International Soil Mechanics and Geotechnic Conference Proceedings

Technical Committee 102

Ground Property Characterisation from In-Situ Tests Comité technique 102

Caractérisation des propriétés des terrains par essais in situ

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

General Report for TC102 In-Situ Testing Rapport général du TC102 Essais in-situ Giacheti H.L.

São Paulo State University, Dept. of Civil and Environmental Engineering, Bauru-SP, Brazil (wwwp.feb.unesp.br/giacheti)

Cunha R.P.

University of Brasília, Dept. of Civil and Environmental Engineering, Brasília-DF, Brazil (www.geotecnia.unb.br/gpfees)

ABSTRACT: An overview of all the technical papers accepted for the in-situ testing session of the 18th ICSMGE is presented. Forty two papers submitted to this conference were considered as part of this session. The papers were grouped into four major categories: site characterization, technological advances, geotechnical analyses and behavior, and soil and rocks properties. The objective of this report is to present an overview of the theme topics and briefly discuss the major contributions achieved by these papers. RÉSUMÉ : Un aperçu de tous les articles acceptés à la conférence dans la session «Essais in-situ» du 18e CIMSG est présenté ici. Quarante-deux écrits soumis à cette conférence ont été considérés comme faisant partie de cette session. Les documents ont été regroupés en quatre grandes catégories: la caractérisation du site, les progrès technologiques, les analyses géotechniques, les comportements et enfin les propriétés des sols et des roches. L'objectif de ce rapport est de présenter une vue d'ensemble de tous les sujets et de discuter brièvement des contributions majeures apportées par ces documents. KEYWORDS: In-situ tests, site characterization, technological advances, geotechnical analysis and behavior, soil and rock properties. 1

Figure 2.a shows the distribution of all papers in this session that used any particular in-situ testing technique as a major site investigation tool. Notice that most of the papers used more than one technique. From this, it is possible to realize that CPT and SPT were the more widely-employed tools amongst the published papers. Figure 2.b depicts, from the universe of papers that solely adopted CPT or SPT (or both) as site tool, which interpretation techniques were adopted. It is clearly evident that empirical approaches still form the dominant interpretation group, although in many papers it has been used together with other complementary methods.

INTRODUCTION

Site characterization is the first step on all geotechnical projects and the objectives generally relate to the definition of the stratigraphic profile and groundwater level, estimation of the geotechnical properties from each soil unit, identification of critical layers, definition of geotechnical design parameters and indication of required, if necessary, additional laboratory tests. The traditional methods for site characterization rely basically on drilling, sampling and laboratory tests. These are usually time consuming and, in some cases, over budget. The “modern” approach, on the other hand, focuses on the rational use of in-situ penetration tools coupled in some cases with geophysical techniques. Of course, the success of an efficient site characterization program depends on clearly defining the scope or objectives of the enterprise and, in some cases, combined site investigation techniques are adopted – as will be demonstrated through the papers of this session. Hence, TC102 sessions of the conference contain papers with distinct investigative approaches and scopes. Some have presented new testing devices; others new characterization or interpretation methods. Some have described real case studies where the site characterization was a major issue, whereas others discussed the interpreted soil and rock properties to be used as input for routine geotechnical analyses. Most of the contributions deal purely with in-situ investigation tools, but many have mixed it with laboratory or numerical investigation techniques. As presented in Figure 1, the majority of the papers are “European’ in essence, which is expected for the 18th ICSMGE held in this continent. South America

North America

a) b) Figure 2. a) Percentage of all papers in the session that used the listed in-situ technique among others site investigation tools and b) Percentage of (only) CPT and/or SPT papers in the session that adopted the listed approach to interpret the data, among other techniques.

Figure 3 shows the types of geotechnical formations that served as the major soil stratum for the employed investigative techniques. It is clear that the great majority of the presented papers are concerned with sedimentary deposits, whereas few of them focused on “less classical” materials such as residual (tropical) soils or man-placed tailings and compacted earth fills.

Australia Asia

Europe

Sedimentary

Residual/ Tropical Earth Fill/ Tailing

Africa

Other

Figure 1. Paper distribution by continents for this conference session.

Figure 3. Percentage of geomaterial types addressed in this section.

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013 th


This report is organized into several major topics and subject areas, as follows: site characterization (4 papers), technological advances (9 papers), geotechnical analyses and behavior (14 papers), and soil and rocks properties (15 papers), leading to a total of 42 papers. The main objective is to present an overview and the advances on the main addressed topics of this Conference Session, hence summarizing and emphasizing the important contributions from the papers. Table 1 presents a summary with the main topics and subtopics addressed for each of the listed papers, together with the adopted investigation tools (in-situ or laboratory), the main soil type, the investigation approach, and a short 1-line summary of the paper’s prime objective & contribution. Given such cited divisions, the papers will be reported under each of the defined topics and subtopics, as it follows next.

3.1. New uses of in-situ technology Kim´s et al. paper has an environmental appeal since it deals with the geological CO2 sequestration as an effective mean of reducing the emissions of carbon dioxide. The problem pointed out in the paper is that forward strategies and technologies of CO2 sequestration in Korea need to be specified depending on the geological conditions of potential sites (in Korea). The authors reviewed the geological characteristics of CO2 storage projects around the World and also discuss the suitability for CO2 sequestration. A systematic and quantitative evaluation method to assess the storage and economic efficiencies of sedimentary basins in Korea using dimensionless values between 0 and 1 was applied (Figure 4). The paper also discusses the potential of using geophysical tests to assess the suitability of field strata for CO2-storing, and to monitor CO2 movement and possible leakages.

2. SITE CHARACTERIZATION In this Conference Session, four papers dealt with site characterization; two of them are related to soil classification and the other two are more focused on geotechnical modeling. 2.1. Soil classification The paper written by Serratrice proposes a classification method for natural soils based on piezocone test data. Two aspects are considered to classify the soils with liquefaction potential, the drained and undrained strength via triaxial tests and the soil’s density. The method is presented and applied in two examples where CPTU data are available in homogeneous clayey deposits. The paper from Baud & Gambin presents a contribution to enhance the Pressiorama® diagram with the extra rheological factor “”, which was originally introduced by Ménard on his design method. The authors used very good quality self-bored PMT tests (STAF technique) in several soil types, from soft clays to rock to obtain EM/p*LM values. They proposed a reevaluation of the rheological factor  and the findings are given by an equation graphically expressed in the Pressiorama®. 2.2. Geotechnical modeling The paper from Ivšić et al. discusses the applicability of the RNK-method for spatial engineering & geological and/or geotechnical modeling. This method was tested on many landslides in Croatia and it allows the differentiation of the minimum shear strength zone, or regions of different hydraulic conductivities and varied soil densities. The proposed model was verified by measurements of lateral movements in the landslide area and by results of stability analyses. They concluded that the RNK-method can be used in the study of landslides and slope stability by searching the zone of minimum shear strength. The paper from Steenfelt et al. presents the use of in-situ and laboratory tests for site characterization on an important ongoing infrastructure project in China. A very extensive site investigation campaign was carried out comprising geotechnical boreholes, CPTUs and seismic testing with associated advanced laboratory testing. The paper described the results and the interpretation technique used to provide ground stratification and stiffness variations to be used in design. They concluded that the CPTU was a important tool for a clear geological unit delineation, which also allowed a robust and safe design. 3.

TECHNOLOGICAL ADVANCES

In this Conference Section, nine papers were selected to be part of the technological advances (main) topic; three of them presented new uses of in-situ testing technologies and six dealt with new types of in-situ testing tools (or apparatuses).

