1 ft ( 104 m). The launching girder, of constant rectangular section, was of the suspension type, being suspended at the one-quarter points by cable stays anchored at the central mast, which extended above the level of the launching girder. The supports were hydraulically adjustable, allowing the girder to cope with different angles of superelevation, Figure

Phase 2: Moving the launching girder forward The girder rolls along on the rear support and the segment trolley, which is rigidly attached to a metal framework known as the temporary translation support, which is fixed to the pier segment. The rear and central supports are equipped with bogies and roll along a track fixed to the bridge deck while the girder is being moved forward. Phase 3: Placing typical segments The launching girder rests on two supports, the central support anchored to the pier segment and the rear support tied with prestressing bars to the end of the previously constructed cantilever.

FIGURE 11.66. Chillon Viaduct, launching-girder in

operation.

513


4th &age lauchmg

Girder

FIGURE 11.67. (:hillon

Viaduct, launching-girder movements. 11.68. The launching girder included three means of adjustment: Lateral movement of the trolley in order to place eccentric segments Adjustment 02: Lateral translation of the central support in order to cope with horizontal curvature of the structure [email protected] 03: Vertical adjustment of bogies to take up the superelevation and so keep the central support vertical. Adjustment

D3


adjustments.

Viaduct, launching-girder

Dl:

In order to follow the horizontal curves the launching girder rotated about the rear support while moving sideways across the central support, Figure 11.69. The mobile temporary front prop was conceived in the same way as the other supports so as to allow the passage of the first segments to either side of the pier segment. The Blois Bridge on the Loire River in France had a 367 ft (112 m) long launching girder weighing 135 tons (123 mt), Figure 11.70. The maximum span length was 300 ft (91 m). The launching girder, of constant triangular section, could be dismantled and transported by road. All of the girder components were assembled with high-strength bolts, ensuring the transmission of

514

Technology and Construction of Segmental Bridges C O N S T R U C T I O N O F HOlzIZ0rifA~

CONSTRUCTION OF

UORIZONTAL

CURVC

CURVE

( STAGE 1. )

( STAGE 2 )

FIGURE 11.69. Chillon Viaduct, curved span con-

struction.

ELEVATION

FIGURE 11.70. Blois Bridge, launching girder. SFCTION A

forces by friction between adjoining plates, Figure 11.71. The use of a very light structural steel framework carried with it the risk of large deflections. These were reduced and controlled by two sets of cable stays, passive and prestressed, which came successively into play during maneuvering of a segment (upper passive stays) and during the launching-girder advancement (lower prestressed stays). This launching girder was later used for the erection of two other structures: the Aramon Bridge on the Rhone River, Figure 11.72, and the 2950 ft (900 m) long Seudre Viaduct. The Saint Cloud Bridge on the Seine, Figure 3.78, is a recent example of the use of a large launching girder. The girder could place segments weighing up to 143 tons (130 mt) in spans of up to 335 ft (102 m) with a minimal radius of curvature in plan of 1080 ft (330 m), Figure 3.79. The weight of the launching girder was 260 tons (235 mt) and its total length was equal to 400 ft (122 m). The adjustments adopted were similar to those used for the Oleron, Blois, and Chillon bridges. The launching girder, which used upper passive stays and lower prestressed stays, was constructed

FIGURE 11.71. Blois Bridge, launching-girder as-

sembly detail.

FIGURE 11.72. Aramon Bridge over the Rhone

River.

with a constant triangular section made up of individual elements assembled by prestressing. This launching girder is notable, apart from its assembly by prestress, for its ability to follow extremely tight curves. The movements used for the Chillon Via-


duct were, of course, used for this purpose. However, in the Saint Cloud Bridge it was necessary also for the launching girder to take up several intermediate positions during the erection of a given cantilever so as to bring each segment to its final position in the structure. The total lateral translation reached 19.7 ft (6 m) at its maximum. Construction speed of the bridge deck was 130 ft (40 m) per week, including all launching-girder maneuvers. Two other structures erected with the help of the Saint Cloud launching girder were the Angers Bridge and the Sallingsund Viaduct. The launching girder used for the Alpine Motorway network was conceived for spans and segment weights of more modest dimensions; it is typical of lightweight universal equipment that can be easily dismantled for reuse in another structure, Figure 11.73. This girder allowed the handling of segments weighing up to 55 tons (50 mt) over spans up to 200 ft (60 m). Reflecting on the launching girders mentioned above, we note that their evolution centers on two major characteristics: the structural conception of the girder and the assembly method (connection types, number of elements, and so on). Launching girders tend more and more to be of the lightweight type, relying on exterior forces to cope with different loadings. These exterior forces are provided by the external active cable stays, which allow the structure to be placed in a condition ensuring a favorable behavior under a given loading. This approach to launching-girder design provides more optimal use of materials than did the first-generation girders of variable cross section. Another advantage of a constant cross section is that it facilitates the construction of standard sec-

FIGURE 11.73. Alpine Motorway launching girder.

515

tions that can be interchanged and assembled on site. In this way the girder length can be varied according to the span length and the weight of the segments. Connections are made with tensioned bolts, Figure 11.74, which reduce considerably the number required and consequently the time needed to assemble or dismantle the structure. These connections have recently replaced those made with high-strength bolts and fishplates, notable on such structures as the Deventer Bridge and the B-3 Viaducts. Means of erection adjustments also have improved, tending to reduce the forces applied to the deck itself by ensuring that the girder supports are located over the piers or at least in the very near vicinity. This natural evolution leads us toward a new type of launching girder, one whose total length is slightly greater than twice the typical span length, allowing the simultaneous placing of the typical segments of cantilever N and the pier segment of cantilever N + 1. Launching Girders Slightly Longer Than Twice the Typical Span The first launching girders of this type were used on the following bridges: Rio Niteroi in Brazil; Deventer in Holland, Figure 3.50; and B-3 South Viaducts in the eastern suburbs of Paris, Figure 3.93. The Rio Niteroi Bridge (Section 3.8), linking the city of Rio de Janeiro with Niteroi, consists of 10 miles (16 km) of bridge deck constructed by four identical launching girders, Figures 3.55 and 3.56. Each 545 ft (166 m) long girder could be completely dismantled. The constant triangular sec-

FIGURE 11.74. Prestressed connections.

516


tion, weighing 440 tons (400 mt), could cope with spans of up to 260 ft (80 m). The connections were identical in principle to those used for the Blois girder. Each installation was equipped with three supports of nontunnel type, one fixed and the other two retractable. The erection sequence was as follows, Figure 1.51: Phase 1: Segment placing The girder rests on three supports, each one over a pier. Two segments are erected simultaneously, one on either side of the double cantilever under construction. The pier segment of the next cantilever is also placed with the launching girder in this position. Phase 2: Moving the launching gder forward The girder rolls on two temporary translation supports, one placed above the pier of the finished cantilever and the other above the pier of the cantilever to be constructed. These temporary supports are attached to the trolleys; the launching girder is lifted, thus freeing the permanent supports; and the trolleys are engaged, enabling the translation of the launching girder to a position to erect the next cantilever. The temporary translation supports are equipped with a mechanism allowing transverse movements, as the structure includes a certain amount of horizontal curvature. The Rio Niteroi girder was equipped with three sets of active stays: lateral stays, central stavs, and launching stays. The lateral stays, positioned on the underside of the two spans and constantly under tension, ensure the resistance of the girder while the load (segment) passes near midspan. The central stays strengthen the girder in the vicinity of the central support. The launching stays, under tension while maneuvering the girder, transfer the front and rear reactions to the central support.

Owing to the length of the bridge and the presence of a large stretch of water beneath the structure, the segments were brought to the launching girder on barges. The cantilever stability of the bridge was assured by the launching girder itself, and ties and props were positioned as construction proceeded. The launching girder used for the Deventer Bridge in Holland, Figures 3.49 and 3.50, were also capable of being entirelv dismantled and of triangular section. Its total length was 5 12 ft (156 m) f o r a w e i g h t o f 1 9 8 t o n s ( 1 8 0 m t ) . T h e maximum span length was 243 ft (74 m). Assembly of the launching-girder elements was consummated by prestress bars normal to the joints. It was supported by the fixed supports, of. which the rear and the central allowed the passage of a segment, and two sets of cable stavs: central stays and launching stays. The translation operations were identical to those of the Rio Niteroi Bridge, even though only one segment could be lowered into place at a time. What was peculiar about this launching girder was its abilitv to raise itself to its working level bv its own means, and this from the ground level where it was assembled. This was made possible bv the central suspension mast, which acted as a lifting jack.

In the case of the B-3 South Viaducts, Figure 3.92, the constantly varying structure supported b\ 200 piers, crossing five railway tracks, the Ourcq Canal, and several urban roadwavs, was mastered by a highly mechanized launching girder. The simultaneous placing of two segments of the same cantilever, each weighing between 33 and 55 tons (30 and 50 mt) either side of the pier, is controlled by a radio-controlled servo mechanism that synchronizes the loading at each end of the girder. Again the length of the launching girder was slightly greater than twice the typical span length,

TYPICAL CROSSw8ECTION

FIGURE 11.75. B-3 South Viaduct launching girder. general la\o11t.

517

References

that is, between four and six segments per day. The average construction speed, including launching-girder maneuvers, was therefore 200 ft (60 m) per week. The B-3 launching girder was recently reused for the Marne-la-Vallee Viaduct, which carries high-speed suburban rail for the Paris transport authority. References FIGURE 11.76. B-3 Sourh Viaduct, segment transport tractor.

which varied between 100 and 164 ft (30 and 50 m), Figure 11.75. The girder support reactions were thus applied in the region of the piers, and the cantilever stability was ensured by the launching girder itself. This stabilizing device can be seen to the left of the central support in Figure 11.75. The segments were supplied by a special eightwheeled tractor moving along the top slab, Figure 11.76. A special device used to unload and store the segments brought by the tractor freed the latter and removed the supply of segments from the erection critical path. The cycle of segment placement and girder advancement is represented in Figure 3.93. The next pier segment was placed during the same phase as the typical segments. About two spans were constructed each week-

1. Anon., Manual for Quality Control for Plants and Production of Precast Prestressed Concrete Products, MNL116-70, Prestressed Concrete Institute, Chicago, 1970. 2 . Anon., ACI Manual of Concrete Practice, Part I, American Concrete Institute, Detroit, 1973. 3. “Proposed Recommendations for Segmental Construction in Prestressed Concrete,” FIP Commission-prefabrication, 3d Draft, September 1977. 4. “Recommended Practice for Segmental Construction in Prestressed Concrete,” Report by Committee on Segmental Construction, Journal of the Prestressed Concrete Institute, Vol. 20, No. 2, March-April 1975. 5. Anon., PCI Post-Tensioning Manual, Prestressed Concrete Institute, Chicago, 1972. 6. Anon., PTI Post-Tensioning Manual, Post-Tensioning Institute, Phoenix, Arizona, 1976. 7. T. J. Bezouska, Field Inspection of Grouted PostTensioning Tendow, Post-Tensioning Institute, Phoenix, Arizona, March 1977.

12 Economics and Contractual Aspects Of Segmental Construction

12.1 BIDDING PROCEDURES

12.2

12.1.1 Single Design 12.1.2 Design and Build 12.1.3 Value Engineering 12.1.4 Alternate Designs 12.1.5 Summary Remarks on Bidding Procedures EXAMPLES OF SOME INTERESTINti BIDDINGS AND cDsrs 12.2.1 Pine Valley Creek Bridge, California 12.2.2 Vail Pass Bridges, Colorado 12.2.3 Long Key Bridge, Florida 12.2.4 Seven Mile Bridge, Florida

12.1 Bidding Procedures A bridge design should on principle be economical and as a practical matter must fall within budgetary restrictions of a particular project. The economic “moment of truth” for a given bridge design occurs when bids are received and evaluated. In a basically stable economy where material and labor costs are predictable within relatively small fluctuations, the selection of structure type and materials is relatively straightforward. This situation prevails when the time required for the design is relatively short and thus is not affected by economic cycles, or, if the design time is relatively long, the economic cycles are mild. In an inflationary economy there is no economic stability, and designers are hard put to make rational choices, as they have no control over economic parameters that can influence their design decisions. In short, the problem is whether economic assumptions made during the course of design are valid at the time of bidding. 518

12.2.5 Zilwaukee Bridge, Michigan 12.2.6 Cline Avenue Bridge, Indiana 12.2.7 Napa River Bridge, California 12.2.8 Red River Bridge, Arkansirs 12.2.9 North Main Street Viaduct, Ohio 12.2.10 Summary of California’s Experience 12.3

INCREASE IN EFFICIENCY IN CONCRETX BRIDGES 12.3.1 12.3.2

Redesign of Chacas Viaducts, Venezuela Comparison between Tancarville and Brotonne Bridges, France

REFERENCES

Obviously, the design and the bidding (tendering) of a project are closely related. Contractual bidding procedures vary from country to country, and current economic pressures are leading to changes in these procedures. The various bidding methods used in various countries can be broadly categorized (with some possible variations) as follows: (1) single design, (2) design and build, (3) value engineering, and (4) alternate designs. 12.1.1 SISGLE DESIGl\ Heretofore, single design was the major method used in North America and Great Britain. In this method, in general, design drawings prepared for bid are very detailed, to the extent that even the length and other dimensions of every reinforcing bar may be given. The bidding period is followed by a tight construction schedule. The contractor bids and executes the project in strict accordance with the bidding documents. No variation from the documents is allowed unless an error in design is

519

Bidding Procedures discovered, or a specific detail proves impractical to consummate, or geological perturbations are discovered that differ from what was assumed in design and delineated in the contract documents. These changes are authorized by a change order, and if there is an increase in cost the contractor is paid an “extra.” This system worked well for many vears when the economy was fairly stable and predictable and when economic changes were gradual over an extended period. Its disadvantage is its lack of flexibility to accommodate an inflationary economy, sudden price changes in materials, a rapidly advancing technology, and the current emergence of specialtv contractors with unique equipment or skills, proprietarv designs, and patented construction methods. Its biggest advantages are ease in administering the contract and absolute control over the final design.

In some European countries, by contrast, bid documents are prepared with the intention that the contractor will prepare and submit his own detailed design for the prqject. Thus, bid plans will be more general and, for a bridge, may show only span lengths, profile, and typical sections. The contractor may then refine the original design or submit an alternate design of his own choice, the responsibilitv for producing the final design and details being his rather than the engineer’s, This procedure allows the contractor to use any special equipment or technique he may have at his disposal. For example, a cast-in-place concrete box mav be substituted for a steel superstructure where the contractor has special know-how in concrete construction, or the change may be less drastic and involve only a reduction in the number of webs in a box girder. Verification of the adequacy of the contractor’s final design is generally carried out by a “proof engineer” who is retained by the owner or is on the owner’s engineering staff. In order to minimize disagreements between the contractor and the proof engineer, European codes have been made very specific. As a result, European contractors usually maintain large in-house engineering staffs, although they may also use outside consultants. The outcome apparently is a savings in construction cost, achieved by the investment of more design time and effort than in the single-design method.

The advantage of the design-and-build method is that in an atmosphere of engineering competition, innovative designs and construction practices advance very rapidly. The state of the art of designing and constructing bridges advances in response to the need for greater productivity. The disadvantage is the lack of control over the selection of the type of structure and its design. There is some concern, too, that quality of construction may suffer as a consequence of overemphasis on productivity and initial cost. However, the contractor is usually required to produce a bond and guarantee his work over some period of time, and any defects that surface during this period have to be repaired at his expense. Whether such a system could be adopted in the United States is debatable. 1 2 . 1 3 VALUE

E.\‘GI.~EERI,~G

Value engineering is defined by the Society of American Value Engineering as “the systematic application of recognized techniques which identify the function of a product or service, establish a value for that function, and provide the necessary function reliability at the lowest overall cost. In all instances the required function should be achieved at the lowest possible life-cycle cost consistent with requirements for performance, maintainability, safety, and esthetics.“’ In 1962 the concept of value engineering became mandatory in all U.S. Department of Defense armed services procurement regulations (ASPR). Before this time value engineering had been applied to materials, equipment, and systems. The advent of ASPR provisions introduced value engineering concepts to two of the largest construction agencies in the United States-the U.S. Army Corps of Engineers and the U.S. Navy Bureau of Yards and Docks. Soon thereafter the U.S. Bureau of Reclamation and the General Services Administration (GSA) adopted and inserted value engineering clauses in their construction contracts, and the U.S. Department of Transportation established a value engineering incentive clause to be used by its agencies. Several value engineering clauses (or’ incentive clauses) are in use today by many agencies. In general, they all have the following features’: 1.

