ACI STRUCTURAL JOURNAL
TECHNICAL PAPER
Title no. 106-S11
Introduction of Transition Transition Zone Design for Bridge Deck Link Slabs Using Ductile Concrete by Shunzhi Qian, Michael D. Lepech, Yun Yong Kim, and Victor C. Li This paper presents an innovative approach to designing the transition zones between concrete deck slab segments and an adjacent highly deformable link slab on a steel girder composite bridge deck. (A link slab provides a special class of jointless bridge for which which only only the bridge deck is is made continuous continuous rather rather than both the deck and girders). The transition zones represent a fraction of the ends of the link slab introduced to divert stress from the potentially weak link slab/deck slab interface. The link slab studied herein is built with an engineered cementitious composite (ECC), an ultra ductile concrete, adopted in a recent demonstration project in Southeast Michigan. Conventional design of concrete link slabs leaves the old/new concrete interface as the weakest part of the bridge deck system. Due to the presence of a link slab debond zone (part of the link slab is debonded from bridge girder to provide hinge flexib flexibili ility), ty), this this interfa interface ce also also experie experience ncess high str stress ess conce concentr ntrati ations ons.. In addi additio tion, n, the the effectiveness effectiveness of the link slab design design depends depends on the integrity of the interface so that imposed rotation and tensile deformation will be accommodated within the highly deformable ECC link link slab. slab. The basis of the suggested suggested approach approach is to isolate isolate the the concrete/ECC interface away from the structural interface between the debond zone and composite zone to prevent interfacial cracking. The shear studs and lap-spliced reinforcement located within transition zones facilitate load transfer between the concrete deck and the ECC link slab. These modifications are expected to cause a shift of the stress concentration from the concrete/ECC interface to the bulk part of the ECC link slab. In support of this new design concept, experimental tests are carried out on the behavior of the ECC link slab-bridge deck-girder connection. Detailing of the transition zone, that is, spacing of shear stud and development length/lap splice length requirement is then laid out in a design procedure based on results of shear stud/ ECC pushout pushout and reinfor reinforcement cement pullout pullout tests. tests.
(Kim et al. 2004). Within concrete link slabs, these two objectives directly contradict each other, simply due to the well-known dependence of controlled crack width in reinforced concrete upon high steel reinforcement ratios. The high reinforcement ratios in turn result in a large flexural stiffness, which is not desirable to achieve a hinge-like behavior for the link slab. Furthermore, saw-cut control joints filled with sealant, which are commonly used in concrete link slabs in Michigan and Texas to relieve stress due to combined mechanical and environmental loads (Caner and Zia 1998; Gilani and Jansson 2004), potentially pose problems similar to those in expansion joints due to the relative large saw-cut width (larger than 0.118 in. [3 mm]) and wearing out of the sealant through debonding, splitting, and damage from noncompressible debris (Purvis 2003). Overall, the brittle nature of normal concrete limits the durability of bridge deck link slabs.
Keywords: cracking; engineered cementitious composite; lap splice; stud.
INTRODUCTION Deterioration of bridge structures with mechanical expansion joints joints betwe between en simpl simply y support supported ed spans spans const constitut itutes es a signi signific ficant ant infrastructure deficiency, resulting in high maintenance and repair costs (Kim et al. 2004; Lepech 2006). Damage due to debris accumulation within the expansion joint leads to leakage of water through the joint or cracked concrete deck, followed by corrosion of girders and girder bearings, as shown in Fig. 1. One of the solutions for these problems is to use link slabs to replace expansion joints (Caner and Zia 1998). Link slabs provide a special class of jointless bridges for which only the bridge deck is made continuous rather than both the deck and girders. To realize a durable jointless bridge deck for improved resistance to aggressive agents, it is crucial that crack widths in the link slab remain tight (less than 0.004 in. [0.1 mm] according to NCHRP Report 380 [Krauss and Rogalla 1996]) while the flexural stiffness of the link slab is minimized to reduce its impact on the simple-span nature of the bridge
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Fig. 1—Corroded 1—Corroded bridge: (a) girder bearing; and (b) girder end reinforcement due to leakage of bridge joint. (Photo courtesy of R. Till, Michigan Department of Transportation.)
ACI Structural Structural Journal, Journal, V. 106, No. 1, January-February 2009. MS No. S-2007-248 received July 4, 2007, and reviewed under Institute publication publication policies. Copyright © 2009, American Concrete Institute. All rights reserved, including the making of copies unless permission is obtained from the copyright proprietors. Pertinent discussion including author’s closure, if any, will be published in the November-December 2009 ACI 2009 ACI Structural Structural Journal if Journal if the discussion is received by July 1, 2009.
ACI Structural Journal/January-February 2009
Shunzhi Qian is a Research Fellow in the Microlab, Faculty CiTG, at the Delft University of Technology, Delft, the Netherlands. He received his PhD from the University of Michigan, Ann Arbor, MI. His research interests include the application of high-performance fiber-reinforced cementitious composites to highway and other infrastructure systems for improved durability and sustainability. Michael D. Lepech is an Assistant Professor in the Department of Civil and Environmental Engineering at Stanford University, Stanford, CA. He received his BSE, MSE, and PhD from the University of Michigan. His research interests include the development and modeling of durable high-performance cementitious composites for use in the design of sustainable infrastructure systems. Yun Yong Kim is an Assistant Professor in the Department of Civil Engineering at Chungnam National University, Daejeon, South Korea. His research interests include the design of fiber-reinforced cementitious composites and their structural applications, including durable retrofit of aged infrastructures. Victor C. Li is a Professor in the Department of Civil and Environmental Engineering at the University of Michigan. His research interests include the design of ultra ductile and green cementitious composites and their application to innovative infrastructure systems and integration of materials, structural, and environmental design.