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Figure 4. Score for suitability for Korean sedimentary basin (Kim et al).

The paper from Fenton & Hicks discusses the uncertainty associated with site characterization and it focused specifically on the effect of number of samples on residual uncertainty. The results can be used to quantitatively select the required number of samples needed to achieve a target maximum residual uncertainty level. A statistical approach was used to study this problem and it was concluded that the accuracy improves as the number of samples and the correlation length increases. Somasundaram et al. present the characterization and settlement modeling of deep inert debris fills. Inert fills can be considered as a non-text book type geomaterial since they are difficult to characterize and model by current geotechnical methods, due to their inherent heterogeneity, very large particle sizes, and nested and voided structure. The authors presented an approach to characterize a 54 m deep inert debris fill, to model its settlement behavior under seismic loading and groundwater level rise, and to develop remedial measures to render it suitable for development. 3.2. New in-situ tools Jacquard´s et al. paper presents a new probe to overcome the limitation of Menard type pressuremeter tests, i.e., the difficulty of reaching large expansion volumes and pressures. This new device allows for the volume of the hole to be doubled, even under high pressures. The authors described the technological innovations that increased the capabilities (and reliability) of the pressuremeter probe as well as presented comparative tests on different sites to demonstrate the advocated technical advance in this enhanced PMT device. In Rito & Emura paper a new type of sampling method called ‘Koken wire line system’ is developed (Figure 5) to retrieve high depth undisturbed samples in deep Pleistocene clay and sand layers at the Kansai International Airport area. The authors also developed two different pore pressure measuring devices, and concluded that both the sample quality and the measured values were respectively of high quality and with reasonably good accuracy to be used in the settlement design of the subsoil of this airport, in Japan.

Technical Committee 102 / Comité technique 102


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Author / Paper

SPT / BPT / SWS CPT

DMT PMT

Panda/DCP

Geophysics Borehole Sampling

Permeability New Device Other Type

Oedometer

Index

Uniaxial / Triaxial Simple / Direct Ring Shear Proctor / CBR Other Type

Sand / Gravel Silt / Clay

Residual / Tropical

Earth Fill / Tailing

Rock / Saprolite

Instrumentation Numerical

Calibration Chamber

Probabilistic Statistical

Field Data

Experimental

Investigation Approach

Analytical

Table 1. Summary of the general characteristics from each of the papers for Technical Section TC 102 (In-Situ Testing). Main Investigation Tool Main Studied Soil Type Topic In Situ Laboratory Subtopics

Other Type

Empirical

Objective or Major Contribution

X X X X X Classification of soil sensibility via CPT tests S Serratrice X X X X X X X Enhancement of soil type interpretation via PMT tests C Baud & Gambin X X X X X X X X Model for landslide analysis via site correlations G Ivsic et al. Site characterization for tunnel design X X X X X X X X X X X X X X M Steenfelt et al. Kim et al. X X E X X Storage of CO2 emissions in sedimentary basins N Fenton & Hicks X X Technique for optimum soil sampling U Somasundaram et al. X X LS X X X X X Site techniques to characterize and analyze debris fills Jacquard et al. X X X X X X X X Details of a new high volume PMT probe Rito & Emura X X X X X X X Details of a new high depth sampler and piezometer N Kayser et al. X X NS X X X X X Scour evaluation for piers via new in-situ probe X X X X X X X Development of seismic SPT for residual soils I Giacheti et al. Frost & Martinez X X X X X X X CPTu upgrade with a new multi sensor device Monnet X X P X X X X X X Development of a new enhanced PMT probe Yasufuku et al. X X X X X X X X X Rational use of in-situ and lab. tests for foundations D Cao et al. X SS X NS X X X X X X Field instrumentation and results of a pile curtain wall ST X X X Dynamic soil-structure analyses for piles I Hokmabadi et al. Potential use of SDMT in a real case investigation Amoroso et al. X X X X X X X X X X X Haza-Rozier et al. SS X X Improvement of the behavior of a soil foundation Svinkin SS X X X Issues on ground vibration by pile driving Matesic et al. X X X LT NS X X X X X Field tests to monitor the foundation of oil tanks Jeon & Mimura X X X X X Soil foundation deformation of an offshore airport F Chou et al. X X Field survey of affected region after Morakot typhoon X X X X X Soil erosion via multiscale sediment monitoring tool C Lin et al. Al-Saoudi et al. X NS GS X Geotechnical properties of gypseous soils via lab. test Shulyatiev et al. LT X NS X X X Parameters from field load tests on barrette type piles Chen et al. X X X Study of cuttability index for tunnel excavation Bellato et al. X X X X X Assessment of cutter soil mixing samples in the lab. Baud et al. X X X X Shear modulus degradation assessment via PMT tests Benz et al. X X X X X X X X New interpretation approach for Panda penetrometer Nishimura et al. X NS X X X Earth fill investigation using probability analyses N Poulsen et al. X X X X X Influence of CPT penetration rate in silty soils T Galaa et al. X X X X X X X X X Hyd. conductivity determination of glacial deposits Phoon & Ching X X FV X X X X X X In-situ parameters via reliability-based approach Motaghedi et al. X X X X X X X X CPTu strength values via capacity-based equation Tumay et al. X X X X X X X Organic content assessment for sedimentary soils Mulabdic X X X X X X X X X X Characterization of a compacted dam via in-situ tests Zabielska-Adamska X X FA X X Assessment of a compacted soil via CBR tests P Chapuis X X X X X X X X Scale effects in the permeability of sandy aquifers X X X X X X X X Deformation moduli from jointed CPT & DMT tests E Mlynarek et al. Liu et al. X X X X X X X Practice and correlations of CPTu tests in China Espinace et al. X X NS X X X X Control of tailing dams with the Panda penetrometer Correlations on drained compressibility parameters Hanza & Shahien X X X X X X X SC=Soil Classification, GM=Geotechnical Modeling, NU=New Uses of In-Situ Technology, NI=New In-Situ Tools,, DI=Design Improvement, FC=Field Conditions/Site Performance, NT=New Theoretical Advances, PE=Parameter Evaluation. SPT=Standard Pen. Test, BPT=Becker Pen. Test,, SWS=Swedish Weight Sounding, CPT=Cone Pen. Test, DMT=Dilatometer Pen. Test, DCP=Dynamic Cone Probing, LS=Large Scale Density, FV=Field Vane Test, SS=Stress Strain Sensors, LT=Load Test, CBR=California Bearing Ratio, P=Permeability, ST=Shaking Table, E=Eletroresistivity, G=Geophysics, NS=Lab Test adopted but Non Specified, GS=Gypseous Soil, FA=Fly Ash.

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Site Characterization Technological Advances Geotechnical Analysis and Behavior Soil and Rock Properties



Attachment Digital Housing Attachment Digital Boards

(1.66)

(1.40)

fa4

fa3

(0.88)

fa2

(0.67)

fa1 u a0

Friction Sleeve Mandrel

u a1

(0.81)


u a2

(1.07)


u a3

(1.33) (1.14)

Mandrel

u a4

(1.59)


Digital Housing

Replaceable Attachment Friction Sleeve Piezo Sensor

Piezo Sensor

Attachment Sleeve Mandrel

Piezo Sensor

Attachment Individual Piezo Sensor

Piezo Sensor

Attachment Individual Load Cell

Piezo Sensor

Digital Board

Figure 5. The Koken wire line sampling system (Rito & Emura).