A paragraph that defines the requirements of a proposal: (a) it must require a change to the contract and (b) it must reduce the cost of the contract without impairing essential functions.

Economics and Contractual Aspects of Segmental Construction

520

2. A “documentation” paragraph that itemizes the information the contractor should furnish with each proposal. It should be comprehensive enough to ensure quick and accurate evaluations, detailed enough to reflect the contractor’s confidence in its practicability, and refined to the point where implementation will not cause undue delay in construction operations. Careful development of this paragraph and meticulous adherence to its requirements will preclude scatter-shot proposals by the contractor and burdensome review by the agency. 3.

A paragraph on “submission.” This paragraph details the procedure for submittal.

4. A paragraph on “acceptance,” which outlines the right of the agency to accept or reject all proposals, the notification a contractor may expect to receive, and appropriate reference to proprietary rights of accepted proposals. 5.

A paragraph on “sharing,” which contains the formula for determining the contract price adjustment if the proposal is accepted and sets forth the percentage of savings a contractor may expect to receive.

As generally practiced by highway agencies in the United States, a value engineering proposal must indicate a “substantial” cost savings. This is to preclude minor changes such that the cost of processing offsets the savings to be gained. Some other reasons for which a value engineering proposal may be denied are as follows: Technical

noncompliance.

Delay in construction such that the cost savings would be substantially nullified. Proposed change would require resubmission of the project for any number of various permits, such as environmental impact statement, wetlands permit, and navigation requirements. Resubmission would in all probability delay construction and nullify any cost savings. Savings resulting from a value engineering proposal are generally shared equally by the agency and the contractor, after an allowance for the contractor’s development cost, the agency’s cost in processing the proposal, or both. As practiced in the United States, all contractors must bid on the design contained in the bid documents, and only the low bidder on the base bid is allowed to submit

a value engineering proposal. This is, of course, value engineering’s biggest disadvantage. An) number of contractors may have more costeffective proposals that they are not allowed to submit because they were not low bidder on the base bid. Its advantage is that to some degree it allows contractor innovation to be introduced.

Alternate designs, as it is developing in the United States, basicallv is an attempt to produce a hvbrid system consisiing of the best elements oi the single-design and the design-and-build methods. It attempts to accomplish the following: Retain for the authorizing agency control over the “type selection” of the structure and its design Provide increased competition between materials (structural steel versus concrete or prestressing strand versus bars) or construction procedures (cast-in-place versus precast segmental or balanced cantilever versus incremental launching, and so 04 Provide contractor flexibility (construction dures, methods, and/or expertise)

proce-

This method has developed, with encouragement from the Federal Highway t\dIninistratiorl, as an anti-inflationary measure to combat dramatic increases in highway construction costs. A technical Advisory2 published by the Federal Highway Administration states: Because qf.fluctuating economic conditions, it isfelt that on multiple repetitive spans, long spans or major bridges, or where there is an extended period qf‘design from conception of the project to a release for bds, there can be no assurance of price stability fbr n particular material or construction methodoloCg. With alternate de.siLgns, no matter how the economy changes, more designs are ctzlailable at the time of biddt’ng that are likely to be suited to the prevailing economic conditions. General recommendations regarding alternate designs from the same document’ are as follows: 1.

To receive the most economical construction between basic structural materials, consistent with geographic, environmental, ecological or other site restrictions, there should be maximum opportunity for competition between structural steel and concrete.

Bidding Procedures 2. Within environmental, aesthetic, site, and other constraints, the plans and bid documents should show or otherwise indicate what alternative types of structures will be allowed or considered. The contractor should be allowed the option to bid any designated alternative design that is consistent with the contractor’s expertise, available equipment, and so on. 3.

Bid documents and the contract plans should clearly indicate the design criteria and what tvpe of alternative designs and/or contractor options will be acceptable. Determination of practical and economical alternatives and/or contractor options should be developed in the preliminary design.

4. Bid documents should be considered as “open” documents in regard to construction method, erection systems, and prestressing svstems. 5.

Consistent design criteria should be used for alternatives; for example, if load factor design is used, it should be used for all alternatives.

6. Span lengths should be identified on the contract plans. However, other than where pier locations are constrained by physical and geological conditions at the site, consideration should be given to allowing a tolerance in pier location to avoid placing a particular alternative at an economic disadvantage. For example, in a typical three-span structure, the side span should be approximately 80 percent of the center span for structural steel, 70 percent for conventional cast-in-place concrete on falsework, and 65 to 70 percent in segmental balanced cantilever construction. 7.

To avoid an economic disadvantage to a particular superstructure alternative, alternative substructure designs may be required. Limitations on the substructure, such as allowable axial load and moment, should be clearly identified on the contract plans. 8. Where specific design requirements are not covered by the American Association of State Highway and Transportation Officials (AASHTO) Bridge Specifications, the contractor should be allowed to use other recognized codes and standards where applicable. However, the alternative design should document where these provisions are to be used, why the AASHTO requirements do not

9.

521

apply, and which articles of the substituted code or standard are to be used. Such provisions should be subject to approval by the engineer and appropriate agencies. Prebid conferences are to be encouraged as a means of communication between the engineer, highway agencies, and contractors.

10.

In order to allow a contractor adequate time to investigate the various alternatives and prepare plans, it is recommended that the advertising time be commensurate with the size and complexity of the project with a minimum of 60 days.

11.

In order to allow adequate review and checking of the low bidder’s proposal, award of contract should be extended commensurate with the size of project.

Specific recommendations* regarding concrete alternates are as follows: 1.

prestressed

To increase the competition in post-tensioned concrete construction, it is recommended that plans and other bid documents allow conventional cast-in-place on falsework, precast prestressed span units, and segmental construction or combinations thereof.

2. Segmental construction should allow the following at the contractor’s option: a. Precast or cast-in-place segmental construction. b. Any of the post-tensioning systems-that is, strand, wire, or bars or combinations thereof. following construction c. Any of the methods: balanced cantilever, span-byspan, progressive placing, incremental launching, or combinations thereof. d.

Exterior dimensions of the cross section should be fixed. At the contractor’s option, the thickness of webs and flanges may be varied to accommodate proposed construction and erection methods and post-tensioning systems, providing that any changes in the dead weight, shear, and so on are accommodated in the design. 3 . T h e c o n t r a c t plans should indicate the maximum and minimun final prestressing force (P,) and moment (Pr x e) required, after all losses, for the final condition of the structure -that is, dead, live, impact, and all superimposed loads. Any increase in prestressing force

522

Economics and Contractual Aspects of Segmental Construction requirements as a result of the method of construction, erection, or type of tendon system should be evaluated at the shop drawing stage.

4.

Changes in eccentricity of prestress should be accompanied with appropriate changes in prestress force to produce the same minimum compressive stress due to prestress.

5.

The minimum prestress force should be such that under any loading condition, both during and after construction, stresses will be within allowable limits. Consideration should be given to secondary moments due to prestress, redistributed moments due to creep, and stresses resulting from thermal gradient (between the top and bottom of the girder and between the inside and outside of webs). 6. Contractor revisions to contract plans, with supporting calculations, should be submitted to the engineer for approval.

12.1.5 SU,MMARY REtMARKS PROCEDURES

OlV BIDDIiVG

All of the bidding procedures described above have one thing in common: they all attempt to produce the lowest initial cost by competition in construction and/or design. All of the last three approaches (design-and-build, value engineering, and alternate designs) require decisions based on comparisons of basic structural materials, structure types, construction methods, and so on. This implies that the basic premise in the selection process is equivalency-comparable service, performance, and life-cycle cost of the facility. Life-cycle costs refer not only to initial cost, but also to maintenance and any rehabilitation costs during the life of the structure. True cost of the project must be considered. What may be initially least expensive may in the long run, when future costs are accounted for, be actually most expensive. Some newer structure types and designs are at the fringe of the state of the art and have only been used in the United States within the last decade or less. Thus, an adequate background of experience is unavailable to evaluate life-cycle costs. The estimation of life-cycle costs may be difficult in many cases, such as for new and progressive bridge designs. Functionally, alternative structures are designed to the same criteria. Only years of operational experience can provide the data base for reasonably estimating life-cycle costs and thereby true equivalency in design insofar as cost is in-

volved. However, the problem of adequacy of data does not diminish the importance of the question and the need to attempt to answer it. Another anti-inflationary measure used in recent years is that of stage construction. This concept may take one of two forms. Major structures, because of their size, lend themselves to stage construction-that is, separate substructure and superstructure contracts. Usually several years will elapse between bidding and awarding of the substructure contract and the superstructure contract. The economic superstructure span range for different alternative types and materials is a variable. In this form of stage construction the substructure is let first; thus the spans for the superstructure design become fixed. This may or may not impose an economic disadvantage to specific superstructure alternates. The substructure must be designed for the largest self-weight superstructure alternative, which may or may not be the successful superstructure alternative. It appears that this form of stage construction may be to some extent selfcanceling or counterproductive to cost savings. With a total alternative design package, the substructure (foundation, piers, span arrangement) can also have alternatives commensurate with the superstructure alternatives. The other form of stage construction concerns a large project, containing many bridges, that is subdivided for bidding purposes into a number of smaller projects. Its primary purpose is to encourage small contractors bv providing prqjects o f manageable size, thus Increasing competition. However, certain construction techniques, by virtue of the investment in sophisticated casting or erection equipment, require a certain volume of work to amortize the equipment and be competitive. Depending upon the size of the subdivided contract, this form of stage construction in some instances may also become counterproductive. The value engineering concept can be divided into two major areas of application: during design and during construction. Value engineering procedures in the design stage may result in very specific recommendations based on a certain set of assumptions at a particular point in time for the design. If conditions change during the interval between the design decision and the actual construction, which can be several years, conditions on which the assumptions were based may have changed. Such changes could make the original value engineering decision incorrect. The alternative design concept, on the other hand, does not make all such specific design decisions at an early

523

Examples of Some Interesting Biddings and Costs stage but retains some options in order to allow a later response to changed conditions. Therefore, there is an apparent incompatibility between the application of value engineering principles in the design stage and the concept of alternative designs for bidding purposes. However, the concept of value engineering is a powerful tool and can be made compatible with the concept of alternative designs if its principles are used to determine whether a given project should require alternative designs and, if so, what structure types should be considered as equivalent alternates.

12.2

12.2.1

Examples of Some Interesting Biddings and Costs

PI.L’E VALLEY CREEK BRIDGE, C‘4LIFORNlA

The Pine Valley Creek Bridge (Section 2.7) was the first segmental bridge in the United States to incorporate the concept of value engineering (costreduction incentive proposal) in the bidding documents. The original design assumed cast-inplace balanced cantilever construction with the prestressing force (P,) and moment (Pr x e) based on strand capabilities. Project plans developed by CALTRANS required construction of the cantilevered sections at piers 3 and 4 before those at piers 2 and 5 (refer to Figure 2.40). Because of rigid specification requirements for the protection of the valley slopes, TABLE

12.1.

the contractor was faced with a difficult and costly erection procedure. At least one cableway spanning the entire valley would have been required to transport men and material to appropriate locations during the construction procedure. Complete site installations on both sides of the valley would have been required, which were accessible only with a great deal of difficulty. The superstructure design alternative submitted under this proposal employed the Dywidag system, which included the following: (1) use of the threaded-bar system especially suited to segmental construction, (2) increased prestress-force eccentricities, since most longitudinal prestressing bars could be placed and anchored in the slab, (3) diagonal prestressing in the webs to cater to the shear stresses, and (4) a modification in construction sequence so as to work from piers 5, 4, 3, and 2, Figure 2.44. Changes proposed under the value engineering clause are summarizied in Table 12.1. Total savings as a consequence amounted to $382,000.3 12.22 VAIL PASS BRIDGES, COLORADO The Vail Pass structures are part of Interstate I-70 near Vail, Colorado, in an environmentally sensitive area. Environmental considerations played a dominant role in the selection of the bridge types and the design thereof. Another factor considered was the relatively short construction season at the high elevations of the sites.

Pine Valley Creek Bridge, Value Engineering Proposal

Original Construction sequence Structural svstem

Long cantilevers before short to minimize creep As Cantilever for D.L., continuous for L.L. + added D.L. No Redistribution

Concrete stresses: Construction Tension Concrete shear

CRIP Reverse to facilitate constr. from abutment

Savings estimated $ 88,000

Continuous for all loads Full redistribution 0.55fi 3*

Slab designs

AASHTO loads, method and distribution reinf.

Hinge

Two diaphragms

Principal stresses diagonal prestress AASHTO loads with “Homberg” graphs, no distribution reinf. No diaphragms

$228,000 22,000 22,000

22,000 $382,000


524

Of the 21 bridge structures in this project, 17 were designed and bid on the basis of alternative One alternative design considered designs. trapezoidal steel box girders composite with a concrete roadway flange. The other alternative design was for precast concrete segmental box girder design, with the Federal Highway Administration requiring that the contractor be given the option of cast-in-place segmental construction. At two locations, the site constraints were such that they were bid without alternatives as steel box girders. Two other locations required 80 ft (24 m) long simple-span underpass structures to provide for wildlife migration. These structures were built of cast-in-place concrete box girder construction. The remaining 17 structures were completely designed and detailed for the two alternatives, one in structural steel and the other in precast concrete segmental (with a contractor option of providing a cast-in-place segmental design). Spans varied in number from two to five and in length from 30 to 260 ft (9 to 79 m). Table 12.2 tabulates the five contracts that involved the 17 bridges bid on the basis of alternative designs and lists them in the order in which they were bid.4 Approximately a year elapsed between the letting of the first and last contracts. Although considerable differences in bid prices are shown in individual projects, for the total project there is less than $80,000 difference out of an approximate total cost of- $17 million, or less than 0.5% difTABLE

Proj. No. 1

2

3 4

5 Totals

12.2.

Bridge No.