Whereas it is very difficult, if not impossible, for concrete to maintain tight crack widths with a minimal amount of reinforcement, engineered cementitious composites (ECC) represent a material solution that can simultaneously achieve tight crack width (typically less than 0.002 in. [0.06 mm]) and minimize steel reinforcement ratios in the link slab. These can only be achieved by ECC-type materials due to their self-controlled crack width along with high tensile strain capacity (tensile strain value corresponding to ultimate tensile strength obtained in a uniaxial tensile test [Fig. 2]). A typical tensile stress-strain curve of ECC under uniaxial tensile loading (material testing without steel reinforcement) is shown in Fig. 3, along with crack width development at different loading stages. As can be seen, the crack widths within ECC material are not a function of reinforcement ratio as in reinforced concrete, but are rather an inherent material property similar to compressive strength or elastic modulus. The tensile strain-hardening behavior, that is, increasing load capacity with increasing straining without localizing into a fracture plane, distinguishes ECC from normal concrete and fiber-reinforced concrete (FRC). Concrete and FRC show tension-softening after first cracking— that is, a decreasing load capacity with increased opening of a localized fracture. The tensile strain capacity of ECC is approximately 500 times that of normal concrete in this case. The significance of crack width in controlling the permeability coefficient of cracked concrete has been demonstrated by Wang et al. (1997). The permeability coefficient indicates the ability of concrete to bypass a fluid with specific viscosity under a pressure gradient. The permeability coefficient of cracked concrete was shown to decrease by seven orders of magnitude (from 10 –2 to 10–9 in./ second [10–4 to 10–11 m/second]) as crack width decreases from 0.02 to 0 in. (0.55 to 0 mm). When crac k width falls bel ow 0.004 in. (0.1 mm), the flow rate is similar to uncracked concrete. Hence even with a large number of surface cracks, ECC may behave like sound concrete with no cracks, by virtue of its tight crack width control (typically less than 0.002 in. [0.06 mm]). Similar results were also observed by Lepech and Li (2005). Given the ability to self-control tight crack width regardless of the amount of reinforcement, the use of ECC as a material for bridge deck link slabs has been proposed (Li et al. 2003; Kim et al. 2004; Kim and Li 2004). The research revealed that the property requirements for link slab applications were satisfied by the mechanical properties of ECC material.
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Compared with normal concrete, the self-controlled tight crack width of ECC material associated with the superior tensile strain capacity is expected to provide significantly enhanced durability to an ECC link slab, resulting in potential realization of highly durable concrete bridge deck systems. This is the motivation behind the adoption of ECC material in an actual implementation of an ECC link slab on a bridge deck in the fall of 2005 in Southeast Michigan. In a jointless bridge deck system, an ECC link slab can be used to accommodate bridge deck deformations due to temperature variation, live load, and shrinkage, replacing a typical expansion joint or concrete link slab. The interface between concrete and ECC, however, may become a weak link due to minimal mechanical interaction and load transfer across the cold joint between the existing concrete deck and newly cast ECC link slab, as the concrete and ECC are not cast at the same time. This phenomenon has been observed by the Michigan Department of Transportation (MDOT) (Gilani and Juntunen 2001; Gilani and Jansson 2004) in a preliminary laboratory investigation, where the design of an ECC link slab followed conventional concrete link slab design procedures. During their monotonic test to failure on a specimen simulating a jointless bridge deck using an ECC link slab, the interfacial crack between ECC and concrete grew noticeably (Fig. 4) while the width of cracks in the ECC link slab were maintained below 0.002 in. (0.05 mm). This indicates that a macroscopically crack-free bridge deck
Fig. 2—Uniaxial tensile test setup for ECC specimen.
Fig. 3—Typical tensile stress-strain curve and development of crack width of ECC.
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system may be difficult to realize due to potential large interfacial cracks, even though ECC material satisfies the property requirements of link slab design. The conventional design procedure for concrete link slabs does not address this potential interfacial cracking problem. In the design of conventional concrete link slabs (Caner and Zia 1998), the debond zone begins at the interface between the link slab and the concrete deck (Fig. 5(a)). This imposes high stress concentrations at this interface. Previous experiments suggest that the bond strength between concrete and ECC (hot joint: concrete and ECC are cast at the same time) is approximately 290 psi (2 MPa) (Zhang and Li 2002). As mentioned previously, the concrete/ECC interface is cold jointed when an ECC link slab is used to replace an expansion joint, considerably lowering the interface bond strength. Overall, this conventional design procedure makes the interface the weakest part of the bridge deck system. In the proposed new approach for designing the transition zone (Fig. 5(b)) at the ends of the ECC link slab, however, the location of the shear studs connecting the steel girder and the deck are extended into part of the ECC link slab, shifting the high stress concentration expected at the end of the debond zone away from the ECC/concrete interface. In addition to shear studs, the existing longitudinal reinforcement is lap
spliced with new reinforcing bars within the ECC link slab. This creates a transition zone between the debonded portion of the ECC link slab and the concrete deck or between the structural interface and the material interface (Fig. 5(b)). The shear studs and lap-spliced reinforcement located within the transition zone facilitate stress transfer between the concrete deck and the ECC link slab. These modifications are expected to prevent any failure of the ECC/concrete joint, but allow the free portion of the ECC slab to inelastically deform in response of tensile stretching, thus functioning as a jointless expansion joint. For constructability reasons, severe coldweather casting of ECC link slabs is not recommended (Li et al. 2005). Due to this, most thermal deformation imposed upon the ECC link slab will be tensile in nature. Therefore, compressive strains due to thermal deformation will not be of significance to the design of ECC link slabs. The purpose of this article is to present the experimental background and the design details of the transition zone between ECC link slabs and adjoining concrete deck slabs. First, full-scale testing of link slab under fatigue loading will be presented to demonstrate the feasibility and advantages of the proposed new design approach focusing particularly on the integrity of the link slab/concrete deck interface. Second, the design details of the transition zone, that is, the spacing of shear studs and development length/lap splice length of the top mat longitudinal reinforcement will be clarified based on the results of shear stud pushout tests and reinforcement pullout tests. Finally, the implementation of this new design approach into a recent (2005) ECC link slab demonstration project in Southeast Michigan is highlighted.