The paper from Kayser et al describes an approach to assess soil scour potential through the use of the In-Situ Erosion Evaluation Probe (ISEEP), which is advanced by water jetting. Soil erosion parameters were assessed for silty sand in terms of a critical stream power (critical shear stress and detachment rate coefficient). Scour depths around a circular bridge pier were also computed using ISEEP data, and compared with an empirical approach available in literature. Giacheti et al briefly describes a test which associates the up-hole technique to the SPT, the “seismic SPT” (Figure 6). This hybrid test allows the determination of the maximum shear modulus (G0) together with the N value in a unique test. The paper also presents and discusses cross-hole, down-hole, SCPT and SPT test data for a Brazilian tropical sandy soil to emphasize the advantage of using the interrelationship between the small strain stiffness (Go) and the ultimate strength (N value) to identify and characterize different soil behaviors. Manual SPT Equipment Trigger & Anvil

1

H2

2

H3 Hi

3 i

u2

Pore Pressure

qc

Tip Load

Figure 7. The multi-piezo-sleeve friction penetrometer along with a standard CPT probe (Frost & Martinez).

Monet presents a new in-situ testing device called the “Geomechameter”, i.e. an evolution of the pressuremeter. This new device uses the forces generated by water flow around the probe. The hydraulic flow allows the control of the level of vertical stress at the test depth. The influence of this stress is hence taken into account in the test interpretation. The new probe can also evaluate the soil permeability and sensibility to erosion. It was validated by direct comparison with mechanical properties from triaxial tests and permeability values from Lefranc type injection tests. 4.

GEOTECHNICAL ANALYSIS AND BEHAVIOR

Fourteen papers in this Conference Session were grouped in the topic of geotechnical analysis and behavior; four of them dealt with design improvement and the other ten addressed field conditions and/or site performance. 4.1. Design improvement

DAQ System

Case with geophones H1

fs

Dual Axis Inclinometer Friction Sleeve

L1

L2

L3

Li

Figure 6. S-SPT test and a seismic refracted path (Giacheti et al).

Frost & Martinez enhances the well-established cone penetration test with an extra multi-sleeve penetration attachment (Figure 7). The new CPT probe incorporates a series of friction sleeves with varying surface textures and a torsional load sensing capabilities along with a series of pore pressure sensors, in addition to the standard smooth friction sleeve and pore pressure sensor located behind the tip. They advocate that the multiple measurements made with this device allow it to provide a new insight into the characterization of soil types, besides of establishing relations between stratigraphic variations and in-situ shear strength with the texture height of the sleeves. The authors really consider that the multi-sleeve technology CPT offers significant benefits over other devices to measure the mechanical response of soils.

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The paper from Yasufuku et al. emphasizes the importance of integrating the geotechnical investigations with pile foundation design. Figure 8 shows the policy and concept of geotechnical investigation & design for the studied case, i.e. the construction of the connecting bridge for New-Kitakyushu airport. A rational method for evaluating the pile bearing capacity was presented which reflected the soil characteristic values and the geological environmental history. They concluded that field and laboratory investigations with a reasonable geotechnical consideration sharply decreased the total cost of the bridge in the studied case. The paper from Cao et al. studied the performance of a deep excavation in downtown Toronto. They presented field measurements of soldier pile walls installed into clayey soils and shaly rock. The authors assessed the method of deducing wall bending moments from inclinometer measurements, among other aspects. The paper provides recommendations for such walls when designed in similar geotechnical conditions. The paper from Hokmabadi et al. studies the seismic response of superstructures on soft soils. Shaking table tests and three dimensional numerical simulations using FLAC3D were carried out to investigate the influence of the soil-pile-structure interaction on the seismic response of a 15-storey moment resisting building, supported by end-bearing pile foundations. The authors observed a good agreement between the numerical predictions and the experimental data confirming the reliability of the numerical approach.


th


Design of foundation (Reassessment)

(Feedback)

Assessments based on Geotechnical considerations Investigation

(Collaboration)

design

Select of possible models Decision of sort and number of field & lab. tests Implementation of site investigation

Verification by site investigations

• Full scale load tests • Field observations

Modeling of ground

• Careful selection of soil parameters • Determination of adequate model

Reconsideration of safety factors et al. Implementation of rational and Economical design in total

Figure 8. Collaboration of geotechnical investigations with design (Yasufuku et al).

Amoroso’s et al. paper presents a case history emphasizing the use of the seismic dilatometer (SDMT) as a powerful site investigation tool on the restoration design of an historical building which was damaged by the 2009 L’Aquila earthquake. The investigation of the foundation also included boreholes and laboratory cyclic simple shear tests. The paper presented the interpretation of SDMT for determination of soil profiling, shear wave velocity, constrained modulus and horizontal stress index, which when, combined with lab data, allowed a better understanding of the building’s response during the earthquake. 4.2. Field condition and/or site performance Haza-Rozier et al. study the behavior of a soil foundation improved by rigid columns to support wind turbines. This foundation was fixed on a rigid slab, lying on a granular layer, improved by 84 rigid columns. The authors monitored the structure behavior during excavation, machine construction, and over a period of time for the working service of the wind turbine. They observed that the working platform induced an important confinement of the columns’ heads with subsequent small levels of displacement. Svinkin’s paper discusses the controversial and contradictory evaluations of ground vibrations from pile driving theories. He pointed out that pile driving is a powerful and wide-spread source of construction vibrations which may detrimentally affect adjacent or remote structures. The paper thus presented several issues in the assessment of ground vibrations generated by pile driving. The paper from Matešić et al. presents a case history with the use of hydro test results for designing steel tanks on improved ground with 660 stone columns. The authors described the conducted hydro tests as part of a technical monitoring assessment from all elements of the tank structure. The paper presents and discusses all experimental data and states that they could be wisely used to improve the tank design. Jeon and Mimura present elasto-viscoplastic FEM analyses to assess the long-term deformation of a reclaimed island over a Pleistocene foundation from the adjacent construction of an offshore (twin) airport. It is a numerical modeling paper where simulation was compared to instrumentation results. The authors introduced the concept of “mass permeability” to model the excess pore water pressure dissipation and concluded that it functioned well to assess the long-term deformation of the foundation, including the interactive construction behavior. Chou´s et al. paper discusses survey results of damaged areas after a flood disaster caused by the 2009 Morakot

Typhoon in Taiwan. A comprehensive site survey was conducted after the flood disaster and ten failure mechanisms were identified depending on the different geological environments. The paper presented the site survey observations, analyzed the causes and mechanisms of failures, and drafted strategies and suggestions for the restoration projects. The paper from Lin et al. uses a multi-scale sediment monitoring device to assess the remediation effectiveness on a watershed reservoir after sedimentation processes were originated by the same typhoon cited on Chou et al. It is stated that it caused unprecedented landslide and sediment-related disasters in mountain areas of the Tsengwen reservoir watershed, drastically reducing its storage capacity. Hence, the paper describes the method and how to systematically study and analyze soil erosion and landslide areas with the aid of sediment accumulation trapping dams and aforementioned device. Al-Saudi et al. is another paper that deals with a non-text book type geomaterial: gypseous soils, another “problematic” soil given its intrinsic characteristics. According to the authors, it covers about 20 to 30 % of total Iraq area. An important characteristic of this soil is the collapsibility, a sudden and large volumetric strain when exposed to water. Proposals for soil treatment are presented, focusing on the control of settlement by reducing or even preventing humidity changes within the soil foundation. Shulyatiev´s et al. paper presents a case study related to the construction of the Okhta-center high-rise tower in St. Petersburg. Static load tests on real scale barrette pile types were carried out to adjust the design soil parameters. The paper also presents a comparison between the derived bearing capacity values and those from Russian and foreign building codes. The authors concluded that pile tests are an effective way to calibrate design parameters for usage in real case designs. The paper from Chen et al. presents a generalized (dimensional analysis type) solution to be used into underground geological-mechanical interaction excavation problems. The model groups the geological characteristics into three categories: brittle (rock-like), ductile (soil-like), and brittle-ductile (gravel-like), with respect to thrust and force cuttings. Two case histories are presented to validate the approach to assess the efficiency of a tunnel cutting machine. Bellato´s et al. paper presents a case study to discuss the quality control of Cutter Soil Mixing (CSM), i.e., a relatively new deep mixing method suitable for various types of ground improvement. The materials and the testing program were described in the paper. The obtained results under an innovative experimental apparatus underline the influence of the physical, and chemical, characteristics of the natural soil on the strength gain of the stabilized materials. 5.