No. Spans

Length (ft)

F-l l-AX F-l l-AW F-l 1-AV F-l l-AU F-12-AK F-12-AM F-12-AN F-12.A0 F- 12-AP F-l I-AP F-l l-A0 F-l l-AN F-l l-AM F-l l-AL F-l l-AK F-12-AT F-12-AS

4 5 4 4 3 2 3 3 3 3 2 4 4 4 3 4 4

727 880 690 668 220 240 350 368 600 310 222 740 744 514 450 726 726

ference in bid prices for the alternatives. Precast segmental was the low bid on project 1 and castin-place segmental was the low bid on project 4. Based upon length (width was constant), the segmental concept was successful in approximately 60% of the total project. The consultants, International Engineering Company, Inc., estimated that the additional cost to produce alternative designs was about 2.5% of construction cost. It is difficult to estimate what savings were achieved by bidding alternative designs rather than a single design; however, overall savings of 7 to 10% of the construction costs are not unreasonable.4 12.2.3 LO.VG KEY BRIDGE, FLORID.4 The Long Key Bridge in the Florida Keys was bid utilizing the concept of alternate designs. Four complete sets of contract plans were prepared for the alternative construction schemes indicated in Table 12.3. Plans for the AASHTO precast, pretensioned I girders were prepared by the Florida Department of Transportation. Plans for the three basic precast segmental schemes were prepared by the state’s consultant, Figg and Muller Engineers, Inc. In the preliminary design stage three methods of segmental construction were considered for this project: balanced cantile\.er, span-by-span, and 1,rogressk.e placing. ‘l‘he progressi\,e placing

Results of Alternative Bids, Vail Pass Bridges

Total Length (ft)

Low Steel Bid

Cost/F?. Steel

2965

$5,992,155

$48.12

f5,527,3 18

s44.39

1778

$3,777.549

$50.59

$4,111,170

$55.05

532

$994,347

$44.50

$1,053,364

$47.14

2448

$4,257,77 1

$4 1.4 1

$4,108,057

$39.96

1452 9175

$2,298,409 $17,320,23 1

$37.69 $44.95

$2.598.938 $17,398,847

$42.62 545.15

LOW

Concrete Bid

CostiFr’, Concrete

525

Examples of Some Interesting Biddings and Costs TABLE 12.3. Long Key Bridge, Alternatives Substructure Precast

Superstructure Precast girders, AASHTO Segmental: Span by span, V piers Span by span, vertical piers Cantilever, vertical piers First option, slab reinforcing Second

option,

barrier

A

B

C E G

D F H

Shafts

R/C epoxy coated Pretensioning Cast-in-place conventional Precast (never integral)

curbs

method was discarded because it was felt to be (at the time) too new for acceptance in U.S. practice. It was later introduced on the Linn Cove Bridge in North Carolina. The basic difference in the two span-by-span alternatives for the Long Key Bridge is in the pier configuration: V piers or vertical piers. Aside from the construction alternatives and pier types, the contractor was offered the option on all segmental alternatives of transversely reinforcing the top flange either with epoxy-coated conventional reinforcing steel or by transversely pretensioning with 4 in. (12.7 mm) diameter strand. Further, he had the option on all segmental alternatives of either precasting or casting in place the traffic barriers. The contractor also had the option of casting the segments rightside up or upside down. Casting the segments upside down was intended to facilitate transversely pretensioning the top flange. However, since no waterproof membrane or wearing surface was specified, the top flange surface of the deck was required to have a grooved or tined surface for skid resistance. If the segment were cast upside down, then, the form would be required to produce the desired texture. Specifications were left open such that strand or bar prestressing tendons could be bid. All conventional steel reinforcement was required to be epoxy coated in all alternatives. The eight basic alternatives for this project produced bids from eight contractors, as indicated in Table 12.4. Note that there were six bids for the span-by-span method, one for the balanced cantilever method, and one for the precast pretensioned AASHTO I girders. The low bid in precast segmental was $2.6 million less than the AASHTO I-girder bid. Low bid was for the span-by-span alternative with precast V

Drilled

Piles

TABLE 12.4. Long Key Bridge, Bid Tabulation Bid Rank

Alternative

Chosen

D F F F B F F H

1 2 3 4 5 6 7 8

Relative

Bid

1.0000 1.0225 1.0539 1.0963 1.1731 1.1844 1.2557 1.3063

piers and drilled shaft foundations. The contractor elected to precast the segments near the project site and cast the segments rightside up, using transverse prestressing in the top flange. He slipformed the cast-in-place barriers after segment erection. Further, he elected to move the scaffolding trusswork from span to span by using a barge-mounted crane as opposed to having the falsework trusses mounted on barges. Table 12.5 presents a cost analysis of the low bid as compared with the AASHTO pretensioned Igirder alternative.5 TABLE

12.5. Long Key Bridge, Cost Analysis of the Low Bid and the AASHTO I-Girder Bid Span-by-Span Segmental

Total cost” Superstructure cost Substructure cost Segments erected Total bid Total area

Precast AASHTO I Girder

$26.63/ft2 $21.43/ft2

$30.95/ftZ $23.59/f@

$ 5.20/f? $19.16/ft2 $15,307,375.91 4 6 8 , 3 0 1 ft2

$ 7.36/f? $17,956,538.75 4 7 0 , 2 7 7 ft*

“The mobilization bid items were proportioned to the structural items in all cases. The Florida Department of Transportation estimate was $14,550,000.


526

TABLE 12.6. Seven Mile Bridge, Alternatives

Substructure Superstructure Precast AASHTO I-girders Segmental: Span-by-span, vertical piers Cantilever, vertical piers First option, slab reinforcing

Precast Piles

Drilled Shafts

A

A

C E

D F R/C epoxy coated Pretensioning Cast-in-place conventional Precast (never integral) Cast-in-place conventional Precast

Second option, barrier curbs Third option, box piers

12.2.4 SEVE,\’

MILE BRIDGE, FLORIDA

12.2.5 ZILWz4CTKEE

BRIDGE. ,MICHIG,4.\

Seven Mile Bridge in the Florida Keys had the same basic alternatives as Long Key Bridge, except that the span-by-span with V piers was eliminated and the contractor had the further option for the vertical piers of casting in place conventionally or precasting, Table 12.6. Six bids were received, with all bidders selecting alternative D, Table 12.7. Low bid was $44,986,942.3 ! 1 here were no bids for the AASHTO I-gil,ders. The low bidder optioned t o reinforce the top slab with conventional reinforcement, epoxy coated; to cast the barrier curb in place; and to precast segmental box piers. The low bid included $5,128,600 for waterline, roadway approaches, and navigational requirements. Analysis of the bid items revealed a superstructure cost of $23.22/ft2 and a substructure cost of $5.68/ft2, resulting in a $28.90/ft2 total cost. The Florida Department of Transportation estimate was $52 million ($7 million higher than the low bid).

This structure was designed with alternatives of steel plate girders and precast segmental concrete box girder. Bids were first taken in November 1978, Table 12.8. The engineer’s estimate for the concrete alternative was $60,609,614.30 and for the steel alternative $71,3 16,854.90. On the basis that the low bid of $80,999,445.50 was 33% higher than the estimate, the bids were rejected. The design underwent revision, and the bid documents allowed the contractor to make design and prestressing system changes under a “cost reduction incentive,” and an escalation clause was introduced. The project was rebid in August 1979 (nine months later), Table 12.9. The engineer’s estimate was $7 1,645,661.50 for the concrete alternative and $7 1,965,5 16.70 for the steel alternative. The low bid of $76,787,252.65 was 7% over the estimate-5% below the previous low bid. By rebidding the project (after nine months of inflation) a savings of $4.2 million was achieved.

TABLE 12.7. Seven Mile Bridge, Bid Tabulation

TABLE 12.8. Zilwaukee Bridge, Ranking of First Bids

Bid Rank 1 2 3 4 5 6

Alternative D D D D D D

Chosen

Relative Bid 1.0000 1.0214 1.0768 1.1404 1.2297 1.2556

Bid Rank

Relative Bid

Alternative

1.oooo 1.0115 1.0562 1.0816 1.1071 1.1375

Concrete Concrete Steel Concrete Steel Steel

527

Examples of Some Znteresting Biddings and Costs TABLE 12.9. Zilwaukee Bridge, Ranking of Second Bids Bid

Rank

Relative

1 2 3 4 5

Bid

1 .oooo 1.0798 1.0829 1.1231 1.1501

12.2.6

Alternative Concrete Concrete Concrete Steel Steel

CLINE AVENUE BRIDGE, INDIANA

The Cline Avenue bid documents were very liberal toward redesign, with the bidder only having to inform the state of the intention to redesign at bid opening. As designed, the plans and specifications provided the option of a steel plate girder or precast segmental box girder structure. The structure was redesigned as cast-in-place on falsework, prestressed concrete box girder, except for the main channel spans which are cast-in-place segmental. The steel option was a composite load factor design. The low concrete bid was for $53,545,770.55 with the engineer’s estimate being $53,560,259.78. Relative bids are listed in Table 12.10. 12.2.7 NAPA

RIVER BRIDGE, CALIFORNIA

The Napa River Bridge (Section 2.11) is another example of the use of alternative designs. For this project, because the lower structure height made falsework feasible, bid documents were prepared for three alternative schemes: A: Conventional continuous cast-in-place box girder bridge B: Trapezoidal continuous structural steel box girder bridge C: Cantilever prestressed segmental concrete bridge with either precast or cast-in-place segments, and erection either by the balanced cantilever method or on falsework Because of poor foundation material and a readily available aggregate supply, all alternatives used lightweight concrete in their superstructures. TABLE 12.10. Cline Avenue Bridge, Ranking of Bids Bid 1 2 3

Rank

Relative

Bid

1 .oooo 1.0252 1.0596

Alternative Concrete Steel Steel

Alternative C used a transverse prestressed deck in order to reduce the number of girders. Strong competition was expected from the steel industry, as this site is readily accessible by water from the yards of two major fabricators. However, the low bidder, G. F. Atkinson Company, selected alternative C and cast the bridge generally in half-span segments on falsework to the ground as a series of balanced T’s. About 60 ft (18 m) of the 250 ft (76 m) span over the navigation channel was constructed in three segments on falsework suspended from the cantilevered boxes on each side.‘j There were six other bidders, of which only one bid the steel alternative and none bid alternative A. Relative bids are listed in Table 12.11. The first six bids were for alternative C, and the last and highest was for alternative B. 122.8 RED RIVER BRIDGE, ARKANSAS This is a seven-span structure with five interior spans of 210 ft (64 m), end spans of 135 ft (41 m), and a roadway width of 32 ft (9.75 m). Estimated cost was $3.3 million. Eight bids were received, six in structural steel and two for concrete segmental. Bids ranged from $3.22 to $4.89 million. The concrete segmental was completely open as to the method of construction (both concrete bids were based on the incremental launching method). Relative bids are listed in Table 12.12. TABLE Bid

12.11.

Napa River Bridge, Ranking of Bids Relative

Rank

1.0000 1.0928 1.1218 1.1837 1.2765 1.4305 1.5210

1 2 3 4 5 6 7

TABLE 12.12. Bid 1 2 3 4 5 6 7 8

Bid

Rank

Red River Bridge, Ranking of Bids Relative 1 .oooo 1.1437 1.2685 1.2800 1.3099 1.3229 1.4267 1.5175

Bid

Alternative Steel Steel Steel Steel Concrete Concrete Steel Steel


528

Structural steel prices varied from a low of $0.65/1b to a high of $0.93/1b with an average of $0.78/1b. Structural steel prices in Arkansas for this type of construction had previously been in the range of $0.80 to $0.85/1b. The low bid price of $0.65/lb represents a reduction of approximately 19 to 23%. All steel prices were for domestic steel. Note that the bid prices included the demolition of the existing bridge. If this item were deleted, the bidding would be rearranged as indicated in Table 12.13. The lump-sum price for the concrete superstructure was, for the low concrete bidder, $3 1.37/ ft2, which compares favorably with the Keys bridges in Florida. However, this price was not competitive. Undoubtedly there are numerous reasons why. One may be that there was no precasting plant within sufficient distance of the site, and thus the cost of shipping the segments may have been prohibitive. The project was not large enough to attract contractors with the expertise to set up a precasting operation at the site. The two concrete segmental bids received were based on incremental launching, and evidently the project was not large enough to adequately amortize the cost of the casting bed and launching equipment on this project to make the method competitive. 12.2.9 NORTH Md4I,\’

Bid Rank” 1 4 2 8 3 5 6 7

12.2.10

12.3

Bid

1 .oooo 1.1147 1.1620 1.2841 1.3379 1.3919 1.4054 1.5276

Alternative Steel Steel Steel Steel Steel Concrete Concrete Steel

“Ranking corresponds with that presented in Table 12.12.

OF CALIFORSIA’S

EXPERIE.VCE

Increase in Efjciency in Concrete Bridges

As stated in previous chapters, prestressed concrete segmental bridges have extended the practical and competitive economic span range of concrete bridges. An interesting comparative exercise is to look back at bridges built in the past and evaluate them in the light of presentday developments.

TABLE 12.14. North Main Street Viaduct, Ranking of Bids

Red River Bridge, Reranking of Bids Relative

SU,ZI,M,4RY

California’s experience with a cost reduction incentive proposal (CRIP) (value engineering) and alternative designs for projects involving segmental construction is summarized in Table 12.1 5.s

STREET VIADUCT, OHIO

The low bid on this project was $25,715,733.00, as compared with the engineer’s estimate of $29,200,000. Probably a major reason why the low bid was 12% under the engineer’s estimate was the competition offered by the two plan alternative designs in concrete and structural steel, resulting in a minimum savings of at least $3,500,000. The competitive situation was further enhanced by allowing bidders to propose additional optional designs. Although no additional steel optional designs were

TABLE 12.13.

submitted, three optional concrete redesigns were bid. As designed, the plans and specifications provided the option of a steel plate girder or precast segmental box girder structure. Bid documents were quite liberal for redesign but required quite a bit of detail with the bid documents. The winning bid was steel girders, as designed, priced at about $87/f? without one abutment, which was to be constructed under another contract. The steel girders were a noncomposite, working-stress design. The approximate]\ 15 million pounds of A588 structural steel was bid at $0.75/1b. It should be noted that additional savings in steel could have been accomplished with a composite design. Table 12.14 is a relative summary of the eight bids. Note that the low concrete bid was only 3.7% above the low bid, which indicates the competitiveness.

Bid Rank 1 2 3 4 5 6 7 8

Relative Bid 1.0000 1.0370 1.0401 1.0579 1.0884 1.1128 1.1508 1.4099

Alternative Steel alternative as Redesign concrete Steel alternative as Concrete alternative Steel alternative as Steel alternative as Redesign concrete Redesign concrete

per plan alternative per plan as per plan per plan per plan alternative alternative

TABLE Date Bid

Bridge

12.15.

Summary of California’s Segmental Prestressed Bridge Experience

Length/Max. span (ft)

Cost-Bridge Work Only

Design Alternatives Provided

2/72 P i n e \‘alle> Creek

17001450

$14.2 M = $69/f-?

A.

Y/72

225oi550

$12.0 M = $127/f? (inch. prov. for future widening)

A.

.Stani.slaus Kiter at New Melones

B.

.3/73 Eel River Br. and Overhead

1 1174

Napa RiteI- at Napa

13871310

$5.0 M = $37/t?

A.

22301250

$8.9 M = $54/f?

A.

B. C.

Y/i5

12176

Colomdo Ri\er at \‘uma

27501220

7.4 M = $34/ft2

A.

Guadalupe River at San Jose

1009/155

$2.6 M = $28/f?

A.

0.

.5/76 San Joaquin River at Anti-

94041460

$33.4 M = $79/f?

A.

och

B.

C.

Cantilever prestressed concrete segmental box girder with either: 1. Prestressed rock anchor footings” 2. Mined rock shaft foundation Cantilever prestressed concrete segmental box girder Structural steel box girder” Both with either: 1. Prestressed rock anchor footings 2. Mined shaft foundations” Conventional twin two-cell cast-in-place prestressed concrete box girders Conventional six-cell cast-in-place prestressed box girder Structural steel box girder Cantilever prestressed concrete segmental two-cell box girder” Segmental prestressed concrete single cell box with criteria provided to convert to cantilever Conventional sevencell prestressed concrete box” A four-cell prestressed box designed for segmental construction Prestressed concrete with three main spans designed for cantilever segmental construction and 20 300 ft approach spans designed for segmental construction with provisions to modify to cantilever Structural steel welded plate girders (unpainted A-588) with 29 200 ft approach spans” Same as A with 200 ft approach spans

Remarks CRIP by contractor revised superstructure design and construction sequence. Savings to state, $191,000.

Seven contractors bid steel and two bid concrete 1.6% separated low steel and low concrete.

CRIP modified design to segmental single-cell boxes. Max. falsework height 92 ft +. Savings to state, $112,824. Six of seven bidders chose C. Falsework heights varied from 64 to 132 ft.

Cantilever construction considered by contractor but not used because contractor owned adequate supply of falsework. Max. f.alsework height 70 ft i_. Maximum falsework height 40 ft over a seasonally dry river. Contractor inexperienced in segmental construction.