RESEARCH SIGNIFICANCE
Fig. 4—Significant interfacial cracking at ECC link slab/ concrete deck slab interface (Gilani and Jansson 2004).
While the conventional design procedure for concrete link slabs does not address the potential interfacial cracking problem, this paper presents a new approach of designing the transition zone between an ECC link slab and the adjacent concrete deck slab based on experimental tests of the behavior of the ECC link slab-bridge deck-girder connection. A detailed procedure for designing the transition zone is laid out for easy adoption by practicing engineers. This procedure has been adopted in an ECC link slab demonstration project in Southeast Michigan. The new transition zone design is expected to prevent interfacial cracking between the ECC link slab and concrete bridge deck, ensuring a successful implementation of the ECC link slab. This should result in a highly durable jointless bridge deck, leading to a significantly enhanced service life of the bridge.
FATIGUE TESTING OF FULL-SCALE LINK SLAB SPECIMENS Background of link slab full-scale testing
Fig. 5—Schematics of link slab design concepts for: (a) conventional method without considering transition zone; and (b) proposed method focusing on transition zone design.
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Three full-scale link slab specimens were tested cyclically to support the contention that interfacial cracking may be prevented by the newly introduced transition zone within link slab and therefore a macroscopically crack-free jointless bridge deck can be designed and constructed with ECC link slabs. While previous laboratory investigations of link slabs (Caner and Zia 1998; Gilani and Juntunen 2001) involved testing of a 1/6-scaled bridge including a link slab with two adjacent spans, the present study focused on testing of a fullscale link slab portion exclusively. According to Kim et al. (2004), the test conditions produce effective load severity five times greater than under field conditions in terms of the
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bending moment imposed by maximum allowable girder end rotations within the AASHTO design code (AASHTO 1998). The experimental investigation of ECC link slabs was conducted using a representative section (28 in. [711 mm] wide) of a link slab between the inflection points of the adjacent deck slabs (128 in. [3250 mm] long). The zero moment condition at the inflection points as well as the boundary conditions at the pier were simulated by roller supports at the specimen end supports and at the load points (Fig. 6). For practical purposes, the test setup represents an inverted orientation of the link slab region. The loading sequence chosen was similar to the procedure adopted by MDOT (Gilani and Juntunen 2001), as shown in Fig. 6. All specimens were subjected to sequential static loading up to two times the deflection, causing a reinforcing bar stress in Specimen LS-1 of 40% of its yield strength, which is the current limit stress criterion for concrete link slab design. The final step of the sequential static loading stage simulates potential overload. In the subsequent cyclic loading procedure, the load at 40% yield of the reinforcement in Specimen LS-1 is chosen as the mean load with an amplitude up to a maximum deflection at 0.00375 rad rotation angle (Fig. 6). This maximum rotation angle (0.00375 rad) corresponds to the allowable deflection of a bridge span under live load according to Section 6.10.5 of the AASHTO code (AASHTO 1998). Cyclic loading was carried out to 100,000 cycles due to restrictions in the availability of the testing equipment. Figure 7 shows the specimen geometry of three link slabs, including the debond zone length (50 in. [1.27 m]) equal to roughly 2.5% of assumed adjacent spans (both 82 ft [25 m] long). The thickness of the link slab was 9 in. (230 mm), corresponding to a typical deck thickness in simply supported composite girder bridges. Specimens LS-1, LS-2,
Fig. 6—Full-scale link slab testing: (a) loading pattern; and (b) test setup.
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and LS-3 were used to simulate the concrete link slab new construction, ECC link slab new construction, and a retrofit of an existing bridge deck replacing mechanical expansion joints with an ECC link slab, respectively. It should be noted that for Specimens LS-2 and LS-3, the proposed transition zone design approach was implemented, meaning that additional shear studs were extended within the transition zone of the link slab and additional reinforcement in the link slab was lap spliced with existing reinforcement in the bridge deck for Specimen LS-3. This is different from the link slab tests conducted by MDOT (Gilani and Jansson 2004), in which the test specimens were designed based on the conventional link slab design approach, that is, the ECC/ concrete interface was adjacent to the debond zone, leaving the interface the weakest component in the bridge deck system.
Results and discussion For success of the link slab application, the integrity of the link slab/deck interface and the overall crack pattern (particularly the crack width) in the bulk part of the ECC link slab are critical. For details of the specimen preparation and experimental program, refer to Li et al. (2003) and Kim et al. (2004). It should be noted that because the incorporation of an ECC link slab in a bridge deck is for durability reasons rather than load capacity, the formation of cracks, if any; their width at the interface; and the crack pattern in the bulk
Fig. 7—Geometry of link specimens for: (a) Specimen LS-1; (b) Specimen LS-2; (c) Specimen LS-3; and (d) cross section of link slab specimens. Units in in. (mm).