SOIL AND ROCK PROPERTIES

In this Conference Session, fifteen papers were selected to be part of this main topic where seven of them presented new theoretical advances as a major subtopic and eight dealt with the evaluation of geotechnical parameters. 5.1. New theoretical advances The paper from Baud et al. discusses stress-strain hyperbolic curves obtained with a self-boring Ménard PMT test. The authors determined E-moduli values by assimilating the pressure-volume plot of a Ménard PMT to a 2nd degree hyperbolic arc. The self-boring Ménard PMT tests were carried out using a self-bored steel slotted tube implemented either by the STAF® technique, or by the ROTOSTAF® method. The authors derived the hyperbolic best fit of the plotted readings to obtain an original equation of the radial borehole expansion, ε = f(G0, po, pLM, PL). After that, they derived the tangent modulus Gt for each reading and the corresponding Gt/G0 ratio as a

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function of ε, and similarly the secant modulus Gs. The authors affirmed that their variation can be well compared with those given by the usual hyperbolic model, except for cases with very small initial strains. They concluded that the results are similar to those obtained by laboratory tests and geophysical surveys. Benz´s et al. paper presents the recent developments achieved on the Panda 3® dynamic penetrometer to improve its use for site characterization. This new improvement is schematically represented in Figure 9 including a typical test result. According to the authors the load-settlement p-sp curve can be derived from the measurement and decoupling of sonic waves created by each impact of the penetrometer, which allows the determination of the strength and deformation parameters. The paper presented calibration chamber test data for two different soils to validate the given results. It was observed a good repeatability and sensibility to the soil conditions. The authors compared the results with those obtained by triaxial and oedometer tests and also found a good agreement for sands. This new test is now currently used in the field to improve the derivation of geotechnical soil parameters via site derived loadsettlement Panda curves.

Figure 9. Schematic representation of Panda 3® dynamic penetrometer with a typical test result (Benz et al).

The paper from Nishimura et al. presents the use of the Swedish Weight Sounding (SWS) test with the objective of making a diagnosis of man made earth-fills, hence increasing their lifetime – especially because their shear strength is generally required for investigations with this scope. The study is justified by the existence of several earth-fill dams for farm ponds in Japan, with some of them under final life stages. Although the strength can also be predicted by the SPT Nvalues, the authors used the SWS test as a simple method for obtaining the spatial distribution of the N-values in short interval exams. The paper also presented an indicator simulation (geostatistical) method to interpolate the spatial distribution of derived N-values. The results are used to determine degraded regions within existing embankments. The shear strength parameter was derived through the empirical correlation with the N-values, and the reliability analysis of the embankments was conducted considering the variability of the internal friction angle of the material. The paper from Poulsen et al. shows how a change in cone penetration rate affects all cone penetration measurements in a silty soil. The authors emphasized the fact that for the standard rate of penetration (20 mm/s) it is generally accepted that undrained penetration occurs in clay, while it is drained in sands. Data from 15 field cone penetration tests with varying penetration rates were conducted at a sandy silt test site. Figure 10 depicts the pore pressure and cone resistance at depths ranging from 4.5 to 11.4 m for CPTs conducted with variable penetration rates (60 and 0.5 mm/s can be observed). The CPT conducted with a penetration rate of 0.5 mm/s corresponds to fully drained penetration conditions, since the measured pore pressure is close to u0. On the other hand, the CPT conducted with a penetration rate of 60 mm/s corresponds to undrained or

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partially drained conditions. The authors did not observe any correlation between sleeve friction and cone penetration rates. They concluded that a correlation between the penetration rate, the cone resistance, and the derived excess pore pressure, do exist. They have also suggested an approach to determine when the penetration is partially drained or not, and how to convert it into a fully drained or undrained condition, hence changing derived geotechnical parameters.

Figure 10. a) Comparison of the pore pressure and b) cone resistance carried out with penetration rates of 60 and 0.5 mm/s, with 3 CPTs test for each rate (Poulsen et al.).

Galaa et al. present a paper describing a methodology for establishing more representative design values for the hydraulic conductivity (K) of glacial deposits, particularly when performing large scale subsurface investigations for tunnels. They justify their study given the known glacial deposit heterogeneities and the difficulties to determine proper design values for K. The authors affirm that conventional pumping tests can not provide reliable design parameters due to their small zone of influence, and inherent variable nature of glacial deposits. Hence, the paper describes a subsurface investigation which involved 400 boreholes, 88 slug tests and 16 pumping tests. The authors established a correlation between K from the field tests (Kfield) and K calculated by the Kozeny-Carman formula (KKC). They observed that the Kozeny-Carmen formula with the incorporation of a site specific correlation factor predicted K values ranging between 1/3 to 3 times the Kfield values. The calculated and measured K values were used to form a statistical analysis of this parameter, and to provide a more reliable design number for dewatering problems. Phoon & Ching present a paper using a statistical approach for a better interpretation of the geotechnical data when considering soil variability. The paper presented the concept of a “virtual site” with the purpose of emulating site investigation efforts as realistically as possible. The authors affirmed that in the present time, it is still not possible to emulate every aspect of a real site deposit. So, the scope was to reproduce the information content arising from a typical mix of laboratory and field tests conducted at a site with the aim of estimating undrained shear strengths (su) for clays and friction angles (') for sands. However, the development of a virtual site does not replace the site investigation need, but it quantifies the uncertainty in the derived su and  design values by incorporating into the analyses the effect of either higher quality or larger numbers of testing results. Motaghedi et al. present a new analytical method to predict cohesion (c) and friction angle () using qc, u and fs from the piezocone test, considering the bearing capacity mechanism of failure at the cone tip and a direct shear failure along the penetrometer sleeve. The authors state that one of the advantages of this method is the improvement of the accuracy in the case of (eventually) using erroneous data related to all three outputs from the CPTu test. The paper presented laboratory test results, together with two sets of nonlinear equations derived by the proposed approach and existing correlations for both c and � parameters. The authors state that the  obtained by current techniques is relatively higher than real measured values. However, when adopting the advocated method, the comparisons indicate a good consistency with lower scatter.