AH five bidders chose steel. Lower-than-anticipated (foreign) steel prices prevailed.

529


530

TABLE 12.15. (Continued) Date Bid l/78

Bridge San Francisco Bay at Dumbarton (appreach spans)

Length/Max. Span (ft)

Cost-Bridge Work Only

46501150

$24.3 M = $59/ft2

Design Alternatives Provided A. Precast delta girders” B. Twin single-cell prestressed box girders designed for segmental construction with criteria for redesign for cantilevering or launching C. Structural steel box girder

Remarks Much bidder interest in all choices during prebid stage. Final results: seven chose A, one chose C. Some uncertainties about criteria provided in B.

“Selected by low bidder.

12.3.1

REDESIGN OF CARACAS VIADUCTS, VENEZUELA

The Caracas Viaducts in Venezuela (Chapter 8) were completed in 1952, approximately thirty years ago. If these viaducts were built today, the chosen structure would probably be very different from that chosen at the time after exhaustive feasibility studies. In 1973 these structures were reevaluated in terms of the more conventional balanced cantilever method of girder construction. Figures 8.30 and 8.31 compare the actual project constructed in 1952 with possible alternative designs in 1973 and 1975. The three-archrib and eight-beam superstructure would be replaced by a variable-depth twin box section (cantilever construction using precast segments) supported on slip-formed piers. Today, with the same span arrangement considered in 1973, possible alternatives might be a single two-cell box similar to that used in the Kipapa Stream Bridge, or a ribbed single-cell box as in the Vejle Fjord Bridge, Figures 4.24 and 4.22, respectively. This approach would require only single shaft piers. \ 12.3.2

Tancarville Bridge designed in 1956 incorporating a 2000 ft (610 m) span steel suspension bridge Brotonne Bridge designed in 1973 incorporating a 1050 ft (320 m) span concrete stayed bridge These two structures are ohly 20 miles (32 km) apart and are located in very similar surroundings topographically and geotechnically (see Figures 12.1 and 12.2). On the left bank, a flat expanse of meadows and fields requires a long approach viaduct to reach the desired altitude of the main crossing above the navigation channel, while a deep formation of soft soil overlying the load-bearing strata requires deep pile foundations. On the right bank, the limestone cliff extends close to the river bank and calls for only a short transition between the main river span and the approach highway. The comparison presented here pertains only to the left-bank approach viaduct of each structure, although interesting comments could be made also on the relative characteristics of their other parts. The Tancarville approach viaduct has eight 164 ft (50 m) spans, having five 140 ton (127 mt), 10 ft

COMPARISON BETWEEN TANCARVILLE AND BROTONNE BRIDGES, FRANCE

Progress is made slowly through accumulated experience, and it is worthwhile to look back periodically and try to measure such progress. With this in mind, and as a conclusion to this chapter, a comparison is offered between two similar concrete structures separated in time by seventeen years. As mentioned in Section 9.8, the Seine River between the maritime inland harbor of Rouen and the English Channel is now crossed twice by two outstanding structures:

FIGURE

12.1. ‘I‘ancarville Bridge, set-ial v i e w from the southwest showing left bank approach viaduct.

f FIGURE 12.2. Brotonne Bridge, aerial view from the southwest showing left bank approach viaduct. (3 m) deep precast girders in each span, Figures

12.3 and 12.4. Piers are founded on precast concrete piles and were cast in place with a box section. When the design was prepared, it represented the most advanced technology in terms of use of materials. The elastic stability of these very long, slender precast girders was even the occasion of interesting innovative studies. Construction methods were also far from conventional. All girders were prefabricated in a yard located at the original ground level. Moving and lifting operations for one girder (see Figure 12.5) included: Placing girder on dollies, moving in two perpendicular directions to bring it at the foot of the supporting piers

Hoisting girder along the piers with special steel rigs, Figure 12.6 Placing girder on top of the pier with the rotating arm of the special rig, Figure 12.7 Transverse displacement of girder to its final position Suspension of the girder at both ends was achieved by the means of special cantilevers to provide the highest safety against lateral buckling during lifting operations of such slender girders, Figure 12.8. The project was carried out smoothly and completed successfully a long time ahead of the other contracts for the entire crossing. Fifteen years later, the same problem of safely and economically building an aesthetically pleasing

FIGURE 12.3. Tancarville Bridge, elevation of approach spans.


532

I L

11'-4"

I:

FIGURE 12.4.

11'-4"

_ _ __ _ f‘ancarville Bridge, lif ring one prec
Cost

Comparison

1. Quantities (per ft”) (super- and substructure) Concrete (yd3) Reinforcing steel (lb) Prestressing steel (lb) 2. Labor (hr/ftz) 3. Cost ($/ftZ) Labor Materials Equipment, plant, and job overhead Subcontracts Design, overhead and fee 4. Total

1 1'4"

1 l'-4"

Tancarville Bridge, typical cross section of approach spans.

FIGURE 12.5.

TABLE 12.16.

*'-o" I

Between

approach viaduct for the Brotonne crossing was solved with very different methods, both of design and construction. The light single box section selected for the stayed structure was also used for the approach spans. Precast piles were replaced in the foundations by cast-in-place slurry load-bearing walls. Box piers were slip-formed instead of incrementally cast in successive lifts. The superstructure was cast in place in balanced cantilever with travelers, Figure 12.9. Today precast segments would probably be preferred, although the characteristics of the deck would remain substantially unchanged. A comparison between quantities of materials per square foot of deck appears in Table 12.16. The savings in concrete volume of Brotonne over

Tancarville

and

Brotonne

Approach

Viaducts

Tancarville, 1956 (Adjusted 1973)

Brotonne, 1973

0.14 11 6.4 4.1

0.11 14 3.1 1.6

14.20 6.90 15.70 3.70 12.90

5.60 6.30 5.00 4.20 4.20

$53.40

$25.30

Increase in ESJiciency

FIGURE 12.6.

I‘ancarville

in Concrete Bridges

533

Bridge, equipment for lifting precast girders in approach

spans.

Tancarville is justified only by the fact that minimum weight was vital for the concrete stayed bridge and was maintained in the approach spans. Weight of prestressing steel is approximately half because the deck is continuous at Brotonne and the box section is more efficient than the I-girder section. More important, however, is the comparison of costs and the components thereof, Table 12.16 and Figure 12.10. One is struck by the total labor requirements for both sub- and superstructure: Tancarville Brotonne

P~nc~no GIRDER

FIGURE 12.7.

or-i P

E R

Placing precast girder over pier cap.

4.1 hr/fP 1.6 hr/ft*

In the 15 years that elapsed between the two projects, the combination of design improvements and more efficient construction methods allowed the labor to be divided by 2.5. A similar trend has been observed in other fields. For example, a complete survey of all hydroelectric

Economics and Contractual Aspects of Segmental Construction 60 $53.40 50 c 40 8 ii * 30 .r :: ho

10 QIRDCR

----I LATERAL

0 Tancarville (readjusted 1973)

SA D D L E S

DETAIL

Brotonne (1973)

FIGURE 12.10. Cost comparison between Iancarville and Brotonne approach viaducts. (5) Design overhead and fee, (4) subcontracts, (3) equipment, plant, and site overhead, (2) materials, (1) labor.

O F

GIRDER SUSPWISIO~

FIGURE 12.8. Lifting device at precast girder ends.

-I----19%

1I K5

I

1960

I

1965

I

1970

YEAR

FIGURE 12.9. Brotonne Bridge, cantilever construction of superstructure of approach spans.

FIGURE 12.11. Increase of productivity on power

projects carried out by French Electricity between 1950 and 1970 showed that the annual value of investment for each worker was multiplied by 2 without allowance for inflation and by 3 including inflation, Figure 12.11. Costwise the true gain would be somewhat less significant, because labor

rates have constantly increased faster than material rates. The comparison between other items of the cost breakdown of Tancarville and Brotonne is equally instructive. Material costs are almost equal, including the value of subcontracts (pile foundations

projects in France.

References and roadway work in both projects). The essential differences are seen in the two following areas: E q u i p m e n t , p l a n t , and job overheads: reduced from $15.70 at Tancarville to $5 at Brotonne. This difference is due essentially to increased efficiency in management but also to a climate of fierce competition. Design, overheads and fees: reduced from $12.90 for Tancarville to $4.20 for Brotonne. The same two reasons explain this drastic reduction, which also reflects the change in the overall operation of large construction companies during the last twenty years from family-owned or controlled craftsmen such as building contractors to modern management industrial companies. When Eugene Freyssinet designed his Plougastel Bridge masterpiece (see Chapter 8), he was personally involved in the project for more than three years and probably involved in little else. One generation later, an experienced engineer would have to control or at least participate in many different projects during the same period. In summary, the comparison of costs between Tancarville and Brotonne approach viaducts with prices of both projects reduced to 1973 levels is: Tancarville

$53.40/f?

Brotonne

$25.30/ftZ

Both projects were bid completely on a designand-build basis and awarded to the lowest bidder. The above costs are a true picture of the technol-

ogy and of the level of prices for the two respective periods. To estimate both projects at the level of today’s prices (1980) it would be necessary to multiply the labor rates by 2.3 and the materials and equipment rates by 1.7.

References 1. “Guidelines for Value Engineering (VE),” prepared by Task Force 19, Subcommittee on New Highway Materials, AASHTO-AGC-ARTBA Joint Cooperative Committee. 2. “Alternate Bridge Designs,” FHWA Technical Advisory T5140.12, December 4, 1979, Federal Highway Administration, Washington, D.C. 3. Richard A. Dokken, “CALTRANS Experience in Segmental Bridge Design,” Bridge Notes, Division of Structures, Department of Transportation, State of California, Vol. XVII, No. 1, March 1975. 4. A. B. Milhollin, and C. L. Benson, “Structure Design and Construction on the Vail Pass Project,” Transportation Research Record 7 17, Transportation Research Board, National Academy of Sciences, Washington, D.C., 1979. 5. James M. Barker, “North American State of the Art Current Practices,” Prestressed Concrete Segmental Bridges, Structural Engineering Series No. 6, Federal Highway Administration, Washington, D.C., August 1979. 6. Donald W. Alden, “California’s Experience with Cost Saving Contracting Techniques,” Prestressed Concrete Segmental Bridges, Structural Engineering Series No. 6, Federal Highway Administration, Washington, D.C., August 1979.

13 Future Trends and Developmenti

13.1 13.2

INTRODUCTION MATERL4IS

13.3 13.4

13.2.1 Prestressiq Tendons 13.2.2 High Strength Concrete 13.2.3 Fiber Reinforced Concrete 13.2.4 Pdymer Concre& 13.2.5 Composite Concrete Materials 13.2.6 Material Limitations SEGMENTAL APPLICATION TO BRIDGE DECKS SEGMENTAL BRIDGE PIERS AND SUBSIXUClVRES

13.1 Introduction As observed in previous chapters, prestressed concrete segmental bridges have extended the practical and competitive economic span range of concrete bridges. The Bendorf Bridge in Germany (Section 2.2) constructed in 1964 with a navigation span of 682 ft (208 m) was a monumental achievement. Because of economic differences between Europe and the United States (primarily the ratio of labor cost to material cost) it was not until the early 1970s with the JFK Memorial Causeway (Section 3.10) with a span of 200 ft (61 m) and shortly therafter the Pine Valley Creek Bridge (Section 2.7) with a span of 450 ft (137 m) that segmental construction was introduced in the United States. Today these spans are somewhat commonplace when one considers the Three Sisters Bridge (Section 1. lo), the Koror-Babelthuap Bridge (Section 2.12), and the Houston Ship Channel Bridge (Section 2.14), with spans of 750 ft (229 m), 790 ft (241 m), and 750 ft (229 m), respectively. When combined with the cable-stay concept, spans increase to 981 ft (299 m) for the Pasco-Kennewick Bridge, 1050 ft (320 m) for the Brotonne Bridge, and 1300 ft (396 m) for the Dame Point and Ruck-A-Chucky Bridges. 536

13.5

APPLICATION TO EXISTING OR NEW BRIDGE TYPES 13.5.1 Overpass Strut 13.5.2 Arches, Trusses, Rigid Frames 13.5.3 Wichert Truss 13.5.4 Stress Ribbon Bridges 13.5.5 !3pwe F- Bridges

13.6

SUMMARY REFERENCES

In earlier years these spans for concrete bridges would have been considered incomprehensible and certainly not economical. The fact that they have become achievable, only within the last decade, stands as a testimonial to rapid technological advances and to the courage and vision of those engineers who participated in this development. In the United States as of November 1980, 19 segmental bridges had been completed and there were 16 under construction, 22 in design, and 29 under study-a total of 86 bridges. We may conclude that segmental prestressed concrete construction is a viable concept for highway bridges and that there are no known major problems to inhibit its use. What, then, is the potential for segmental bridge construction in the 198Os? This chapter will look at this potential in terms of new or improved materials, potential application in bridge decks, piers, and substructure, and application to existing or new bridge superstructure types.

Z3.2 M a t e r i a l s During the nineteenth century timber, stone, and masonry were the common materials for bridge construction. Then iron, steel, concrete, and rein-

Material

forced concrete emerged successively as favorite materials, culminating in the twentieth century with prestressed concrete. The present materials used in bridge construction have some or all of the following disadvantages: weight, cost, or inherent weaknesses in one form or another. In the recent past,‘development of improved bridge systems has evolved primarily by more exact methods of calculation made feasible by the electronic computer or by innovative bridge systems such as the cablestayed and segmental types of bridges. Intensive development of the materials themselves has barely begun. 13.2.1 PRESTRESSING TENDONS

Until recently, corrosion of prestressing tendons has caused few problems and little concern. However, with the advent of segmental construction and transverse prestressing on the top flange, an increasing concern has been expressed about the potential deterioration of the tendons resulting from their closeness to the deck surface and exposure to the action of de-icing chemicals. Current methods of alleviating this concern are the use of polyethylene ducts or the possibility of epoxy coating the duct, epoxy coating post-tensioning bar tendons, and possibly epoxy coating the prestressing strand. A research effort is required to determine the production feasibility and cost; the effect, if any, that nonmetallic coatings might have on the bond of strand to concrete; and the compatibility of strains between the coating and the tendon. Another potential method uses individual unbonded strands with successive coatings of teflon, a corrosion inhibitor, and polypropylene, Figure 13.1. An old idea that may need to be resurrected is that of using glass fibers for prestressing. This material was being investigated in the 195Os,’ but for either technical or economic reasons it never reached fruition. There were problems of chemical reaction of the glass fibers with the cement; how-

7 Wire prestressing

537

ever, Owens-Corning Fiberglass Corp. has dev e l o p e d a c o a t i n g f o r g l a s s f i b e r s f o r fiberreinforced concrete. Perhaps this coating could be used for a glass fiber prestressing strand. An ultimate strength of 400 ksi (2758 MPa) and a low modulus of elasticity ranging from 6000 to 10,000 psi (41 to 69 MPa) might be expected. The high strength and low modulus would indicate a low percentage of prestress losses-a decided advantage. The high strength would produce, for a given required prestress force, fewer or smaller tendons, thus reducing congestion. Smaller tendon sizes would reduce web thickness, thus reducing dead weight and prestress force requirements, and so on. Obviously, suitable end anchorages would have to be developed. 13.2.2 HIGH-STRENGTH CONCRETE

Early prestressed concrete designs were based on 3000 psi (20.7 MPa) strength concrete. As knowledge of concrete properties and quality control increased, it became more feasible to use a 6000 psi (41.4 MPa) strength concrete for many prestressed concrete structures. In the Pacific Northwest an 8 0 0 0 p s i ( 5 5 . 2 MPa) s t r e n g t h i s r e a d i l y a n d routinely available. Use of such concrete has permitted the design of longer-span, lighter-weight concrete structures. Within the past few years it has been found that strengths of 10,000 psi (68.9 MPa) and higher can be obtained where special attention is given to (1) selecting the constituent materials, (2) proportioning the concrete mix, and (3) handling, placing, and curing the concrete. It has recently been demonstrated that the application of ultrahigh-strength concrete is not only practical but also economically feasible. Highstrength concrete, 9000 to 11,000 psi (62 to 75.8 MPa), has been used in the columns of five highrise buildings in Chicago. The concrete was produced in a local ready-mix plant and trucked to the

Bonded low frlctlon polymer (Teflon)

Corrosion Inhibitor /

’ Polyproplylene covering

FIGURE 13.1. Corrosion-resistant strand tendon.