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part of link slab, including crack width and distribution are among the most important characteristics considered. For Specimens LS-2 and LS-3, the test results revealed no cracking at the interface between the ECC link slab and concrete deck slab during cyclic testing, whereas the intended microcracks were mostly confined within the debond span of the link slab up to the location of the shear studs (Fig. 8). This indicates that cracks would have appeared at the concrete/ECC interface had it been located at the end of the debond zone. The modification of the design to locate the concrete/ECC material interface away from the structural interface (Fig. 5(b)) between the debond zone and girder/deck composite zone prevented cracking at the material interface. Furthermore, the additional shear studs placed between these two interfaces provided composite action between the girder and ECC slab, effectively holding the material interface together to prevent cracking. As a result, concrete/ECC interfacial cracking caused by stress concentrations is prevented. Cracking is limited within the debond segment of the ECC link slab, where sufficient strain capacity exists to accommodate large deformation demands. Therefore, this new approach of designing the transition zone in link slabs will provide enhanced integrity of the concrete/ECC interface, preventing undesirable material interfacial failure. Although structural failure did not occur in all the specimens during fatigue testing, the cracking patterns of ECC link slab specimens were much more desirable compared with that of the concrete link slab specimen, as shown in Fig. 9. For Specimen LS-1, a small number of large cracks were observed. During fatigue cycles, no additional cracks were seen to form and the existing cracks generated during the preloading stage gradually grew wider. The crack widths in concrete ultimately reached approximately 0.025 in. (0.64 mm)
Fig. 8—Crack pattern at tension surface of link slab in: (a) Specimen LS-2; and (b) Specimen LS-3.
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at 100,000 load cycles. In contrast, a large number of hairline cracks were observed in ECC link slab Specimens LS-2 and LS-3 during the preloading stage. Additional microcracks appeared as the number of loading cycles increased, whereas the existing crack widths were maintained below 0.002 in. (0.05 mm), slightly opening and closing at the maximum and minimum loads, up to 100,000 cycles. Such tight crack width results in low permeability (Lepech and Li 2005) and diffusion coefficients (Sahmaran et al. 2006), and enhance the durability of an ECC link slab particularly under severe environmental conditions, such as in regions where deicing salts are frequently used. This distinction of ECC link slabs from concrete link slabs can only be achieved with the assured integrity of the concrete deck/ECC link slab interface, thus allowing the ECC link slab to function effectively as an expansion joint.
REVIEW OF SHEAR STUD PUSHOUT AND REINFORCEMENT PULLOUT TEST Having investigated the behavior of the ECC link slabbridge deck-girder connection and demonstrated the feasibility and advantages of the proposed new design
Fig. 9—Crack pattern marked with ink pen for clarity after cyclic test for: (a) Specimen LS-1; (b) Specimen LS-2; and (c) Specimen LS-3.
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approach for transition zones in full-scale link slab tests, more detailed experimental investigations have been carried out on the local behavior of the load transfer mechanisms in the transition zone, that is, shear stud pushout tests (Fig. 10) and reinforcement pullout tests (Fig. 11). Detailed experimental programs for both tests can be found in Li et al. (2003) and Qian and Li (2006). Only results from these tests relevant for clarifying the design details of the transition zone are reported herein. Specifically, the spacing of shear studs and development length of the top mat longitudinal reinforcement required for load transfer across the transition zone are analyzed in the following. The mixture proportions and results for shear stud pushout tests are summarized in Table 1, where three different ECC mixtures (ECC 1, ECC 2, and ECC 3) were tested along with one concrete mixture as a comparison. As can be seen, stud anchors in ECC (ECC 1) can achieve much higher shear strength and slip capacity (structural ductility) when compared with concrete of comparable compressive strength due to the extreme ductility of ECC, which suppresses the
Fig. 10—Geometry of: (a) shear stud; and (b) pushout specimen. Units in in. (mm).
brittle fracture mode typical of concrete response. It should be noted that ECC 3 specimens have two rows of shear studs, a total of four studs inside both slabs, whereas all other series (ECC 1, ECC 2, and concrete) have only one row of two shear studs on both slabs, as shown in Fig. 10. The ECC 3 specimens were conducted specifically to investigate the effect of the spacing (in the longitudinal direction of the bridge deck) on the structural response of shear stud anchor connections. The spacing chosen was 4 in. (102 mm), which is slightly smaller than the lower limit suggested by the AASHTO code (AASHTO 1998) (six times stud diameter, 4.5 in. [114 mm] in this case) for conservativeness. The diameter of the stud connector chosen was 0.75 in. (19 mm), which is common in composite bridge construction (An and Cederwall 1996; Badie et al. 2002). The shear strength test results summarized in Table 1 provide the basis for the calculation of the number and spacing of studs in ECC for transition zone detailing purposes. As shown in Fig. 11, reinforcing bar pullout tests were carried out to obtain the appropriate development length of No. 3 reinforcing bars embedded within ECC material. According to the AASHTO code (1998), minimum steel reinforcement ratios for resisting shrinkage and temperature stresses are required to provide at least 0.125 in. 2 /ft (265 mm2 /m) in each direction. To provide this amount of reinforcement, No. 3 reinforcing bars with a spacing of 10.5 in. (267 mm) were placed longitudinally in the deck. This minimum reinforcement
Fig. 11—Reinforcing bar pullout test setup.