Technical Committee 102 / Comité technique 102 th


5.2. Parameter evaluation

of structures. The paper presents deformation characteristics estimated from CPTU and SDMT tests in clays, and focuses on a method to identify soil preconsolidation and to establish relationships between deformation moduli derived from CPTU and SDMT tools. The authors concluded that the simultaneous use of CPTU and SDMT provides a continuous picture of the changes in stiffness of heterogeneous subsoil. They emphasized the need for establishing specific calibration functions for each soil type, which may be a useful tool in the construction of a model for the subsoil’s rigidity based on G0 or M0 values. Liu´s et al. paper reports the practice and development of the piezocone test in the geotechnical engineering field of China. In this paper, the history and current development status of CPT and CPTu in China practice were systematically presented. The most used (standard) cone has the 10 cm2 tip area, but both 15 and 20 cm2 CPT probes are frequently used in China. The relationship between international standardized CPTu and China’s CPT is based on a large data bank of testing results related to a great number of soils. The paper presents a comparison review of the soil characterization methods in China, including the determination of stress history, deformation, consolidation and permeability characteristics. The paper from Espinace et al. presents their 10 years of experience on the use of Panda® penetrometer to assess the stability of Chilean’s tailings dams. The authors report around 40 cases of mechanical instability from tailing dams in Chile, which were mainly due to liquefaction, slipping of banks, or settlement. The paper presents the main results that have allowed the proposition of a new methodology to control and diagnose tailing dams. It is based on in-situ determination of the geomechanical parameters (internal friction angle and density index) using the Panda® penetrometer in order to characterize the constituent materials and their variability. The authors also pointed out that this methodology allows taking into account the variability concept for stability and liquefaction risk studies when using a probabilistic approach. Hamza & Shahien´s paper studies the compressibility parameters of Egyptian cohesive soils via piezocone tests. The major objective was to provide additional data on drained compressibility parameters, focusing on the constrained modulus (Mo) and on the overconsolidation ratio (OCR) for cohesive soils from geotechnical investigations at seven major sites of the Nile delta river deposit in Egypt. Enhanced propositions to estimate the OCR and Mo for the studied clays are presented, allowing settlement analyses to be done with the proposed equations. The authors believe that the presented data and correlations are a valuable contribution, since it improves the current state of the art in estimating the compressibility parameters of sedimentary soils with the CPTU test.

The paper from Tumay et al. discusses the challenge for the effective identification of organic content in the soil based on traditional CPT and CPTU methodologies. It is very important to overcome this interpretation limitation since the cone is a popular and handy tool for subsurface investigations and soil characterization. The paper presents a comprehensive CPT/CPTU-based organic content identification method using a probabilistic soil classification system. The paper describes the probabilistic method, which employs a non-traditional modeling approach that takes the uncertainty of the correlation between soil composition and soil behavior into account. The authors affirmed that the use of the compositional soil classification (U) and in-situ behavior (V) indexes for organic profiling improves the capability of determining organic material at any given depth. A detailed description of the proposed methodology and the discussion of its effective application are included in the paper. Mulabdic´s paper presents the use of penetration testing devices, including the CPT and SDMT, for site characterization of a compacted earth dam. This is a case study of a small earth dam for which the remediation work was necessary given construction errors and the possible damage to the earth structure during the filling stage of the reservoir. The site investigation campaign consisted of drilling boreholes and carrying out in-situ tests (4 CPTs and 3 SDMTs) along the crest of the dam, complemented with laboratory tests. The paper focused on assessing the potential of these in-situ tests in describing physical and mechanical properties of the compacted (man-made) clay strata, since the traditional interpretation methods were developed for natural soils. The authors concluded that both CPT and SDMT clearly detected the inhomogeneous clay conditions. They also showed remarkable repeatability and proved to be valuable tools in characterizing the embankment quality, both in terms of non homogeneity and of physical and mechanical properties. Zabielska-Adamska & Sulewska present the use of both static (classic) and dynamic CBR methods to establish relationships between the bearing ratio and degree of compaction of fly ash. The objective was the use of the compaction degree, and also the California Bearing Ratio, as an indicator of the soil bearing capacity in compacted material. The dynamic CBR test is described in the paper, where fly ash samples were compacted by the standard and modified Proctor methods without soaking to replicate field conditions during earth structure construction. Test results indicate that both the dynamic CBR as well as the classic CBR are closely connected with the characteristics of compaction, and can therefore be used to assess the compaction of fly ash and cohesive soils. The authors suggested that the dynamic CBR test should be widely used as an alternative way to the classical method of quality control to assess the subgrade capacity of the soil. The paper from Chapuis discusses “scale effects” in the permeability of sandy aquifers. The author’s initial hypothesis is that the large-scale tests are more likely to meet preferential flow paths, so yielding larger K values than small-scale tests, which may be viewed as some sort of scale effect. In the paper, the small scale was simulated via lab soil samples, the middle scale from field permeability tests, and the large scale with site pumping tests. The paper presents and discusses some few real case studies, observing that for all of them the K distributions provided consistent images of the aquifers. It was finally concluded that scale effect was not of importance for the test interpretation in such phreatic deposits. Mlynarek´s et al. paper discusses the interrelationship between deformation moduli from CPTU and SDMT tests in overconsolidated soils. The authors point out that glaciations in Poland overconsolidated its deep soil layers. So, it is imperative to take it into account in calculations of differential settlements

6.

FINAL REMARKS

Site characterization using in-situ testing techniques has considerably changed in the last two decades along with the rapid transformation and advances of the technology, either by the development of newer and economical electronic devices operated by laptop computers or by new mathematical and software approaches based on multi-variable, statistical or probabilistic calculations. Besides of such remarkable accomplishments, the traditional “old fashion” (past century….) laboratory and site investigation methods are still widely in use, sometimes as the preferential or unique available method. It was clear from aforementioned review that, on the 21 st century, the proper site investigation, material characterization and soil behavior prediction for the geotechnical design cannot solely rely in one isolated test technique, or on simple “local” unadjusted correlations that are probably not universally valid. Higher sensorial levels of testing tools and combined investigation procedures are surely now available that can be

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applied on a more regular basis, especially for large or important jobs. Improved interpretation methods or background geotechnical theories, advanced computer software codes, and more prominent hardware equipment, should further be explored in research as well as in practical in-situ testing settings. This is a challenge for the overall community as a whole, from practitioners to academicians, manufacturers, and designers. Nevertheless it can not be accomplished without a strong change in mentality from the geotechnical field itself, moving from a “priced-based” design to perhaps a more expensive and sound “quality-based” criteria. The papers presented in this Conference Session show how in-situ testing technology has developed, matured, and used to tackle several geotechnical problems of difficult order, for instance from the uncertainty in site characterization and understanding of different geomaterials, to the challenging task of retrieving high quality soil samples in a difficult environment. Sometimes, in standard project cases, only traditional tests were required and used for the site characterization. However, in more complex or ambivalent conditions, the usage of today’s available technological advances was surely an asset for the design. Although not directly mentioned throughout this review, the human factor, i.e., the good education based on solid concepts of the geotechnical area together with the access to a free flow of technical information and knowledge, will be the key factor for the transformation of our field, and the future society, as we all envisage – with rational use of resources and technology, selfsustained projects, quality based design and environmentally safe site procedures. 7.

ACKNOWLEDGEMENTS

The authors acknowledge the hard work of some of the Ph.D. candidates from the Geotechnical Graduation Program of the University of Brasília during the early stages of paper compilation and description. Therefore they are indebted to the work of the engineers Osvaldo Neto, Edgar Rincon and Raydel Lorenzo. Additionally, this report would not be possible without the use of the language skills from engineer Adrien Treguer, native in French, who is a student from Clermont-Ferrand University and fortunately happens to be at the moment in an undergrad exchange program with the University of Brasilia. 8.