538

Future Trends and Developments

project site. An economic study for a short tied column indicated that the cost per foot varied with the concrete strength and steel percentage as follows: $15.50 for 9000 psi (62 MPa) concrete and 1% steel compared to $39 for 4000 psi (27.6 MPa) concrete and 8% steel. Over the past five to ten years considerable research has been conducted on high-strength concrete, dealing primarily with selecting materials, developing concrete mix design criteria, and determining basic physical properties of the concretes. Very little, if anything, has been done regarding the implementation of high-strength concretes, especially in bridge structures. In an interim report,* “Applications of High Strength Concrete for Highway Bridges,” prepared by Concrete Technology Corporation for the Federal Highway Administration, a segmental bridge segment at a pier was redesigned with high-strength concrete. The purpose was to determine to what extent the thickness of the lower flange could be reduced, and in turn what effect this reduction would have on the overall moments. For purposes of this study, the Shubenacadie Bridge in Nova Scotia (Section 2.15.4) was selected as a design example. Overall dimensions of the bridge are shown in Figure 13.2. It has a 700 ft

(213.4 m) main span and 372 (113.4 m) side spans. The bridge was constructed with a 5000 psi (34.5 MPa) concrete strength and used lf in. (32 mm) diameter Dywidag bars for post-tensioning. For the analysis, the top flange was assumed to be uniform and 11; in. (292 mm) thick. The bottom flange was assumed to taper uniformly from its thickest point at the support piers to 6 in. (152 mm) at midspan. The centroid of the prestressing force was assumed to be located 59 in. (146 mm) below the top of the section-that is, centered in the top flange. AASHTO HS 20-44 was used for loading, as in the actual bridge. Prestress force was provided by la in. (32 mm) diameter Dywidag bars with a minimum yield stress of 150 ksi (1034 MPa). These bars were assumed to provide 104 kips (0.46 MN) of final prestress force each. This assumes a jacking force of 70% of the minimum yield strength and 20 ksi (137.9 MPa) losses. Maximum allowable compressive stress in the concrete was assumed to be O.+f“,, and an allowable tensile stress was assumed to be zero. Significant benefits were found in the use of high-strength concrete to reduce the thickness of the lower flange. As shown in Figure 13.3, the total flexural prestress demand is reduced by approxi-

700’. 0” 1

I

dI-

372’- 0”

I

I

AJ

ELEVATION 35’- 6”

SECTION B-B

I’- 6

,-

1’.

6”

r

+I SECTION

A-A

FIGURE

13.2.

Shubenacadie Bridge.

539

Materials 36.000

‘-6”

‘7-6” Deflection

.

FIGURE 13.4. Schematic load-deflection diagram 31,000 1 5

I 6

I 7

I a

I 9

1

10

Concrete strength SKI

FIGURE 13.3. Variation of prestress force with concrete str-ength-Shubenacadie Bridge. mately 10% as a result of the reduced dead load. The optimum lower flange thickness is about 1 ft 8 in. (508 mm), obtained at 8 ksi (55 MPa) concrete strength.

13.23 FIBER-REINFORCED CONCRETE A relatively new material that has not yet seen much application in structures is fiber-reinforced concrete. Fibers have been used to reinforce brittle materials since ancient times; straws were used to reinforce sun-baked bricks, horsehair was used to reinforce plaster, and more recently various fibers have been used to reinforce Portland cement.3 A state-of-the-art report4 prepared by AC1 Committee 544 defined fiber-reinforced concrete as “concrete made of hydraulic cements containing fine or fine and coarse aggregates and discontinuous discrete fibers.”

Several types of fibers along with several of their properties are listed in Table 13. 1.3,4 As can be seen, fibers have been produced from steel, plastic, glass, and natural materials in various shapes and sizes. Two stages of behavior in the load-deformation curve have been generally observed when fiberreinforced concrete specimens are loaded in flexure. The load-deformation curve may be considered as approximately linear up to point A in Figure 13.4. Beyond this point the curve is significantly nonlinear, reaching a maximum at point B. The load or stress corresponding to point A has been calledjirst-crack strength, elastic limit, or proportional limit, while the stress corresponding to point B has been termed the ultimate strength. Two theories have been suggested for predicting the first-crack strength of fiber-reinforced concrete: the spacing concept and the composite-materials concept. The spacing concept attempts to explain or determine the first-crack strength by a crack-arrest mechanism derived from the field of fracture mechanics. The basic mechanism that controls the

TABLE 13.1. Typical Properties of Fibers

Type of Fiber

Tensile Strength (ksi)

Young’s Modulus ( lo3 ksi)

Ultimate Elongation (SC)

Specific Gravity

Acrylic Asbestos Cotton Glass Nylon (high tenacity) Polyester (high tenacity) Polyethylene Rayon (high tenacity) Rock wool (Scandinavian) Steel

30-60 so- 140 60-100 150-550 110-120 105-125 100 60-90 70-l 10 40-600

0.3 12-20 0.7 10 0.6 1.2 0.02-0.06 1.0 10-17 29

25-45 0.6 3-10 1.5-3.5 16-20 11-13 10 lo-25 0.6 0.5-35

1.1 3.2 1.5 2.5 1.1 1.4 0.95 1.5 2.7 7.8

540

Future Trends and Developments

WITHOUT FIBERS

WlTH

FIGURE 13.6. hending.

Section A-A FIGURE

13.5.

Schematic

of

arrest

mechanism.

first-crack strength depends primarily on the spacing of the fibers.5 The crack-arrest mechanism for fiber-reinforced concrete, as presented by Romualdi and Batson, can best be described with the aid of Figure 13.5, which represents a mass of concrete in tension. The reinforcement consists of a rectangular array of rods at a spacing h and located parallel to the direction of tension stress. At some interior location an internal flaw exists in the form of a flat disk-shaped crack. The basic rationale is illustrated by section A-A, which is a side view of an internal crack between two fibers. In the presence of a gross stress the extensional strains in the vicinity of the crack tip, by virtue of the stress concentration, -are larger than the average strains. These strains, however, are resisted by the stiffer fiber, and there is created a set of bond forces (assuming that the bond between the mortar and the steel is intact) that act to reduce the magnitude of stresses at the crack tip. Under proper conditions of fiber spacing and diameter, an internal flaw could be prevented from propagating to join up with other flaws into microcracks which then join with other microcracks to form macrocracks. The basic philosophy is that if the internal flaws can be locally restrained or retarded from extending into adjacent material, thereby restraining crack propagation, the tensile-strength characteristic of the concrete is improved. The crack-arrest mechanism in bending may be idealized as indicated in Figure 13.6. When a critical strain is reached, the beam cracks; unlike the nonreinforced beam, however, the cracks do not propagate through the beam but are arrested by the fibers that span the cracks.

FIBERS

Idealized crack-arrest mechanism in

The composite-materials concept hypothesizes that the properties of fiber-reinforced concrete, including the first-crack strength, can be predicted from the individual properties of matrix and fibers. It assumes that fiber-reinforced concrete can be analyzed as conventional reinforced concrete, the main difference being that the reinforcement is shorter, thinner, and randomly distributed. Table 13.2 summarizes the improvement of the properties of a steel-fiber-reinforced concrete as compared to plain concrete.3 13.2.4 POLYiMER

CONCRETE

Concrete produced with Portland cement and airentraining agents can contain approximately 13% voids, which are interconnected and distributed throughout the mass. When this concrete is heated to drive out the chemically unbonded moisture and TABLE 13.2. Concrete Reinforced with U.S.S. Fibercon Steel Fiber

Properties Compression Flexural modulus of rupture Tensile strength Impact strength Crack and spa11 resistance Fatigue strength to 2 million cycles Abrasion Shear and torsional strength Corrosion resistance Freeze-thaw Conductivity (thermal and electrical)

Approx. Improvement over Plain Concrete 10-30s 70-300% 50-300s 150-1000% 70-300% 100% 30% 50-300% Good as or better Good as or better Conducts both

541

Materials TABLE

13.3.

Summary of Properties of Concrete-Polymer Material

Property Compressive strength (psi) Tensile strength (psi) ‘Modulus of elasticity (psi) Modulus of rupture (psi) Flexural modulus of elasticity (psi) Coefficient of expansion (in./in. “F) Thermal conductivity at 73°F (23°C) (BTU/ft-hr-“F) Water permeability (ft/yr) Water absorption (%) Freeze-thaw durability: Number of cycles Percent weight loss Hardness-impact (“L” hammer) Corrosion by 15% HCl (84-day exposure), % weight loss Corrosion by sulphates (300-day exposure), % expansion Corrosion by distilled water

then impregnated with a chemical monomer, such as methyl methacrylate (MMA), and irradiated with gamma rays, some startling changes in its properties are produced, Table 13.3.’ Tensile and compressive strength are almost quadrupled. Modulus of elasticity is increased by a factor of 1.8 and modulus of rupture by more than 3.5. A compressive stress-strain curve for this material shows complete linearity up to more than 75% of failure load, Figure 13.7.’ Thus far, research with this material has aimed toward its application in bridge decks. Problems in polymerizing large units such as bridge segments have yet to be solved. Practical resolution of these problems could offer a tremendous advantage for concrete structures.

Concrete with up to 6.7 Weight % Loading of Polymethyl Methacrylate, Co-60 Gamma Radiation Polymerized

Concrete Control Specimen (Type II Cement) 5,267 416 3.5 x 106 739 4.3 x 10” 4 . 0 2 x 1O-6 1.332 6.2 x 5.3

20,255 1,627 6.3 x lo6 2,637

6.2

x 10”

5 . 3 6 x 1O-6 1.306 0 0.29

lo-’

2,420 0.5 55.3 3.6

590 26.5 32.0 10.4 0.144 Severe

0 No attack

attack

18

,6

,

-

,

,

, ( ,

,

,

,

,

,

,

y , I

Impregnated CP 5.4 w t % MMA

COnCrete

,

,

,

Failure

14

12 -z ,” 1 0 8 2 E G

8 Unimpregnated CP concrete 6

13.2.5 COMPOSITE CONCRETE MATERIALS

Assuming that the materials previously discussed can be developed to a point of practical usage, what improvement in properties might be expected if these materials were combined? Sukiewicz and Viralas have presented flexural-load-versus-deflection data for concrete and composite materials, Table

loo0

2ooa

3000

4000

Compressive strain (microinches/inch)

FIGURE

1 3 . 7 . Compressive stress-strain curve for

MMA-impregnated concrete.

542

Future Trends and Developments TABLE 13.4. Relative Load Versus Relative Deflection

Plain concrete Steel-fiber reinforced Polymer impregnated Polymer impregnated and steel-fiber reinforced

Approx. Relative Max. Load

Approx. Relative Midspan Deflection at Max. Load

5 18 20 100

5 20 13 105

13.4. It is obvious that a vast improvement in behavior and toughness can be expected. 13.2.6 AMATERIAL LIMITATIOSS With improved material properties, not only would structures become lighter but also the depth of superstructure and thickness of individual elements would be reduced. There are some practical limitations, however, as to how much the thickness of a web, for example, may be reduced. The practical limitations of placing the concrete in the forms and of congestion of supplemental reinforcement and prestressing tendons must still be considered. To some extent this could be alleviated by the use of external tendons, as implemented in the Long Key Bridge in Florida, Figure 6.53. This also has the advantage of reducing the complexity of fabrication for precast segments. Perhaps a more important limitation in using materials with improved properties is that at some point in the design, stress no longer becomes the controlling criterion. Deformations, both global and local, may govern. Because of the reduced section required from a strength point of view, there may be more concern not only with flexibility of the structure in a global sense but also with the possibility of web buckling and limberness of the deck slab.

13.3

prestressing has been used to a greater extent in the construction of segmental bridges-to provide greater load capacity and load distribution for large overhanging flanges and between adjacent single-cell box girders. Although a few bridge designs have included transverse prestressing, much greater use could be made of it for more economical bridge structures. For replacement of the decks on existing bridges, precast prestressed concrete segmental construction offers great advantages, only some of which can be associated with identifiable costs. As with the segmental box girder, a full-depth segmental panel bridge deck may be precast in short segment lengths longitudinally and may be full deck width or partial deck width, Figure 13.8, depending on the width of deck required for a particular application. Also, in addition to the transverse prestressing, segmental bridge decks may be conceived as having expoxied transverse joints and longitudinal prestressing. A transversely prestressed segmental full-depth panel bridge deck, Figure 13.9, has been proposed by T. Y. Lin International as an alternative design for SR 182, Columbia River Replacement Bridge, in the state of Washington. This proposal has the following features: 1.

Precast full-depth panels of crete to reduce dead load.

lightweight

2. Transverse prestressing to achieve large cantilever overhangs and thus economies in the superstructure. 3.

Attachment of the panels to the superstructure with shear studs in block-outs of the panel to achieve composite action with the superstructure.

Segmental Application to Bridge Decks

To date, there has been very little use of precasting, prestressing, and segmental construction for bridge decks. Transverse prestressing has been used in the top flange of large, cast-in-place on falsework, concrete box girders. Lately, transverse

con-

I

I

FIGURE 13.8. Deck configurations.

I

543

Segmental Bridge Piers and Substructures

made to act compositely by means of shear-transfer devices placed at regular intervals through blockouts cast into the deck. To date, the incremental launching method has been implemented for the construction of ten bridge decks in Switzerland.lo*ll

13.4

Segmental

Bridge

Piers

and

Substructures

FIGURE 13.9. SR-182 Columbia River Bridge. 4. Longitudinal prestressing of the deck to maintain a compression across the transverse joints. Another segmental method of constructing a bridge deck is a transfer of technology from the incremental launching of segmental box-girder bridges.9~10~11 This methodology, as applied to bridge decks, has been pioneered in Switzerland and consists of the following operations: 1.

The casting of a convenient segment length of bridge deck behind an abutment, Figure 13.10, or at midlength of the bridge, Figure 13.11, whichever is more convenient. Segment length is normally 65 to 80 ft (20 to 25 m).

2. The jacking forward of the segment, Figure 13.12, onto the flanges of the steel superstructure, Figure 13.13, thus freeing the casting bed. 3.

Preparing the casting bed for concreting the next segment.

4. Repeating the cycle until completion of the bridge deck. The finished deck, therefore, consists of segments that have been incrementally cast and longitudinally launched. As in conventional structures the deck is attached to the superstructure and

Piers do not have to be massive solid cross sections; a tubular cross section may be more effective and more economical. In the United States it is generally felt to be more economical to cast a solid pier. However, for tall piers the economics of solid-pier casting should be evaluated against the cost of the additional dead load the pier is supporting and transferring to the foundations. It may be desirable to precast the pier as tubular segments that are prestressed bertically to each other as well as the foundation. In the Vail Pass Bridge structures the piers were constructed of diamond-shaped segments, stacked vertically and post-tensioned to the foundations, Figure 13.14. Footings were cast in place with ducts to allow the placing of prestressing tendons. Other examples of segmental application to piers and substructure, previously discussed, are: 1.

Linn Cove Viaduct, Section 6.3.2, Figure 6.17

2.

I-205 Columbia River Figures 5.25 and 5.26

3.

Sallingsund Bridge, Section 5.4.2, Figures 5.20 and 5.21

Bridge,

Section

5.4.3,

The Long Key Bridge, Section 6.5.1, used precast V piers; and the Seven Mile Bridge, Section 6.5.2, used precast segments stacked vertically. It is to be hoped that these concepts will be refined and utilized for future structures where applicable.