Table 1—Mixture proportions, material properties, and structural response of concrete and ECC materials pushout specimens No. of specimens Mixture tested
Type I portland cement
Water
Silica sand*
High-range Coarse Type F fly water-reducing † aggregate ash admixture
PVA fiber
εu,
‡
%
0.01
f c′, ksi (MPa)
Qm,§ kips (kN)
S c,|| in. (mm)
5.5 ± 0.2 29.1 ± 0.4 0.067 ± 0.004 (38.0 ± 1.4) (129.6 ± 1.9) (1.7 ± 0.1)
Concrete
3
1
0.45
2
2
0
0
0
ECC 1
2
1
0.58
0.8
0
1.2
0.03
0.02
2.5
±
0.4
6.7 ± 0.1 36.2 ± 4.0 0.228 ± 0.012 (46.0 ± 0.4) (160.9 ± 17.7) (5.8 ± 0.3)
ECC 2
3
1
0.53
0.8
0
1.2
0.03
0.02
2.5
±
0.3
8.7 ± 0.3 43.2 ± 2.6 0.252 ± 0.051 (60.0 ± 2.1) (192.3 ± 11.7) (6.4 ± 1.3)
ECC 3
3
1
0.53
0.8
0
1.2
0.014
0.02
2.2
±
0.3
9.0 ± 0.3 34.2 ± 3.4 0.197 ± 0.031 (62.4 ± 2.0) (152.1 ± 15.1) (5.0 ± 0.8)
*
F110 sand for ECC1, ECC 2, and ECC 3, and ASTM C778 sand for concrete. Maximum size of 0.75 in. (19 mm). ‡ Uniaxial tensile strain capacity (assumed value for concrete). § Measured shear strength per stud. || Slip capacity (average slip at peak load). †
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ratio for the top mat of longitudinal reinforcement is also employed in standard MDOT bridge slabs (MDOT 2001). Considering the retrofit of an existing bridge using an ECC link slab, No. 3 reinforcing bars were expected to be lap spliced with new reinforcing bars of the ECC link slab. From the pullout test results (Li et al. 2003), a 6 in. (152 mm) embedment length is found to be adequate to develop the yield strength of both plain and epoxy-coated reinforcing bars in ECC, meaning that the development length of these reinforcements in ECC is 6 in. (152 mm). The AASHTO code (1998), however, requires a 12 in. (305 mm) minimum development length for No. 3 reinforcing bars in concrete. Hence, if the AASHTO code (1998) requirements of development length were directly adopted for ECC, it will be at least as conservative as that of concrete due to high ductility of ECC, which can totally suppress the bond splitting failure as sometimes seen in reinforcing bar pullout test from concrete (Li et al. 2003).
TRANSITION ZONE DESIGN Shear stud detailing The deformation (strain) demand on the ECC link slab due to shrinkage, thermal effects, and live load must be accommodated within the debond zone of the link slab. Therefore, corresponding forces (stress) developed in debond zone must be transferred through the transition zone back into the composite girder of the bridge by both the shear studs and lap spliced reinforcements, in addition to the concrete/ECC interfacial bond. For conservativeness, however, the contribution from the reinforcement and interfacial bond will be ignored for shear stud detailing purposes. Based on the shear strength of studs in ECC (capacity side) from shear stud pushout tests, the maximum strain demand of ECC (Li et al. 2003, 2005) and corresponding stresses (forces, demand side), a simple design procedure for transition zone detailing in terms of spacing of shear studs is developed. Demand side— The maximum tensile strain demand of ECC (εls) is determined from the imposed strain due to the combined effect of shrinkage, thermal loads, and live loads, as shown in following equation
εls
α T ⋅ ∆ T ⋅ β ⋅ Llong = -----------------------------------------+ ε sh + ε LL L dz
(1)
where εls is the required tensile strain of ECC due to combined effects of temperature, shrinkage, and live loads; αT is the coefficient of thermal expansion of concrete and steel (0.0000065 in./in./°F [0.0000117 m/m/°C]); ∆T annual
Fig. 12—Typical tensile stress-strain behavior of ECC designed for link slab application.
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temperature variation at bridge location (approximately 90 °F [50 °C] for Michigan according to the AASHTO code [1998]); β is the support type factor, β = 2.0 for joints with two roller bearings (one for each adjacent span), and β = 1.0 for joints with one roller bearing and one pin bearing for adjacent spans or joints with two pin bearings; Llong is the span length of longer adjacent bridge span (larger one of L1, L2), inches; Ldz is the length of link slab debond zone (= 5% ( L1 + L2) + GAP), inches; L1, L2 is the length of adjacent bridge spans, inches; GAP is the gap between adjacent bridge girders, inches; εsh is the shrinkage strain of ECC (= 0.001 from Weimann and Li [2003]); ε LL is the maximum tensile strain due to live load (= 0.0008( c + d )/ d , more detailed derivation can be seen in Li et al. [2005]); c is the distance from top of slab to centroid of tensile reinforcement, inches; and d is the distance from neutral axis to centroid of tensile reinforcement, inches. Force demand F d—The corresponding stress σls of the derived maximum tensile strain demand is determined from the representative tensile stress-strain relation of the ECC material used in the link slab. As shown in Fig. 12, a corresponding stress (800 psi [5.5 MPa]) can be found for a given tensile strain (assumed value of 2%). In addition, the link slab width (same as bridge deck width ws) and thickness t s are needed to calculate the total force demand F d developed in the link slab, which must be transferred through the transition zone and is the product of σls, ws, and t s. That is F d = σlswst s
(2)
where F d is the force demand in the link slab that needs to be transferred by shear studs; σls is the corresponding stress of the derived maximum tensile strain demand of ECC material used in the link slab; ws is the width of the link slab (same as the bridge deck width); and t s is the thickness of the link slab Capacity side—As shown in Table 1, the shear strength of individual studs in the ECCs ( Qm) has been experimentally investigated by Qian and Li (2006) along with a new result of pushout test with two rows totaling four shear studs in each slab. According to the ACI Building Code (ACI Committee 318 2005) for anchorage in concrete (the AASHTO code [AASHTO 1998] has no corresponding provision), the required reduction in shear strength of grouped anchor studs may be ignored when the spacing between shear studs is greater than three times the embedment length of the stud. This condition may apply to ECC very conservatively given that load redistribution among different shear studs (in the longitudinal direction of the bridge) may occur due to superior structural ductility of the studs within ECC. This means that the shear strength of ECC 1 and ECC 2 (36.2 and 43.2 kips [160.9 and 192.3 kN], respectively) may be used as the lower bound of the actual structural response when spacing between shear studs is more than three times the embedment length (3 in. [76 mm] in the case of Fig. 10). For closer spacing of shear studs (less than three times the embedment length), the result of ECC 3 (34.2 kips [152.1 kN]) may be used conservatively for design purposes. Force capacity F c—The number of shear studs per row on a steel girder flange may be assumed to be two or three, depending on the width of the girder flange as allowed by the AASHTO code (AASHTO 1998). Given the shear strength of an individual shear stud in ECC ( Qm), the number of the girders ( N g), and the number of shear studs per row ( N r ), the total shear strength for one row of shear studs on all girders