REFERENCES

Al-Saoudi N.K.S.; Al-Khafaji A.N., Al-Saoudi N.K.S. Challenging problems of gypseous soils in Iraq. Amoroso S.; Totani F., Totani G. Site characterization by seismic dilatometer (SDMT): the Justice Court of Chieti. Baud J.P., Gambin M. Détermination du coefficient rhéologique  de Ménard dans le diagramme Pressiorama®. Baud J.P.; Gambin M., Schlosser F. Courbes hyperboliques contrainte– déformation au pressiomètre Ménard autoforé. Bellato D.; Simonini P.; Grisolia M.; Leder E., Marzano I.P. Quality control of Cutter Soil Mixing (CSM) technology – A case study. Benz M. A.; Escobar E.; Gourvès R.; Haddani Y.; Breul P., Bacconnet C. Mesures dynamiques lors du battage pénétromètrique– Détermination de la courbe charge enfoncement dynamique en pointe. Cao L.F.; Peaker S.M., Ahmad S. Performance of a deep excavation in downtown Toronto. Chapuis R.P. Permeability scale effects in sandy aquifers: a few case studies. Chen L.; Chen Y.C.; Chen W.C., Liu H.W. A study of cuttability Indices for tunnel penetration. Chou J. C.; Huang C. R., Shou K. J. Survey results of damaged areas in flood disaster of Typhoon Morakot and suggestions for restoration projects.

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Espinace R. A.; Villavicencio G. A.; Palma J.; Breul P.; Bacconnet C.; Benz M.A .N, Gourvès R. Stability of Chilean’s tailings dams with the Panda® penetrometer. Experiences of the last 10 years. Fenton G.A.; Hicks M.A. Site sampling: assessing residual uncertainty. Frost J. D., Martinez A. Multi-sleeve axial-torsional-piezo friction penetration system for subsurface characterization. Galaa A.; Manzari M., Hamilton B. Hydraulic properties of glacial deposits based on large scale site investigation. Giacheti H. L.; Pedrini R. A. A., Rocha B. P. The seismic SPT test in a tropical soil and the G0/N ratio. Hamza M., Shahien M. Compressibility parameters of cohesive soils From piezocone. Haza-Rozier E.; Vinceslas G.; Le Kouby A., Crochemore O. Comportement de la structure de sol amélioré par inclusions rigides, supportant une éolienne. Hokmabadi A.S.; Fatahi B., Samali B. Seismic response of superstructure on soft soil considering soil-pile-structure interaction. Ivšić T.; Ortolan Ž., Kavur B. Applicability of the RNK-method for geotechnical 3D-modelling in soft rocks. Jacquard C.; Rispal M.; Puech A.; Geisler J.; Durand F.; Cour F.; Burlon S., Reiffsteck P. Une nouvelle sonde permettant de mesurer sans extrapoler la pression limite pressiométrique des sols. Jeon B.G., Mimura M. Long-term Deformation of the reclaimed pleistocene foundation of the offshore twin airport. Kayser M., Gabr M. Assessment of scour potential using in-situ jetting device. Kim A. R.; Cho G.C.; Kwon T. H., Chang I. H. Practical reviews on CO2 sequestration in Korean sedimentary basins and geophysical responses of CO2-injected sediments. Lin B.S.; Ho H.C.; Hsiao C.Y.; Keck J.; Chen C.Y.; Chi S.Y.; Chien Y.D., Tsai M.F. Using multi-scale sediment monitoring techniques to evaluate remediation effectiveness of the Tsengwen Reservoir watershed after sediment disasters induced by Typhoon Morakot. Matešić L.; Mihaljević I.; Grget G., Kvasnička P. The use of hydro test results for design of steel tanks on stone column improved ground a case history. Młynarek Zb.; Gogolik S., Sanglerat G. Interrelationship between deformation moduli from CPTU and SDMT tests for overconsolidated clays. Monnet J. Le Géomécamètre, un nouvel essai in-situ adapté à la mesure des caractéristiques hydro-mécaniques du sol. Motaghedi H.; Eslami A., Shakeran M. Analytical approach for determining soil shear strength parameters from CPT & CPTu data. Mulabdic M. Use of penetration testing for determination of soil properties in earth dam. Nishimura S.; Shuku T., Suzuki M. Diagnosis of earth-fills and reliability-based design. Phoon K.K., Ching J. Construction of virtual sites for reliability-based design. Poulsen R.; Nielsen B. N., Ibsen L. B. Correlation between cone penetration rate and measured cone penetration parameters in silty soils. Rito F., Emura T. Sampling method and pore water pressure measurement in the great depth (-400m). Serratrice J.F. Une méthode de classification de la sensibilité des sols au moyen du piézocône. Shulyatiev O.; Dzagov A.; Bokov I., Shuliatev S. Correction of soil design parameters for the calculation of the foundation based on the results of barrettes static load test. Somasundaram S.; Khilnani K.; Shenthan T., Irvine J. Characterization and modeling settlement of deep inert debris fills. Songyu. L; Guojun. C; Anand J. P., Yanjun D. Practice and development of the piezocone penetration test (CPTu) in geotechnical engineering of China. Steenfelt J.S.; Yding S.; Rosborg A; Hansen J.G., Yu R. Site characterization of the HZM immersed tunnel. Svinkin M.R. Controversial and contradictory evaluations in analysis of ground vibrations from pile driving. Tümay M. T.; Hatipkarasulu Y.; Marx E. R., Cotton B. CPT/PCPTbased organic material profiling. Yasufuku N.; Ochiai H., Maeda Y. Geotechnical challenge for total cost reduction related to construction of connecting bridge with pile foundations. Zabielska-Adamska K., Sulewska M.J. CBR as a method of embankment compaction assessment.

Challenging Problems of Gypseous Soils in Iraq Des problèmes difficiles des sols gypseux en Irak Al- Saoudi N.K.S.

University of Technology-Baghdad-Iraq

Al- Khafaji A.N., Al- Mosawi M.J. University of Baghdad – Baghdad- Iraq

ABSTRACT: Gypseous soils are classified as one of the problematic soils due to their complex and unpredictable behaviour. They exist in many parts of the world, concentrated mainly in arid and semi-arid regions. In Iraq gypseous soils cover about 20 to 30 % of its total area concentrated primarily on the west desert and extended to the southern parts and directed towards south west. Gypsum soils experience sudden collapse upon exposure to water, losses of serviceability of many structures were observed in different parts of Iraq. Extensive research was made in Iraq to investigate and understand the behavior of Gypsum soils and to set safety limits for the collapse and suggest practical precautions during construction. The enormous amount of data collected from different research sources revealed wide spectrum of information covering the overall performance of Gypsum soils under different environmental and climate conditions The present paper focuses on the main geotechnical properties of gypseous soils and their effect on the collapsible mode of failure, some practical solutions are also proposed that provide safety precautions RÉSUMÉ : Les sols gypseux sont classés comme des sols problématiques à cause de leurs comportements complexes et imprévisibles. Ils existent dans plusieurs régions du monde, principalement dans des régions aride et semi-aride. En Irak, les sols gypseux couvrent entre 20 et 30 % du pays et sont principalement concentrés dans le dessert de l’ouest et s’étendent vers les régions du sud et orientées vers le sud-ouest. Les sols gypseux s’effondrent soudainement lorsqu’ils sont soumis à l’eau et beaucoup de structures inutilisables ont été observées dans différentes zones en Irak. Des recherches approfondies ont été menées en Irak pour étudier et comprendre le comportement des sols gypseux afin de déterminer les limites avant l’effondrement et de suggérer des précautions concrètes lors de la construction. L’énorme quantité de données recueillies auprès de différentes sources a révélé un large spectre d’informations couvrant l’ensemble des performances des sols gypseux sous différentes conditions environnementales et climatiques. Le présent document se concentre sur les principales propriétés géotechniques des sols gypseux et sur leurs effets sur les écroulements de structure, enfin quelques solutions pratiques sont aussi développées pour proposer des mesures de sécurité. KEYWORDS: Gypsum,collapsibility, Gypseous Soils, Problematic Soils 1. INTRODUCTION Gypseous soils are one of the most complex materials that challange the geotechnical engineers. Structures or dams founded on gypseous soil may experience unpredictible deformations that ultimatley may cause catostrophic failure. In iraq it has been reported that several structures have experienced different patterns of cracks and uneven deformations generated primearly from the exposion of the supporting gypseous soils to water. It is a well known fact that gypseous soils demonstrate high bearing capacity and very low compressibility when they are in the dry state. On the contrary a sudden collapsibile behaviour was reported when the gypseous soils are exposed to water.The collapsibility of gypseous soils results from the direct contact of water. The dissolution of different types of salts contained inside the mass of gypseous soil will generate new pores inside the soil skeleton and loosen the cementing bonds between the particles. This process creates a meta stable structure that facilitates the sliding of particles into a more dense state. The rate of dissolution of gypsum depends primarily on environmental changes in moisture content generating from fluctuation of ground water table and /or surface water, climate changes typically temperature, permeability and state of flow conditions in addition to the type and content of gypsum. During the last three decads many attempts were made in Iraq through intensive research programs set in many institution to investigate and underestand the behaviour of gypseous soils under various enviromental and loading conditions. The first ob-