FIGURE 13.10. Incremental launching, behind abutment, from reference 10 (courtesy of American Society of Civil Engineers).

544

Future

Trends

and

Developments

FIGURE 13.11. Incremental launching, midlength of bridge, from reference 10 (courtesy of American Society of Civil Engineers).

. Prestressmg

tendons

1 Precast segments

FIGURE deck slab.

13.12.

Incremental

launching,

jacking

of

FIGURE 13.14. Vail Pass Bridges, segmental pier.

FIGURE 13.13. Incremental launching, deck on superstructure.

13.5

Application to Existing or New Bridge Types

With the exception of the Pasco-Kennewick and Dame Point cable-stayed bridges, the implementation of the concept of segmentally constructed bridges in the United States has been limited to the girder type of bridge. In other parts of the world,

the segmental concept has been applied not only to the cable-stayed bridge, but also to rigid frames, arches, trusses, and to a limited extent to overpass structures. Segmental construction is versatile and should not be stereotyped to girder bridges only; it can be applied to other types of bridge construction. 13.5.1 OVERPASS STRCrCTURES

The main application of prestressed concrete segmental bridge construction has been to the long or intermediate span range and to viaducts. However,


545

this method of construction has been applied to highway overpass structures, proving its versatility. Examples of overpass structures previously presented are: 1. Rhone-Alps Motorway, Section 3.15 2 . Motorway Overpasses in the Middle East, Section 8.6.4 The Federal Highway Administration (FHWA) in cooperation with the American Association of Highway and Transportation Officials (AASHTO) is embarking on a study of the feasibility of standard sections for segmental box girder bridges. If feasible, standardization of sections, especially for overpass bridges, could provide additional economy for bridge construction. An important and costly problem in building overpass structures over heavily traveled highways and freeways is that of traffic control during construction. An idea that might minimize this problem can be borrowed from the construction procedure for the Vienna Motorway cable-stayed segmental bridge. Because construction was not allowed to interfere with navigation on the canal, the structure was built in two 364 ft (111 m) halves on each bank and parallel to the canal. Upon completion, the two girder halves were swung into final position, Figure 9.66, and a cast-in-place closure joint was made. In other words, each half is constructed as a one-time swing span. This concept was considered for a long skewed overpass in Illinois, but the contractor elected a more conventional procedure. 13.5.2 ARCHES, TRUSSES, RIGID FRAMES

The adaptation of segmental concepts to arches, trusses, and rigid frames has yet to be implemented on the North American Continent. As indicated in Chapter 8, there are ample examples to indicate that segmental technology can be used for these types of structures. As previously noted and adequately illustrated, segmental construction should not be stereotyped to girder bridges only. 13.53 WKHERT TRUSS

The resurrection of the Wichert truss principle,12 Figure 13.15, might yield economies in segmental cantilever construction. This type was developed for structural steel trusses and has the curious property of providing a fixing moment while remaining statically determinate. The fixing moment is provided as a function of the geometry of the

FIGURE 13.15. Wichert truss principle.

quadrilateral support, any desired degree of fixing moment being obtained by arranging the geometry of that quadrilateral, with the structure remaining determinate. Consequently there is no danger of complications from settlement of supports or parasitic moments caused by prestressing. The Wichert truss principle has, among others, the following advantages: 1. 2. 3. 4.

Economy of girder material Economy of foundations Elimination of intermediate hinges Stresses unaffected by temperature difference between chords or flanges

The Smithy Wood Footbridge, Figure 13.16, in the United Kingdom is one of three footbridges constructed in about 1970 over the M-l Motorway using the principle of the Wichert Truss. 13.5.4 STRESS RIBBON BRIDGES

Another new type of bridge, introduced relatively recently, is the Spannbandbriicke or stress ribbon bridge. 13~~ Its origin is obscure and can be traced back to early societies. Basically the early versions consisted of wood planking supported directly on main catenary cables. It is still used in certain parts of the world, Figure 13.17. The first modern attempt at the implementation of this concept was in 1958, when Ulrich Finsterwalder unsuccessfully entered the concept in the Bosporus Bridge competition, Figure 13.18, and again in 1961 for the Zoo Bridge at Cologne. The first successful construction of a stress ribbon structure was m 1963 and 1964 in Switzerland for a conveyor-belt bridge at the Holderbank-Wildeck Cement Works, with a span of 710 ft (216.4 m). The Freiburg, Germany, footbridge, Figure 13.19, constructed in 1969 and 1970 has an overall length of 448 ft (136.5 m) with a center span of 130 ft (39.5 m). Notice that the tops of the piers, in effect, form a large-radius saddle for the catenary cables. The deck has a width of 14.4 ft (4.40 m) and a thickness of 10 in. (0.25 m). In 1971 the Rhone Genf-Lignon stress ribbon bridge, Figure 13.20,

FIGURE

13.16.

Smithy Wood Footbridge,

FIGURE 13.17. Early stress ribbon bridge in the Orient.

FIGURE

13.18.

FIGURE 13.19. Freiburg Stress Ribbon Footbridge.

Stress ribbon concept for the Bosporus Bridge.

was constructed with a single span of 446 ft (136 m) and a width of 10.2 ft (3.10 m). This type of structure has used prefabricated transverse and longitudinal prestressed precast segments supported on the main catenary cables. Wind-tunnel tests have indicated that a stress rib-

bon bridge is safe against torsional oscillation.15 It can have a relatively flat sag, such that the grade at the abutments and piers can be kept at approximately 4%. Its largest disadvantage is the large abutments required to sustain the large tensile force in the main cables.


FIGURE 13.20. Rhone Bridge.

547

Genf-Lignon Stress Ribbon

13.5.5 SPACE FRAME BRIDGES

FIGURE 13.21. Elevations of Bul)i)arl Bridge Proposal prepared by Bouygues, Paris.

In 1980, one of the authors served in an advisory capacity to the Kuwait Ministry of Public Works to evaluate responses to a request for proposals for the Bubiyan Bridge Project. An interesting proposal was submitted by Bouygues, a French firm. This proposal consisted of a three-dimensional truss or space frame concept, Figures 13.21 through 13.23. The concept of prestressed concrete trusses is not a new one. Concrete trusses have been used in building construction and in bridges (Chapter 8) in various projects throughout the world. For example: The Mangfall Bridge in Austria (Section 8.7.2) is a three-span, cast-in-place, prestressed concrete structure. It may be described simply as a box girder consisting of solid top and bottom flanges connected by two vertical webs, which are trusses. The Rip Bridge in Australia (Section 8.7.3), L. just north of Sydney, is a three-span cantilever arch-truss structure. The upper chord (roadway slab), diagonal and vertical truss members, and lower chord are composed of precast elements, which are made integral by cast-inplace concrete and post-tensioning. 3. At least three prestressed concrete cantilever arch-truss bridges have been constructed in Yugoslavia, including the Kirk Bridge (Section 8.5,4), which presently holds the record for the longest concrete arch in the world, a 1280 ft (390 m) span. 4. Other prestressed concrete truss bridges have been constructed in France, the U.S.S.R., and Japan.

FIGURE 13.22. Isometric of Bubiyan Bridge Proposal prepared by Bouygues, Paris.

FIGURE 13.23. Construction-stage model of Bubiyan Bridge Proposal.

All the structures mentioned above have one thing in common: the prestressed concrete trusses are all oriented in a vertical plane. The concept is the same as in conventional truss bridges constructed of structural steel members. The three-dimensional truss concept presented for the Bubiyan Bridge is essentially a multitriangular-cell concrete box girder wherein the longitudinal solid webs are replaced by an open lattice system of trusses. Because the lattice truss webs are oriented in an inclined plane, as opposed to a vertical plane, adjacent trusses have common node points (intersection of diagonal and vertical truss members with the flanges). This spatial geometry then forms in the transverse direction another system of trusses. Thus, the flanges are connected by a system of inclined orthogonal trusses (a system of mutually perpendicular trusses), Figure 13.24. Because the trusses are inclined to each other, with the diagonal and vertical members intersecting at common node points, they form a space frame composed of interconnecting pyramids. Thus, the structural behavior of the bridge with regard to distribution of load resembles that of a two-way slab in building construction. This structural concept is new in regard to its application to a bridge structure. However, the concept of a space frame truss has been previously applied to roof structures for large column-free sport facilities, auditoriums, civic centers, and the like. These space structures have been constructed primarily of metallic (steel or aluminum) tubular sections. There is no reason to believe that with the current state of the art in prestressed concrete, segmental construction and existing concrete truss construction, a prestressed concrete space frame

FIGURE 13.24. Model of typical segment with top

flange removed (note external post-tensioning tendons), Bubiyan Bridge Proposal.

concept cannot be consummated-in particular for a bridge structure. The advantage of this concept is that as a result of the “openness” of the trellis framework of trusses the dead or self-weight of the superstructure is much less than that of conventional prestressed concrete construction. This comparative reduction in weight of the superstructure reduces the dead load to be transmitted to the substructure and thus reduces the mass of the substructure, with resulting economies. Further, there is an advantage in the manner the load is distributed throughout the structure. That is to say, there are many load paths. In the unforeseen event of a member failure, the load would redistribute by seeking an alternative load path. Therefore, there is a greater degree of redundancy, which means that there is greater safety from a collapse failure. Or, in other words, by virtue of the spatial geometry there is an inherent reserve capacity. In all other respects the fabrication and erection of the superstructure is consistent with state-ofthe-art conventional prestressed, precast segmental construction, including the external prestressing. Although the concept of a space frame structure is new to bridge construction, its newness is only in assembling existing concepts into a single concept.

13.6 Summary The last decade has seen considerable changes in bridge design and construction, many of which have been evolving since the 1950s. As we move into the 198Os, we must remain aware of change. Research already underway on inproved materials may have dramatic impact in the indusry. Applications of new systems to existing bridge types are being attempted along with new and improved types of bridge structures. Many of the improved materials and new concepts will reach practical application; others will be abandoned for technical or economic reasons. Unforeseen improvements in materials and new types of bridge structures are certain to evolve in the next decade, which proniises to be one of excitement and challenge. As engineers whose basic responsibility is the betterment of mankind, we must be constantly open to new concepts and ideas that will technically and economically improve the structures we attempt to build. However, at the same time we must anticipate a new generation of problems that changes in methodology are certain to bring. Of

549

References prime importance, we must not fall into a trap of oversophistication in d e s i g n a t t h e e x p e n s e o f simplicity and thus economy in construction. As we strive for longer spans and improved means of constructing bridges, it would be well to remember the words of F. Stiissi, an eminent Swiss engineer: The poblem of‘ Long Spans has always fascinated the specialist as well as the layman. The realization of a bridge with a length of span hitherto unattained not only requires great technical knowledge and capability, but also intuition and creative courage; it signifies a victory over the forces of nature and propess in the battle against human insufjc&y.

This philosophy applies not only to the achievement of longer spans, but also to the changing technology of the future.

References 1. I. A. Rubinskv and A. Rubinsky, “A Preliminary Investigation of the Use of Fiberglass for Prestressed Concrete,” Magazine of Concrete Research, September 1954. 2. James E. Carpenter, “Applications of High Strength Concrete for Highway Bridges,” Interim Report, July 1979, Concrete Technology Corporation, Contract No. DOT-FH-1 l-9510 (unpublished report). 3. W. Podolny, Jr., “Properties of Fiber-Reinforced Concrete,” Highway Focus, Vol. 4, No. 5, October 1972, Federal Highway Administration, Washington, D.C. 4. “State-of-the-Art Report on Fiber Reinforced Concrete,” Reported bv AC1 Committee 544,Journal uf the .4merican Concrete Institute, November 1973.

5 . J. P. Romualdi and G. B. Batson, “Mechanics of Crack Arrest in Concrete,” Proceedings of the ASCE, Vol. 89, No. EM 3, June 1963. J. P. Romualdi and G. B. Batson, “Behavior of Rein6. forced Concrete Beams with Closely Spaced Reinforcement,“journa/ of the American Concrete Institute, Vol. 60, Title No. 60-40, June 1963. 7 . “Polymer Concrete,” FIP ‘Votes, 38 January/ February 1972, Federation Internationale de la Precontrainte. 8 . Jan Sulkiewicz and Juhani Virola, “Future Aspects of Bridge Construction,” International Civil Engineering Monthly, Vol. III, No. 2, 1974 (Jerusalem). 9 . G. H. Beguin, “Vergundbrticken-.4usfuhrungsprobleme beim Fahrbahn plattetr-Schiebever fahren, Der Stahlhaum, Heft 12, Dezember 1975. 10. G. H. Beguin, “Composite Bridge Decking b\ Stage-Deck Jacking,” Journal qf t h e S t r u c t u r a l D i v i sion, AXE, Vol. 104, No. ST 1, January 1978. 11. Rene Rvser, Discussion of “Composite Bridge Decking by Stage-Deck Jacking,” Journal of the Structural DiGsion, .4SCE, Vol. 104, No. ST 10, October 1978. 12. D. B. Steinman, The Wichert Truss, Van NostrandReinhold, New York, 1932. 13. Rene Walther, “Stressed Ribbon Bridges,” International Civil Engineering Monthly, Vol. II, No. 1, 1971/1972 (Jerusalem). 14. Heinz Nehse, “Spannbandbriicken,” in Festschrift Ultih Fiwterwalder 50 Jahreftir Dywidag, Dykerhoff U Widmann, A.G., Munich, 1973. 15. Ulrich Finsterwalder, “Free-Cantilever Construction of Prestressed Concrete Bridges and MushroomShaped Bridges,” First International Symposium, Concrete Bridge Design, American Concrete Institute

Publication SP-23, Detroit, 1969.