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( Rs = N g N r Qm) can then be calculated. The product of Rs and the required number of rows of shear studs ( N ) will yield the total shear capacity of the transition zone provided by the studs. That is F c = Rs N
(3)
Number and spacing of shear stud rows within transition zone— By equating the demand side (Eq. (2)) and capacity side (Eq. (3)) of the transition zone, the number of rows of shear studs needed ( N ) can be derived as follows N = F d / Rs = σlswst s / N g N r Qm
(4)
According to Li et al. (2003, 2005), the length of the ECC link slab is the sum of 7.5% of each adjacent girder span ( L). Within this length, 5% of each girder span should be debonded from the ECC link slab to reduce its stiffness so that the stress developed in the link slab can be minimized. The 5% debonding length was based on results from Caner and Zia (1998) that showed that the simply supported configuration of the multiple span bridges would not be affected by a debonding length of up to 5% of the adjacent girder span. Additionally, it was found necessary to extend the length of the ECC link slab 2.5% further into each adjacent span (length of transition zone) to help transfer load from the girders into the ECC link slab through additional shear studs. Knowing the length of transition zone, the spacing of shear studs in transition zone ( S st ) can then be determined by the following equation S st = 2.5% L / N
Conversely, as long as both adjacent spans are longer than 68.5 ft (20.9 m) and a typical thickness of 9 in. (229 mm) deck slab is used, the requirement for additional reinforcement bars in the ECC link slab can be satisfied with No. 3 reinforcing bar at a spacing of 10.5 in. (267 mm) (detailed derivation is given in the Appendix). According to Gilani and Jansson (2004) and Li et al. (2005), six out of seven link slabs already built by MDOT (including recently built ECC link slabs) have adjacent span lengths longer than 68.5 ft (20.9 m), suggesting No. 3 reinforcing bar may be used at a spacing of 10.5 in. (267 mm) in most cases if ECC were to be used in these link slabs. Only the determination of the top mat of the longitudinal reinforcing bar is covered within this paper. Other reinforcement, such as transverse reinforcement in both top and bottom mats, along with any minimum reinforcement within the bottom mat is not addressed. These reinforcements should be designed following AASHTO design procedures. Similarly, all reinforcement detailing for sidewalks, barrier walls, or other bridge features should be completed following AASHTO design procedures.
FIELD IMPLEMENTATION The proposed new approach of designing the transition zone in link slabs has been adopted by MDOT in a recent ECC link slab demonstration project in Southeast Michigan. The project site is Grove Street over Interstate 94 in Ypsilanti,
(5)
Reinforcement detailing As mentioned previously, No. 3 reinforcement has been tested to investigate its development length requirement in ECC. While overly conservative for ECC, the AASHTO code (AASHTO 1998) will be adopted for calculation of the development length and lap splice length of existing No. 3 reinforcing bar in ECC for easy adoption by DOT designers. It should be noted that the existing top mat reinforcement should be run into the transition zone far enough to allow for a lap splice with the additional reinforcement in the ECC link slab. The detailed calculation for development length and lap splice length can be found in Li et al. (2005) based on the AASHTO code (AASHTO 1998).
Fig. 13—Finished ECC link slab.
ACI Structural Journal/January-February 2009
Fig. 14—Intact interface was observed on ECC link slab after 1 year in service.
Fig. 15—Details of transition zone area in link slab demo nstra tion project. Units in in. (mm).
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approximately 45 minutes outside of Detroit. Through the cooperation of state transportation officials, industry consulting engineers and contractors, and research personnel at the University of Michigan, this first infrastructure demonstration project of ECC material in the U.S. (Fig. 13) was completed in mid-November 2005. After opening to traffic for 1 year, the ECC link slab revealed excellent performance. No interfacial crack was observed, as can be seen in the shoulder region in Fig. 14. The portion of interface in the traffic lanes cannot be directly observed due to very high traffic volume. Nevertheless, it is expected that the difference, if any, should be very minor between these two portions of the interface. This short-term observation indicates the promise of the proposed new transition zone design approach. The overall design procedure for an ECC link slab can be found in Li et al. (2003, 2005). The actual reinforcement layout of the transition zone area in the ECC link slab is shown in Fig. 15, where the steel girder, shear studs, and reinforcement details can be clearly seen. For this application, 30# roofing paper was used to form the debond zone. According to the AASHTO code (AASHTO 1998), overall requirements for center-to-center spacing (pitch) of shear studs shall not exceed 23.6 in. (600 mm) and shall not be less than six stud diameters (4.5 in. [114 mm] in this case). Specifically, the longitudinal spacing of the shear studs in the ECC link slab transition zones was found to be approximately 7 in. (178 mm), with three studs in the transverse direction for each steel girder using the procedure described in the previous section. As for the reinforcement detailing, it was found that No. 3 reinforcing bar at a spacing of 10.5 in. (267 mm) can satisfy the requirements of this link slab application. Because the ECC link slab was incorporated within new bridge deck construction, the existing reinforcement was run continuously into the ECC link slab. Therefore, the consideration on development length and/or lap splice length was not necessary in this application. If these reinforcements must be terminated at the transition zone, however, they can still be designed the same way as in the case of a retrofit, which requires a lap splice length of 14 in. (356 mm) and should be staggered a minimum of 6 in. (152 mm), as recommended by AASHTO code (AASHTO 1998).