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jective of these research programs was to determine the physical proerties of the gypseous soils and to check whether staandard tests can be used and if not what modifications are required. Following that several attempts were made to determine the geotechnical properties such as compressibility, collapsibility and shear strength parameters under various flow and environmental conditions and loading conditions. The tests were performed using triaxial and Rowe cells allowing soaking and leaching of the soil samples. Plate load tests were also performed under different soaking periods to moniter the generated deformation with time. Numerical techniques were also used to simulate the disolution process of gypsum under soaking and leaching conditions. The abundant amount of data obtained from the lengthy research programs revealed in many cases contradicting results due to the complexity of the gypseous soils. So no regid conclusions are yet been drawn. The paper sheds the light on the distribution of gypsum in Iraq. A summary of main physical and geotechnical properties with emphises on the collapsibility is presented and a some remidied are proposed.

Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

2. DISTRIBUTION OF GYPSEOUS SOILS IN IRAQ Gypseous soils exist mainly in arid and semi arid regions, concentrates in contenents like africa,central and souther asia. Iraq is among the contries of south asia where gypsum covers about 12 % of its total area. (FAO 1990), although more recent study (Ismail 1994) reported that gypseous soils cover 31.7 % of the total area of iraq.the first map demonstrating the distribution of gypsum in iraq was presented by (Buringh 1960) indicating five zones as shown in figure 1. The primary gypsum is located in the extreme north area between tigres and euphraties rivers. The second zone where primary gypsum mixed with limstone located below and parrallel to the euphraties river extending from the west desert to the south. The secondary gypsum is identified in two areas, one in the north below the first zone and one in the south – west. The fourth zone is gypsiferous alluvium extens from the north in a narrow band and gradually widened towards the south. The fifth zone representing the non gypsiferous soil, mainly limestone is identified in two ares one in the north east and the other in the west desert.

This indicates how serious the challanges are for geotechnical engineers when dealing with such unpredictible material.

Slightly over gypsum bedrock Moderately to highly gypseous soils over gypsum and anhydrate rock Gypsum desert Highly gypsiferous soils on Pleistocene terNon to slightly gypseous soil Moderately to highly gypsiferous associated with lime

Figure 2 Distribution of gypsum in Iraq (Al Barazanji 1973) 3. PROPERTIES OF GYPSEOUS SOILS Figure 1 first map of distribution of gypsum in iraq A more refined map exhibiting the distribution of gypsum in iraq was presented by (Al-Barrazanji 1973). He investigated thoroughyl the type and gypsum content in different parts of iraqand proposed the map shown in figure 2.Six zones are distiguished according to their origin and gypsum content.Zone one of slightly gypseous over gypsum bedrock denoted by narrow parallel lines taking the shape of a triangle in the upper north of Iraq. The second zone is of moderately to highly gypseous soils over gypsum and anhydrate rock denoted by wider parallel lines, located in the north part between the Tigress and Euphrates rivers. Zone three is gypsum desert denoted by a mesh of small squares, located between zones one and two in the north. Zone four contains highly gypsiferous soils on Pleistocene terraces covering two narrow strips on the left and right of Tigress River denoted by moderately dense dots. The fifth zone is non to slightly gypseous soils denoted by parallel hashes, extends from the upper mid third of Iraq up to the Kuwaiti borders in the south. The sixth zone is moderately to highly gypsiferous soil associated with lime denoted but heavily condensed dots, covering the west jazeria. The two maps comply each other in most of their subdivisions with slight divergence in others, although different terminologies have been used. Based on figure two, if the soil in zone four is considered as non gypseous soil that does not possess any hazardous impact then most likely 50 to 60 % of the totalarea of Iraq is covered with active gupsum.

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The Physical, chemical and geotecnical properties of gypseous soils collected from different parts of Iraq are summarized and discussed below. 3.1. Physical properties The physical properties of natural gypseous soils varies considerably with the amount and type of gypsum soil in addition to the texture and constituents of the soil 3.1.1. Specific gravity (Schultz and Cleaves 1955) reported that the specific gravity of gypsum ranges between 2.31 to 2.33, increased to 2.95 for anhydrite type. Thus any increase in gypsum content of specific gravity less than 2.33 will lead to a decrease in specific gravity of thr soil. (Saleam 1988,Nashat 1990 and Al- Mufty 1997) reached to the same finding unless the gypsum of the unhydrated type. 3.1.2. Maximum dry unit weight The results of maximum dry unit weight showed contradicting relationship with gypsum content.(Khattab 1988 and AlDulaimy 1989) found that the dry unit weight increases with increasing gypsum content up to a certain limit followed by a gradual drop. On the other hand (Subhi 1987 and others) reported a decrease in


dry unit weight with increasing gypsum content. Further more,(Al Heeti 1990) showed an increase in dry unit weight with increasing gypsum content. This descrepency may be due to the type of gypsum ( hydriate or anhydriate), type of soil and range of gypsum content considered in the investigation.

The same researchers and many others showed that the coefficient of consolidation remains unchanged with increasing gypsum content.

3.1.3. Soil constiuents and texture

Gypseous soils are distinguished by their collapsible behaviour upon wetting. The term collapse potential is used to classify the hazarduse state of collapsibility. ( Jennings and Knight 1957) proposed a double oedometer collapse test to predict the collapsibility of the foundation soil. Two identical samples are tested, one at natural water content and the other after submergeing in water for one day.The collapse potential C.P. is defined as

The samples of gypseous soils collected from different parts of Iraq showed that gypsum exists primearly in Sandy soil and silty sand and less in silty clay or clayey silt. The presence of appreciable amount of gypsum creats problems in determining the constituents of the soil. During sedementation test disolution of gypsum will occur causing the flucculation of silt and clay particles.Pretreatment with water was suggested by (Al-Khashab 1981 and Mohammed 1993). EDTA was suggested by ( ALKhuzaie 1985 and others). Most of the treated agents used cause distruction of bonds and most likely revealed an unreliable amount of constituents. 3.2. Chemical properties

3.3.2. Collapsibility

C.P. = ∆e / 1+ eo

(1)

Where ∆e is the difference in void raio of the two samples at a specific stress eo is the natural void ratio The severity according to the collapse potential is shown in table 1.