Index of Bridges Abeou

Aqueduct

(France),

345

Changis

(France), 360

Akayagawa (Japan), 379,458,463 Algiers (Algeria), 393 Alpine Motorway (France), 83, 129-134, 135, 139, 390, 489,

Chillon Viaduct (Switzerland), 18, 99-103, 162,,167, 226, 254, 256, 503, 512-514 Choisy-le-Roi (France), 12, 13, 17, 20, 83-88, 139, 155, 165, 169,

510,515,545 A-l Motorway Overpass (France), 352

226, 253, 254, 255, 361,362,468,494, 509 Clichy (France), 233, 234, 449-452, 478, 479, 482 Cline Avenue (Indiana, U.S.A.), 527

Ampel (Holland), 510 Anet (France), 360 Angers (France), 122, 515 Aramon (France), 98, 100, 5 14

Colorado River at Yuma (California, U.S.A.), 529 Columbia River at Astoria (Oregon, U.S.A.), 238 Columbia River, I-205 (Oregon, U.S.A.), 142-144, 158, 161, 162, 169, 238, 241, 543 Columbia River, SR 182 (Washington, U.S.A.), 542, 543 Conflans (France), 95, 96,469, 509

Ardrossan Overpass (Canada), 8, 11 Arnhem (Holland), 57-59 Arrabida (Portugal), 375 Ashidagawa (Japan), 458, 462

Corde (France), 228 Courbevoie (France), 20, 88, 89, 165, 226, 254, 468, 494, 509 Criel Viaduct (France), 333, 335,344

Balduinstein, see Lahn River Barranquilla (Colombia), 401-404 Barwon River (Australia), 401, 402 Basra (Iraq), 346 Bayonne (France), 510 Bear River (Canada), 83, 84, 108-l 12 Bendorf (Germany), 19, 35-37, 164, 167, 283-284, 296, 536 Benton City (Washington, U.S.A.), 401, 402 Bezons (France), 31-33 Black Gore Creek, see Vail Pass

Dame Point (Florida, U.S.A.), 401, 402, 405, 433-436, 536, 544 Danube Canal (Austria), 401, 402, 427-431, 545 Danube River at Worth (Germany), 26, Dauphin Island (Alabama, U.S.A.), 256, Denny Creek (Washington, U.S.A.), 22, Deventer (Holland), 20, 105, 106, 162,

345, 257, 34, 515,

346 258 304311 516

Digoine (France), 455, 456, 458 Dijon (France), 476

Blois (France), 98, 100, 233, 245, 468, 513, 514 Boivre Viaduct (France), 350 Bonhomme (France), 15, 387-392

Dnieper River (U.S.S.R.), 401, 402 Drummondville (Canada), 8,9 Dumbarton (San Francisco, Calif., U.S.A.), 530

Bonpas (France), 509 Borriglione Viaduct (France), 344

Dusseldorf-Flehe (Germany), 21, 23, 303-307

Bospoms Bouguen

Eastern Scheldt (Holland), 134136, 139 East Huntington (West Virginia, U.S.A.), Elztalbrucke, (Germany), 296-300

(Turkey), 545, 546 (France), 34, 38-40

Bourg-Saint AndCol (France), 509 Briesle Maas (Netherlands), 385-388 Brotonne (France), 17, 29, 30, 198, 199, 233, 234, 236, 374, 401, 405, 419-427, 481, 482, 530-535, 536 Bubiyan, Kuwait, 547-548

Eel River (California, U.S.A.), 529 Esbly Bridge (France), 5, 360 Eschachtal

(Germany),

73,

80

B-3 Viaduct (France), 14, 18, 124, 128, 139, 145, 162, 165, 468, 507, 510,515, 516, 517

Felsenau

Calix (France), 139, 140 Canadians Interchange (France),

Flying Lever Bridge, 3 Fontenoy (France), 25 Foyle River (N. Ireland), 401, 402

476-478

Canal du Centre (Belgium), 401,402 Capt. Cook (Australia), 20, 136-139 Caracas Viaducts (Venezuela), 363-372, 375, 530 Carpineto Viaduct (Italy), 401, 402, 404 Center Canal (France), 455 Cergy-Pontoise (France), 444, 445 Chaco/Corrientes (Argentina), 401-404, Champigny (France), 173, 343

408-410,

401,402

(Switzerland), 168

F-9 Freeway (Australia), 145 Firth of Forth (U.K.), 3

Freiburg (Germany), 545, 546

432

Garden (Frame), 509 Gennevilliers (France), 52-55, 164, 478, 480 Givers (France), 151, 152, 163 Gladesville (Australia), 371-374, 375 Grand’Mere (Canada), 55-58

551

Index of Bridges

552 Great Belt (Denmark), 14, 401-405, 430-433 Gronachtal (Germany), 345 Guadalupe River at San Jose (California, U.S.A.), 529 Guadiana

Viaduct (Portugal), 300-301

Marne River Bridges (France), 5, 357, 361, 362, 363 Marolles (France), 344 Medway (U.K.), 7 1 Miller Creek, see Vail Pass Mirabeau (France), 21, 509

Hartel (Holland), 103-105, 494, 509

Morand

Herval (Brazil), 31 Hobbema (Canada), 8, 10 Hokawazu (Japan), 378

Morlaix (France), 163 Moulin a Poudre (France), 163 Moulin-les-Metz (France), 476, 477

Holderbank-Wildeck (Switzerland), 545 Houston Ship Channel (Texas, U.S.A.), 28, 68-71, 161, 169,

Mount Street (Australia), 401, 402 M-25 Overpass (U.K.), 401, 402

212-222, 232, 238-239, 241, 536 Hyobashigawa (Japan), 457, 459

Muhlbachtalbrucke (Germany), 338-339 Muscatuck (Indiana, U.S.A.), 84

Incienso (Guatemala), 71, 74, 75

Napa

Ingolstadt (Germany), 345 Inn (Germany), 345 Interstate I-266, Potomac River Crossing, see Three Sisters Bridge

529 Napa River (Vallejo, California, U.S.A.), 10

Iquacu (Brazil), 375 Islington Avenue Ext. (Canada), 84

Neckarburg (Germany), 370, 376-380 Neckarrews (Germany), 33 New Melones (California, U.S.A.), see Stanislaus New Orleans, Greater (Louisiana, U.S.A.), 3 Niesenbachbrucke (Austria), 378, 380-382

JFK Memorial Causeway (Texas, U.S.A.), 27, 83, 84, 109, 112114,536 Joinville (France), Juvisy

164,

479

84,

142,

143

Kettiger Hang (Germany), 293-294 Kimonkro (Ivory Coast), 344 Kipapa Stream (Hawaii, U.S.A.), 72, 76, 169, 530 Kirk (Yugoslavia), 15, 16, 374, 375, 379, 382-384, 547 Kishwaukee (Illinois, U.S.A.), 84, 141, 142, 143, 169 Kisogawa (Japan), 458,461 Koblenz (Germany), 162, 167 Kochertal (Germany), 45-46, 47, 48, 168 Koches Valley (Germany), 345 Koror Babelthuap (Pacific Trust Territory), 13, 30, 61-63, 81, 536 Krahnenberg (Germany), 293-295, 296 Krummbachbrucke (Switzerland), 379 Kwang

Fu

(Taiwan),

River (Napa, California, U.S.A.), 59-61, 81, 527,

Natorigawa

(Japan),

457,460

River

Oissel (France), 163, 476 Oleron Viaduct (France), 2, 18, 165, 175, 177, 241, 243-245, 511, 512,514 Olifant’s River (South Africa), Oh Viaduct (France), 333-335,

20, 22, 96-99, 108, 157, 158, 249, 468, 494, 504, 506, 508, 26, 331-333, 343, 441, 452-454 344

Oosterschelde (Holland), see Eastern Orleans (France), 478 Ottmarsheim (France), 145, 165 Overstreet Ounasioki

Scheldt

(Florida, U.S.A.), 145, 146 (Finland), 281, 284-285

Pag (Yugoslavia), 375, 378 Paillon (France), 344 Paris Belt, Downstream, (France), 91-94, 155, 233, 234, 263-266, 468, 509, 511

401,402

Lacroix Falgarde (France), 34, 40-41 Lahn River (Germany), 11, 33 Lake Maracaibo (Venezuela), 400-403, 405-408,

Paris Belt, Upstream, (France), 94, 95, 245-248, 468, 509, 511 Paris-Lyon high-speed railway line (France), 453-457 Parrots Ferry (California, U.S.A.), 72, 77 432

La Voulte (France), 34, 441, 442 Lievre River (Canada), 83, 84 Linn Cove Viaduct (N. Carolina, U.S.A.), 23, 35, 84, 178-181, 228, 233, 282, 290-293, 308, 503, 525, 543 Loisach (Germany), 301-303 Long Key (Florida, U.S.A.), 18, 22, 23, 84, 159, 169, 308-313, 489, 524-525, 526, 542, 543 Luc Viaduct (France), 333-335, 344 Luzancy (France), 11, 13, 82, 357, 359, 360 Madison County (Tennessee, U.S.A.), 5 Magliana (Italy), 401-404 Magnan Viaduct (France), 72, 77, 256, 257, 479 Mainbrucke (Germany), 401,402, 405,410-412 Mangfall (Austria), 395, 396, 547 Marne la Vallee

442-444

North Main Street Viaduct (Ohio, U.S.A.), 528 Nuel Viaduct (France), 344

(France), 95, 96, 233, 254, 509

Kakogawa (Japan), 441,462 Kennedy (France), 476 Kentucky River (Kentucky, U.S.A.),

(France),

Viaduct (France), 441, 444-448, 517

Pasco-Kennewick (Washington, 418-419, 536, 544

U.S.A.),

17,

84,

401,

402,

405,

Penn DOT Test Track (Pennsylvania, U.S.A.), 84 Pfaffendorf (Germany), 393 Pierre Benite (France), 21, 89-91, 92, 468, 510 Pine Valley, California (U.S.A.), 27, 29, 34, 35, 46-52, 81, 161, 169, 523, 529, 536 Pleichach Viaduct (Germany), 295-296 Plougastel (France), 15, 31, 355, 358, 359, 375, 535 Polcevera Viaduct (Italy), 401-403 Pontchartrain (Louisiana, U.S.A.), 6, Pretoria (South Africa), 401, 402 Puente de1 Azufre (Spain), 7 1, 73 Puteaux (France), 73, 78, 151 Pyle (France), 229 Quebec (Canada), 3 Querlin Guen (Germany), 345

7

Index of Bridges Ravensbosch Valley (Holland), 329-331 Reallon (France), 151, 153 Red River (Arkansas, U.S.A.), 527, 526 Rhone Genf-Lignon (Switzerland), 545, 547 Richmond-San Rafael (California, U.S.A.), 238 Rio Caroni (Venezuela), 24, 321-327, 345 Rio Caroni railway (Venezuela), 464 Rio Esla (Spain), 375 Rio Niteroi (Brazil), 20, 22, 106-108, 139, 169, 176, 178, 190, 200, 201, 515, 516 Rio Tocantis (Brazil), 71, 72 Rip Bridge (Australia), 14, 395-397, 547 Roche (France), 455, 456 Rombas Viaduct (France), 282, 288-290, 317 Ruck-A-Chucky (California, U.S.A.), 401, 402, 436-439, 536 Saint Adele (Canada), 37-39 Saint Andre de Cubzac (France), 101, 113-116, 122, 162, 165, 489,504,510 Saint Clair Viaduct (France), 375 Saint Cloud (France), 14, 18, 29, 30, 114, 117-122, 139, 150, 160, 162, 165, 205, 208, 234, 503-505, 508, 514, 515 Saint Isidore (France), 479 Saint Jean (France), 41-44, 163, 235, 265-269 Saint Michel (France), 384-385 Sallingsund (Denmark), 101, 122-l 25, 158, 167, 236-238, 240, 489,5 15,543 Sando, (Sweden), 375 San Joaquin River at Ant&h (California, U.S.A.), 529 San Mateo-Hayward (California, U.S.A.), 238 Saone River (France), 454-458 Sathom (Thailand), 348 Saulieu (France), 455 Shubenacadie (Canada), 71, 74, 81, 538,539 Schwarzwasserbrucke (Switzerland), 379 Sebastian Inlet (Florida, U.S.A.), 10, 12 Seine River (France), 455,456 Serein (France), 455, 456 Setubal (Argentina], 72, 75, 76 Seudre Viaduct (France), 98, 101, 165 SevenMile (Florida, U.S.A.), 22, 28, 159, 169, 312-314, 489, 526, 543 Sheldon (New York, U.S.A.), 11 Shepherds House (U.K.), 339-343 Shogun’s Bridge (Japan), 2 Sibenik (Yugoslavia), 375, 378

553

Siegtal (Germany), 19, 43-45 Smithy Wood Footbridge (U.K.), 545, 546 Stanislaus River (California, U.S.A.), 529 Sugar Creek (Indiana, U.S.A.), 84 Tancarville (France), 419, 530-535 Tarento (Italy), 168 Tauemautobahn (Austria), 141, 142 Tempul Aqueduct (Spain), 400-402,405 Tet Viaduct (France), 344 Three Sisters Bridge (District of Columbia, U.S.A.), 27,34, 536 Tie1 (Netherlands), 401, 402, 412-418 Tonneins (France), 355 Torcy Viaduct (France), 447-449 Tours (France), 509 Tourville (France), 476 Trent Viaduct (U.K.), 140, 141 Tricastin (France), 73, 79, 80, 151, 153 Trilbardou (France), 360 Turkey Run State Park (Indiana, U.S.A.), 84 Ussy, France, 360 Vail Pass (Colorado, U.S.A.), 84, 140, 141, 228, 285-288, 523, 524,543,544 Val Ristel (Italy), 25, 327-330 Van Staden (South Africa), 375, 378 Var Viaduct (France), 345 Vejle Fjord (Denmark), 63-67, 162, 167, 530 Veurdre (France), 354, 355 Villeneove (France), 355, 356 Viosne (France), 164 Waal River (Holland), see Tie1 Wabash River (Indiana, U.S.A.), 26, 34, 84, 324, 335-338 Wadi Kuf (Libya), 401-403, 407, 408, 432 Walnut Lane (Pennsylvania, U.S.A.), 5, 6 Wandipore Bridge (Bhutan), 2 Weirton-Steubenville (West Virginia, U.S.A.), 401, 402 Woippy (France), 479, 481 Worms (Germany), 11, 13 Zilwaukee (Michigan, U.S.A.), 84, 144, 145, 158, 162, 169, 526, 527 Zoo (Germany), 545

Index of Personal Names Abrahams, M. J., 30, 147, 224 Alden, Donald W., 535 Ballinger, C. A., 30, 147, 224 Barker, James M., 535 Batson, G. B., 540, 549 Baumgart, E., 31 Baur, Willi, 24, 30, 323, 352 Beloff, G., 20, 138, 139 Bender, Brice, 104, 105, 330, 331 Benson, C. L., 535 Bequin, G. H., 549 Best, K. H., 353 Bezouska, T. J., 5 17 Bezzone, A. P., 81 Bockel, Manfred, 296, 298, 302 Bouchet, Andre, 147 Breen, John E., 109, 112, 113, 147 Brown, R. C., Jr., 147 Brunei, 339 Bryant, Walter, 419 Bums, N. H., 147 Caquot, Albert, 31 Carpenter, James E., 549 Chang, F. K., 439 Cooper, R. L., 147 Dill, R. E., 4 Dohring, C. E. W., 4 Dokken, Richard A., 81, 535 Downing, Dale F., 81 Eiffel, Gustave, 41, 43 Esquillan, M., 34, 375

Hadley, Homer M., 401, 439 Hale, Phil, 61 Heinen, Richard, 81, 411, 412 Henneberger, Wayne, 109 Hoyer, E., 4 Jackson, P. H., 4 Kashima, S., 147 Kingston, R. H., 353 Lacey, G. C., 147 Lenglet, C., 440 Leonhardt, Fritz, 24, 25, 26, 30, 149, 224, 323, 325, 345, 405, 420,439 Lin, T. Y., 30,436,440 Lindberg, H. A., 67 Lu, H. K., 440 Macintosh, D. W., 81 Maddison, M., 353 Mangnel, G., 4 Mathivat, J., 147, 260, 280 Milhollin, A. B., 535 Moth, Elizabeth B., 30 Morandi, Riccardo, 400, 401, 403, 404, 407, 430, 432, 439 Muller, Jean, 12, 30, 147, 224, 399 Navier, Claude Louis Marie, 83 Nehse, Heinz, 549 Podolny, W., Jr., 30, 147, 224, 320, 439, 549 Pope, Thomas, 3

Feige, A., 439 Finsterwalder, Ulrich, 11, 23, 30, 33, 36, 81, 320, 401, 402, 404,410,432, 5 4 5 , 5 4 9 Fleming, J. F., 439 Ereyssinet, E., 4, 5. 11, 31, 210, 226, 354, 355, 363, 375, 392, 394, 398, 399,535

Redfield, C. M., 440 Resal, 196 Romualdi, J. P., 540, 549 Rothman, H. B., 439 Rubinsky, A., 549 Rubinsky, I. A., 549 Ryser, Rene, 549

Gallaway, T. M., 147 Garrido, L. A., 403 Gerwick, Ben C., 147, 238, 280 Graham, H. J., 440 Grant, Arvid, 325, 326, 353, 418, 419, 420, 439, 440 Gray, Normer, M., 401, 439

Scalai, J. B., 30, 439 Schambeck, H., 320,439 Schell, Herb, 3 10 Shama, Robert, 399 Steiner, C. R., 4 Steinman, D. B., 549 Sttissi, F., 549

555

556 Sulkiewicz, Jan, 541, 549 Tang, Mar&hung, 30,62, 81, 288, 319 Thul, H., 81, 319, 320, 439 Torroja, E., 400, 401, 405, 439 Troitsky, M. S., 439 Tyrrell, H. G., 30 Vatshell, J. L., 308-311 Virola, Juhani, 541, 549

Index of Personal Names Walther, Rpz, 549 Whatley, M. J., 353 Whitman, R. A., 81 Wittfoht, Hans, 43, 282, 319