stress concentration shifting, no cracking is observed at the bridge deck/link slab interface under fatigue loading, whereas desirable microcracking formed over the debond zone of the link slab up to the location of shear studs; 2. Based on the experimental studies on shear stud pushout tests, it was concluded that stud anchors in ECC can achieve much higher shear strength and slip capacity (structural ductility) when compared with concrete of comparable compressive strength due to the superior ductility of ECC. From the reinforcing bar pullout tests, a 6 in. (152 mm) embedment length was found to be adequate to develop the yield strength of both plain and epoxy-coated reinforcing bars in ECC, meaning that the development length of these reinforcements in ECC is 6 in. (152 mm). However, the AASHTO code (AASHTO 1998) requires a 12 in. (305 mm) minimum development length for No. 3 reinforcing bars in concrete. Hence, if AASHTO code (AASHTO 1998) requirements of development length were directly adopted for ECC, it will be at least as conservative as that of concrete due to high ductility of ECC that can totally suppress the bond splitting failure as sometimes seen in reinforcing bar pullout test from concrete; 3. Experimental investigations demonstrated the feasibility and advantages of the proposed transition zone design in the ECC link slab. Overall, the proposed transition zone design approach combined with advanced ECC material makes ECC link slab a feasible solution for bridge deterioration problems associated with mechanical expansion joints, and potentially much more durable compared with concrete link slabs designed using conventional methods; and 4. The proposed transition zone design approach has been materialized through a detailed design procedure based on experimental results of shear stud pushout tests and reinforcement pullout tests. Further, this new design method has been implemented into a current ECC link slab demonstration project in Southeast Michigan, validating it as a useful and reasonable design tool for practicing bridge engineers.
ACKNOWLEDGMENTS This work has been supported by a research grant from the Michigan Department of Transportation to the University of Michigan with project managers D. Juntunen (2001) and R. Till (2002-2006). This support is gratefully acknowledged. G. Fischer, S. Wang, M. Weimann, M. Li, and E. H. Yang contributed to this project.
CONCLUSIONS To achieve the potential realization of a durable concrete bridge deck system by incorporation of an ECC link slab, an innovative approach of designing the transition zone in a link slab has been proposed. This design approach was also implemented into an ECC link slab demonstration project in Southeast Michigan. Intact interface after 1 year of service suggests the promise of the new transition zone design approach. The basis of the suggested approach is to locate the concrete/ECC interface away from the structural interface between the debond zone and girder/deck composite zone, preventing interfacial cracking. The shear studs and lap spliced reinforcement located within the ECC link slab transition zone further assist in load transfer between the concrete deck and the ECC link slab. The following conclusions can be drawn from this investigation: 1. The results from the full-scale link slab experiment suggest that the addition of shear studs in the transition zone is very effective in shifting the stress concentration from the interface to the bulk part of the link slab. As a result of this
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REFERENCES AASHTO, 1998, “AASHTO LRFD Bridge Design Specification,” second edition, American Association of State Highway and Transportation Officials, Washington, DC. ACI Committee 318, 2005, “Building Code Requirements for Structural Concrete (ACI 318-05) and Commentary (318R-05),” American Concrete Institute, Farmington Hills, MI, 430 pp. An, L., and Cederwall, K., 1996, “Push-Out Tests on Studs in High Strength and Normal Strength Concrete,” Journal of Construction Steel Research, V. 36, No. 1, pp. 15-29. Badie, S. S.; Tadros, M. K.; Kakish, H. F.; Splittgerber, D. L.; and Baishya, M. C., 2002, “Large Shear Studs for Composite Action in Steel Bridge Girders,” Journal of Bridge Engineering, May-June, pp. 195-203. Caner, A., and Zia, P., 1998, “Behavior and Design of Link Slab for Jointless Bridge Decks,” PCI Journal , May-June, pp. 68-80. Gilani, A., and Juntunen, D., 2001, “Link Slabs for Simply Supported Bridges: Incorporating Engineered Cementitious Composites,” Report No. MDOT SPR-54181, Michigan Department of Transportation, July, 88 pp. Gilani, A., and Jansson, P., 2004, “Link Slabs for Simply Supported Bridges,” Repor t No. MDOT SPR-54181, Michigan Department of Transportation, June, 129 pp. Kim, Y. Y.; Fischer, G.; and Li, V. C., 2004, “Performance of Bridge Deck Link Slabs Designed with Ductile Engineered Cementitious Composite (ECC),” ACI Structural Journal, V.101, No. 6, Nov.-Dec., pp. 792-801.