Yang, Y. c., 440 Zellner, Willhelm, 377, 380, 439 Zublin, 3 7 7 , 3 8 0 , 399

Index of Firms and Organizations American Association of State Highway and Transportation Officials (AASHTO), formerly American Association of State Highway Officials (AASHO), 5, 6, 8, 30, 55, 148, 149, 190, 193, 199, 2.30, 243, 363, 460, 521, 524, 525, 526, 535, 538, 545 American Concrete Institute (ACI), 30, 36, 81, 97, 98, 147, 193, 199, 280, 295, 319, 324, 325, 326, 353, 404, 439, 466, 517, 539,549 American Consulting Engineers, Inc., 335 American Institute of Steel Construction (AISC), 439 American Railway Engineers Association (AREA), 148, 190 American Road and Transportation Builders Association (ARTBA), 535 American Society of Civil Engineers (ASCE), 30, 81, 352, 364, 365, 367, 399,409,410,439,440, 543,544, 549 Arvid Grant and Associates, Inc., 419 Associated General Contractors of America (AGC), 535 Associazione Italiana Cement0 Armato E Precompresso (AICAP), 353 Associazione Italiana Economica Del Cement0 (AITEC), 353

Florida Department of Transportation, 524, 525, 526 French National Railways, 453 Freyssinet International, 11,439, 440 General Services Administration (GSA), 519 G. F. Atkinson Company, 527 Heavy Construction News, 81 Hensel Phelps Construction Company, 304 Howard Needles Tammen & Bergendoff, 433, 434, 435, 440 Indiana State Highway Commission, 335 Institution of Civil Engineers, 340, 341, 342, 353, 439 International Association for Bridge and Structural Engineering (IABSE), 149,224,319,439 International Engineering Company, Inc., 524 International Road Federation, 30, 147, 224 Julius

Berger-Bauboag

Aktiengesellschaft,

403,

405-408

Kuwait Ministry of Public Works, 547 Beton-und-Stahlbetonbau, 35, 36, 81, 284, 320 Bouygues, 547 British Rail, 340 Bullen and Partners, 340 BVN/STS, 104,105, 330, 331 California Department of Transportation (CALTRANS), 10, 27, 47, 59, 61, 81, 523, 535 California Division of Highways, see California Department of Transportation Concrete Society (London), The, 341, 342, 343, 353 Concrete Technology Corporation, 538, 549 Der Bauingenieur, 297, 298, 299, 320 DRC Consultants, Inc., 62, 288 Du Pont de Nemours, 249 Dyckerhoff & Widmann, 13, 19, 30, 33, 35, 62, 63, 76, 81, 284, 285,296,301, 303,307,320,404,412,432, 549 Engineering News-Record, 37, 38, 39, 81, 353, 399, 439, 440 Entreprises Campenon Bernard, 12, 122, 419 Federal Highway Administration (FHWA), 30, 147, 224, 310, 419,439,440,520,535,538,545,549 Federation Internationale de la Precontrainte (FIP), 224, 517, 549 Federation Internationale de la Precontrainte, ComitC Europeen du Beton (FIP-cEB), 190, 193 Figg and Muller Engineers, Inc., 313, 524

Leonhardt and Andra, 24,323,419 L’Industria Italiana de1 Cemento, 403, 404 Main Roads Department, Australia, 20, 138, 139, 147, 399 Museum of Modern Art, New York, 30 Owens-Corning Fiberglass Corp., 537 Polensky-und-Zollner, 43 Portland Cement Association (PCA), 5, 8, 9, 10, 11, 12, 37, 49, 50, 56, 74, 81, 135, 136, 147, 309, 320, 353 Post-Tensioning Institute (PTI), 81, 320, 353, 517 Preston Corporation, 419 Prestressed Concrete Institute (PCI), 5, 6, 8-12, 30, 81, 110-112, 147, 224,320,466,517 Ralph Rodgers Construction Co., 335 Roads and Transportation Association of Canada, 81 Siemens-Bauunion, 296 Society of American Value Engineering, 5 19 Structural Engineers Association of California, 190 Texas Highway, 27, 109 Transportation Research Board, National Academy of Sciences, 30, 319,439,535 Travaux, 1 4 7 , 2 8 0 T. Y. Lin International, 436, 542

557

558

Index of Firms and Organizations

U.S. Army Corps of Engineers, 5 19 U.S. Bureau of Reclamation, 5 19 U.S. Coast Guard, 61 U.S. Department of Defense, 519 U.S. Department of Transportation, 5 19 U.S. Navy Bureau of Yards and Docks, 519 United States Steel Corporation, 540 University of Texas at Austin, Center for Highway Research, 109,147

VSL Corporation, 308, 335 Washington Department of Transportation, 307-311 Wayss & Fretag, 296 Weddle Bros. Construction Co., Inc., 335 White, Young and Partners, 401-404,430 Zement and Beton, 283, 294, 295,319

Index of Subiects J

AASHTO-PC1 I-girders, 5, 6, 8 Abutments, 8, 9, 24, 56, 132, 150, 205, 207, 232-235, 271-276, 333,336, 363 Aesthetics, 232-234, 354, 424 Alignment, 25 Alternate designs, 29, 59, 60, 63, 114, 285, 371, 520-522 Arches, 12, 15, 16, 31, 47, 354-382 frames, 12 Assembly truss, see Truss, falsework Balanced cantilever, see Cantilever construction Ballast, 62 Bars: Dywidag, 36, 52, 284, 330, 377, 380, 382, 404, 408, 411, 412,432,433,538 high-strength, 34 Macalloy, 340, 341, 342 Beam and winch, 20,91, 114-116, 128, 140, 510 Bearings, 25, 29, 40, 41, 43, 83, 125, 132, 138, 203-205, 460, 461 elastomeric, 68, 83, 96, 151, 158, 241-253, 418,426 lateral guide, 325, 338, 341 sliding, 326, 333, 414 temporary, 325, 338, 341 Bidding procedures, 5 18-523 alternate designs, 520-522 design and build, 519 single design, 5 18, 5 19 value engineering, 5 19, 520 Box girder, 10, 12, 14, 15, 21, 35, 37, 41, 43, 47, 48, 49, 58, 59, 68, 83, 88, 89, 103, 107, 109, 117, 138, 159, 160, 203, 322 cast-in-place on falsework, 28, 29, 103 efficiency, 159, 160 torsional moments, 169, 170, 205 transverse load distribution, 164, 169, 170, 202 transverse moments in deck slab, 170 variable depth, 37, 52, 203, 205 Box sections, see Box girder; Cross section Braking force, 331 Cable-stay, 12, 14, 16, 28, 400-440 advantages, 401403 structural style and arrangement, 403-405 temporary system, 333, 349, 350 Camber, 40, 63, 71, 97, 205-210, 482-485 Cantilever construction, 2-4, 11, 12, 17, 18, 23, 29, 31-146, 149 cast-in-place, 17, 18, 29, 31-80

J

cast-in-place operation sequence (cycle), 18, 19, 38, 43, 63 precast, 17, 18, 20, 29, 34, 82-147 Casting yard and factories, 20, 103, 135, 138, 504-507 Climbing forms, 45,49 Competitive bidding, see Alternate designs Concrete: composite, 541 creep, 40, 62, 71, 82, 83, 151, 155-158, 206, 231, 363, 365 curing, 114, 126, 466-469, 506 design and properties, 466 fiber-reinforced, 539, 540 high-strength, 537-539 lightweight, 58, 60, 72, 145 polymer, 540, 541 shrinkage, 17, 62, 71, 82, 83, 365 shrinkage damage, 324 Construction: critical-path, 82 joints, 21 push-out, see Incremental launching segmental, see Segmental construction sequence, see Erection sequence speed, 18 Continuity, 34, 49, 85, 87, 158, 206 Continuous superstructures, 155-158 Contractual aspects, 518-535 Contraflexure, 9, 21, 97, 113, 124, 150, 205 Cranes: barge (floating), 20, 22, 86, 87, 91, 109, 112, 138, 139, 237, 238, 239, 311, 312 crawler, 20, 509 floating, see Barge portal, 20, 103, 105, 509 swivel, 23 truck, 20, 128 Creep, see Concrete Cross section, 162-169, 202, 203, 217-219 Deflections, 17, 40, 157, 158, 159, 205-210, 483 Demolition, 352 Depth-to-span ratio, 35, 43, 52, 69, 89, 149, 322,402, 405 Design of segmental bridges, 148-223 Deviation blocks, 310 Diaphragms, 41, 53, 54, 57, 62, 67, 85, 90, 112, 124, 203, 204, 205, 207,332 Differential settlements, 276-280 Dimensional control, see Geometry control Double-T, 21, 322, 333

559

Zndex of Subjects Earthquake, 49, 59, 71, 363, 390,418 Ecology, 28,47, 304 Economics, 5 18-535 Environment, 18, 23, 27, 28, 29, 48, 99, 122, 304 Epoxy, 488-493 joints, 17, 83, 87,96, 103, 112, 124, 127, 150, 199-202, 414 Erection sequence, 66, 105, 127, 132, 133, 136, 341, 367, 397, 410,417 Expansion joint, 12, 29, 41, 57, 68, 83, 97, 99, 107, 113, 125, 155,158, 205, 208, 303, 331, 345,414,418 Expansion of long bridges, 158-159 External post-tensioning, 23 Falsework bents, 25 Fatigue, 210-212, 441 Finger joints, 138 Form traveler, 1, 17-21, 23, 36, 38, 40, 43, 50, 54, 58, 62-64, 70, 206, 302, 303, 475-478 Free cantilever, see Cantilever construction Friction, 321, 323, 330, 351 Foundations, 225-280 Geometry control, 321, 322, 495-497, 499-504 Hinges, 21, 34, 35, 49, 50, 62, 124, 126, 150, 205, 208, 363 Freyssinet concrete, 357, 365 mid-span, 151, 155-159, 354 I-girder, 8-10, 28, 29, 525, 526 Ice breaker, 124, 238, 239 Incremental launching, 12, 18, 24-27, 34, 84, 321-353 alignment requirements, 343 balanced casting, 322, 323 casting area, 349 construction principle, 343-346 continuous casting, 322 demolition by, 352 design, 343-352 launching methods, 349 piers and foundations, 350-352 span arrangement, 343-346 superstructure type, shape and dimensions, 343-345 Investment, 18, 29 Jacking, lift-and-push, 323 Jacks, Freyssinet flat, 360, 365, 374 Joints, see Construction; Expansion joint; Finger joints; Hinges; segments Keys, 11, 112, 124, 136, 199-202,409 Launching gantries, 20, 22, 64, 67, 96, 98, 103, 105, 106, 109, 117, 120-122, 124, 126-128, 134, 135, 141, 145, 511517 Launching girder, see Launching gantries Launching nose, 25, 46, 322, 325, 331-333, 338, 341, 346-350, 380 Launching sequence, see Erection sequence Live load requirements, 149 Long-bed, see Long-line method Longitudinal bending, 2 12, 2 13 Longitudinal closure strip, 112 Long-line method, 85, 97, 103, 142, 143, 493-494

Match-cast, 11, 17, 43, 82, 83, 85, 96, 97, 117, 124, 130, 199202.409.418 Match-cast joint, 199-202 Material quantities, 219-220, 223 Mayreder system, 377 Median frame, 9, 11 Model, 109, 112, 200, 201 Mushroom girder, 21, 296, 298 Natural frequency, 418 Negative moment tendons, see Cantilever construction; Prestressing tendons Neoprene bearings: double-row, 230, 241-253 influence of thickness and arrangement, 251-253 properties, 245-248 see also Bearings, elastomeric Piers, 39-41, 49, 63, 68, 89, 91, 99, 100, 113, 225-280, 301, 304, 308, 321,424 aesthetics, 232-234 caps, 49 deformation, 230, 231, 248-250, 257-261 double elastomeric bearing, 241-253 elastic stability with flexible legs, 261-263 flexibility, 83, 158, 229, 230, 331 foundation, 49, 68, 234-241 for incremental launching, 350-352 loads, 230-232 moment-resistant, 228, 230, 234-241 precast, 23, 228 properties with double-row neoprene bearings, 250, 251 properties with flexible legs, 257-261 segmental, 543, 544 stability during construction, 263-271 temporary, see Temporary bents temporary bracing, 105, 270, 27 1 temporary stays, 271 twin flexible legs, 228, 230, 253-263 Polyethylene pipe, 310 Positive-moment tendons, see Continuity; Prestressing tendons Precast outrigger struts, 45, 73 Precast piles, 135 Prestressed concrete: evolution, 4, 5 evolution of bridges, 5-8 precast girders, 8-10 Prestressing: continuity, 178, 179, 182 future provisions, 212 longitudinal, 69, 107, 109, 125, 213 temporary, 83, 103, 132-134, 334, 348, 414 three-dimensional, 36, 63, 69, 117 transverse, 17, 49, 60, 61, 69, 94 vertical, 17, 41, 69, 70, 125 Prestressing tendons: anchors, 470, 471 cantilever, 87, 109, 114, 173 continuity, 87, 109, 114, 149, 173 corrosion protection, 537 ducts, 470 friction losses, 47 l-475 glass fiber, 537

561

Index of Subjects grouting, 475 layout, 173, 174, 183, 184, 313, 471 straight, 175, 176, 177, 348 unbended, 475 Prestress losses, 61, 190, 206 Principal stresses, 197, 198, 201, 216 Problem areas, 220-222 Progressive placement, 12, 18, 22-25, 35, 37, 84, 281-293 design aspects, 314-319 Push-out construction, see Incremental launching Pylons, 16, 17, 408, 410, 411, 415, 418, 424, 426, 428 Railway, 26, 441-464 design aspects, 458-464 horizontal force, 459, 460 loading, 441, 458, 459, 463 Redistribution of moments and stresses through concrete creep, 183-189, 213-215 Reinforcement, compression, 36 Relaxation. 155, 231 Resal effect, 196, 263 Rigid frames, 12, 14, 15, 382-392 Segmental bridge decks, 542, 543 Segmental construction, 10-l 2, 17, 29, 30, 485-489 Segments: cast-in-place, 17, 18 dimensional tolerances, 469 fabrication, 82, 83, 85 formwork, 469,470 handling, 507-509 precast, 17, 18, 22, 23 size, 18, 83 weight, 18, 83 Shear, 193-199, 202, 203, 215-217 Short-line method, 90, 91, 109, 124, 126, 130, 494498 Shrinkage, see Concrete Slip-form, 43, 49, 99, 243, 298, 321, 424

Space frame bridges, 547, 548 Span arrangement, 149-l 5 1 Span-by-span construction, 12, 17, 18, 20-22, 34, 84, 282, 283, 293-314 Span-to-depth ratio, see Depth-to-span ratio Splice, 9-11 Standardized cross section, 130 Stationary forms, 24 Stray currents, 461 Stress ribbon bridges, 545, 546 Superelevation, 29, 103, 125, 302 Superstructure: cross section, see Cross section shape in elevation, 160, 161 Swing span, 428 T-beams, 21, 301 Teflon, 49,83,132, 323, 326,338,414,418,430 Temperature, 285, 302, 332, 365 Temperature gradient, 170-l 73, 199 Temporary bents, 323,333, 336,338,341, 345,396 Thermal movement, 62, 83, 231, 345, 390 Torsional moments in box girder, see Box girder Transverse load distribution, see Box girder Transverse moments in deck slab, see Box girder Transverse ribs, 67, 114 Traveler, see Form traveler Truss, 12, 14, 47, 392-399 auxiliary, 31, 43, 45, 50, 52, 125, 128 falsework, 22 Two-stage casting, 478, 479 Ultimate bending capacity, Uplift, 151, 204, 205

190-193

Value engineering, 143, 519-520 Wichert truss, 545 Width-to-depth ratio,

149

Podolny and Muller - Construction and Design of Pre Stressed Concrete Segmental Bridges

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