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Kim, Y. Y., and Li, V. C., 2004, “Fatigue Response of Bridge Deck Link Slab Designed with Ductile Engineered Cementitious Composite (ECC),” Proceedings of the International Conference on Concrete under Severe Conditions—Environment & Loading (CONSEC’04), Seoul, Korea, June, pp. 832-841. Krauss, P. D., and Rogalla, E. A., 1996, “Transverse Cracking in Newly Constructed Bridge Decks,” Report 380, National Cooperative Highway Research Program (NCHRP), Transportation Research Board. Lepech, M. D., 2006, “A Paradigm for Integrated Structures And Materials Design for Sustainable Transportation Infrastructure,” PhD thesis, University of Michigan, Ann Arbor, MI. Lepech, M. D., and Li, V. C., 2005, “Water Permeability of Cracked Cementitious Composites,” Paper 4539 of Compendium of Papers, ICF 11, Turin, Italy. (CD ROM) Li, V. C., 1993, “From Micromechanics to Structural Engineering—the Design of Cementitious Composites for Civil Engineering Applications,” JSCE Journal of Structural Engineering and Earthquake Engineering, V. 10, No. 2, pp. 37-48. Li, V. C., 2002, “Reflections on the Research and Development of Engineered Cementitious Composites (ECC),” Proceedings of the JCI International Workshop on Ductile Fiber Reinforced Cementitious Composites (DFRCC)—Application and Evaluation (DRFCC-2002) , Takayama, Japan, Oct., pp. 1-21. Li, V. C.; Fischer, G.; Kim,Y. Y.; Lepech, M.; Qian, S.; Weimann, M.; and Wang, S., 2003, “Final Report on Durable Link Slabs for Jointless Bridge Decks Based on Strain-Hardening Cementitious Composites,” Michigan Department of Transportation, Oct., 96 pp. Li, V. C.; Lepech, M.; Li, M.; Lynch, J.; and Hou, T., 2005, “Field Demonstration of Durable Link Slabs for Jointless Bridge Decks Based on Strain-Hardening Cementitious Composites,” Final Report submitted to Michigan Department of Transportation, Dec., 147 pp. MDOT, 2001, “Standard Bridge Slabs (Load Factor Design and Empirical Design),” Michigan Department of Transportation, Bureau of Highway Technical Services, Issued on Nov. 27, 2001. Purvis, R., 2003, “Bridge Deck Joint Performance,” NCHRP Synthesis 319, Transportation Research Board, 58 pp. Qian, S., and Li, V. C., 2006, “Influence of Concrete Material Ductility on Shear Response of Stud Connections,” ACI Material Journal, V. 103, No. 1, Jan.-Feb., pp. 60-66. Sahmaran, M.; Li, M.; and Li, V. C., 2007, “Transport Properties of Engineered Cementitious Composites under Chloride Exposure,” ACI Materials Journal, V. 104, No. 6, Nov.-Dec., pp. 604-611. Wang, K.; Jansen, D. C.; and Shah, S. P., 1997, “Permeability Study of Cracked Concrete,” Cement and Concrete Research, V. 27, No. 3, pp. 381-393. Weimann, M. B., and Li, V. C., 2003, “Hygral Behavior of Engineered Cementitious Composites (ECC),” International Journal for Restoration of Buildings and Monuments , V. 9, No. 5, pp. 513-534. Zhang, J., and Li, V. C., 2002, “Effect of Inclination Angle on Fiber Rupture Load in Fiber Reinforced Cementitious Composites,” Composite Science and Technology, V. 62, No. 6, pp. 775-781.
(267 mm) as long as both adjacent spans are longer than 68.5 ft (20.9 m), assuming a typical 9 in. (229 mm) thick deck slab is present. The detailed derivation of this conclusion is presented herein for the reader’s interest, which is based on the design procedure of ECC link slabs in Li et al. (2003, 2005). 1. The AASHTO minimum requirement of reinforcement ratio is found to be ρ = 0.11/(10.5 × 9) = 0.001164 (the crosssectional area for a No. 3 reinforcing bar is 0.11 in. 2 [71 mm 2]); 2. Length of link slab debond zone
APPENDIX I—DERIVATION OF LOWER LIMIT OF SPAN LENGTH TO USE AASHTO MINIMUM REINFORCEMENT RATIO
7. Check if moment resistance is larger than moment induced by the maximum girder end rotation
While the transition zone detailing procedure was shown previously mainly for the design of shear stud spacing and lap splice length under combined bending and tensile deformation, this appendix is for the design of additional reinforcement in the ECC link slab (debond zone), where the major load case considered is bending as long as the tensile strain capacity of ECC is not exceeded. The requirement for additional reinforcement in the ECC link slab can be satisfied with AASHTO minimum reinforcement requirements for bridge decks, that is, No. 3 reinforcing bar at a spacing of 10.5 in.
M ls ≤ M R
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Ldz = 0.05 ( L1 + L2) + GAP ≈ 0.05 (L1 + L2) = 0.1 L (A-1)
(assuming the length of both adjacent bridge spans is equal to L); 3. Maximum beam end rotation θmax is equal to 0.00375 rad., which is constant for all span lengths. More details can be found in Li et al. (2003, 2005); 4. Moment of inertia of link slab (per foot width of bridge deck) 3
3 12 × t s 12 × ( 9 ) I ls = ---------------- = ---------------------- = 729 in. 4 (0.0003 m4) (A-2) 12 12
5. Moment in link slab due to max girder end rotation: 2 E EC C I ls ( 2900 ) ( 729 ) M ls = ----------------------------------------------- 0.00375 θma x = 2------ L dx 0.1 L 158,557.5 = ------------------------ kip-in./ft L
(A-3)
58,772 ---------- kN-m/m ------ L
6. For a 9 in. (229 mm) thick link slab with AASHTO minimum reinforcement ratio, moment resistance (per foot width) of the link slab can be calculated by the following equation (Li et al. 2005) M R = 10,211ρ + 180.9 = 10,211 × 0.001164 + 180.9 = 192.78 kip-in./ft (71.46 kN-m/m) (A-4)
L ≤ 192.78 158,557.5/ L ≥ 822.5 in. = 68.5 ft (20.9 m)
Therefore, the lower limit of span length is 68.5 ft (20.9 m) to use AASHTO minimum reinforcement ratio, that is, No. 3 reinforcement at spacing of 10.5 in. (267 mm).
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