3.2.1 Chemical compsition of gypsum Pure chemical proportions of gypsum as reported by (Nashat 1990) are 20.9 % combined with water 46.6 % sulphur trioxide SO3 32.5 % calcium oxid CaO (Majeed 2000) observed that the alkalinity increases with increasing gypsum content. On the other hand the electrical conductivity, cation exchange capacity and exchangeble sodium percentage decrease with increasing gypsum content. 3.2.2. Solubility of gypsum The most effective parameter in the general behaviour of gypsous soils is the solubility. Gypsum is classified as a moderate soluable salt. The solubility of the hydrated type in pure water is 2g/l ( Hesse 1971). Some higher values, 2.41g/l and 2.6g/l , were reported for Iraqi gypseous soils ( Sirwan etal 1989, Seleam etal 1988)

Table 1. Collapse identification ( Jennings and Knight 1975) Severity No prob- Moderate Trouble Severe Very lem Severe C.P. % 0-1 1-5 5-10 10-20 > 20

(Saleem 1988, Nashat 1990 and many others), found that the collapse potential under a constant stress of 200kN/m2 increases with increasing gypsum content. The gypsum content of the tested samples ranged between 20 - 80% revealed a moderate type of 4 % maximum collapse potential. 3.3.3. Moduluse of deformation

The rate of dissolution of gypsum is responsible for the development of cavities and sinkholes. It is very complex to be evaluated as it is affected by many environmental conditions such as temperature,source of water,time, concentration of sodium chlorid and calicum sulphate etc.

Al Khafaji etal 2009 investigated the deformability of gypseous soils through plate load tests performed on natural and soaked soils. The tests were performed on two sites GP-GM soil and SM soil and socking period extended to 7 -11 days under 300 mm head of water. All types of stiffness moduli were calculated, the initial tangent moduluse, the permissible secant modulus at half the yeild, the yeild secant modulus at the yeild and the yeild tangent modulus after the yeild. The outcomes revealed that soaking decresed the stiffness moduli in the range of 2 to 5 folds for GP-GM soil and from 2 to 3.5 for SM soil. The field tests highlights on the hazardius degree of constructing structures on gypsious soils without awarness of the expected generated settlements that may result from the contamination of water.

3.3. Geotechnical properties

3.3.4. Hydraulic conductivity

The geotechnical prperties of gypseous soils cover, compressibility, collapsibility, permeability and shear strength parameters ( c and Ø)

Hydraulic conductivity or coefficient of permeability of gypseous soils is very hard to predict. Standard constant head test on sandy gypseous soils does not reveal reliable results as the gypsum disolves during flow creating more free space for the soil particles to reorient themselves to a closer state of packing, causing a suddent fluctuation of rate of flow during test This phenomenon is very difficult to evaluate as the dissolution process is influenced by many factores like type and amout of gypsum, hydralic gradient, initial placement of soil sample. Attempts were made to perform leaching permeability tests under different stress levels using Rowe cell (Al-Kaisi 1997 and many others). (Al- Qaissi 2001 and many others) using triaxial permeability leaching apparatus cited that the variation in hydraulic gradient combined with diffusion of gypson encountered serious difficulties in predicted reliable values of the coefficient of permeability.

3.2.3 Rate of dissolution of gypsum

3.3.1. Compressibility More than ten researchers have investigated the influence of gypsum on the copressibility characterstics.(Al-Khashab 1981 and many others) reported a decrease in the copression index with increasing gypsum content. It is hard to judje about the contradicting results as many parameters such as the placement conditions, degree of disturbance, and testing methodology. Similar contradicting results were reported for the recompression index. Most of the researchers demonstrated an increase in the secondary compression index with increasing gypsum content. This phenomenon is attributed to the contieous dissoltion process of gypsum with timeas reprted by (Saleam 1988 and Nashat 1990).

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4. PROPOSED REMEDIES FOR GYPSEOUS SOILS

5. CONCLUSIONS

The invenitable geotechnical problems associated with the abandance of gypseous soils in Iraq create real challanging issues. Based on that proposals were made for treatment of soils focussing on controlling the settlement and reducing the coefficient of permeability or preventing any contact of water between the foundation soil and any source of water. All the proposed treatments are based on elemt tests and not verified by field applications

Inspite of the abandant data collected concerning the the geotechnical properties of gypsefeous soils and the attempts to investigate and understand the behaviour of gypseous soils under different stresses and environmental conditions.The challanges still exist due to the scarceyt and complexity of such natural material. No real firm solution or a general improvement technique can be proposed. It is advised that geotechnical engineers must investigate each case seperatly depending on the type of structure, characteristics of site, environmental conditions coupled with the engineering judgement of the consultant.

4.1. Chemical treatment The treatment materials proposed are basically cement, lime and petroleuum products 4.1.1. Treatment with cement (Khattab1986) reported that sulphate resisting cement improved the unconfined compression strength of granular gypsified soil, but a substaintial amount of reduction in strength and stiffness upon immersion in water. 4.1.2. Treatment with lime Al-Obaidy 1992 and Al-Zory 1993 showed that mixing 5 -7 % lime with gypsous soil of 43 % gypsum content exhibited an increase in strength and high resistance to leaching. It is reported that the soil became practically impermeabile after 28 days curing. 4.1.3. Treatment with petroleum products Various types of petroleum products such as kerosene, automobile oil ,fuel oil and bitumenous materials such as S-125 and R250 were proposed as improvement agents for gypseous soils. (Saleam 1988) found that treating soil of gypsum content between 40-50 % with kerosene caused a decrease in compressibility and permeability by delaying the removal of gypsum.(AlAqaby 2001) observed a reduction in cohesion of soil of gypsum content between 30 -67 % upon immersion in water or kerosene. The angle of internal friction was reduced by 6 degrees upon soaking in kerosene. (Al-Kaisi 1997) found that 4 % automobile oil caused a reduction in the coefficient of permeability by not less than ten folds. (Al–Hassany 2001)perfromed consolidation tests on two samples of gypsum content 26 % and 51% treated with fuel oil. The fuel oil tends to to fill the pores of soil and prevent water perculation and hence reduce the permeability. The presence of fuel oil also reduced the copmressibility and collapsibility. Bitumenous materials S-125 and R-250,emulsified asphalt, Cut-Back MC-30 were used by (Al-Morshedy 2001 and many others). Gypseous soils treated with one of the above materials showed reduction in coefficient of permability as well as compressibility and collapsibility. 4.2. Physical treatment (Al-Khafaji 1997) developed simple and quick equations for estimating the optimum water content and maximum dry unit weight to control field compaction of soils with gupsum content ranging between 0.5 -50 %.

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6. REFERENCES FAO. 1990 . Management of gypsiferous soils. Food and Agricultural Organization of United Nations Rome. Internet http//fao.org/docrep/to323e/ro323e03.htm Ismail H.N. 1994. The use of gypseous soils. Symposium on Gypsiferous soils and their effect on structures. NCCL Baghdad. Iraq Buringh P. 1960. Soils and soil conditions in Iraq. Ministry of Agriculture . Baghdad. Iraq Al-Barazanji A.F. 1973. Gypsiferous soils in Iraq. PhD. Dissertation. Ghent University. Belgium. Shultz J.R. and Cleaves A.B. Geology in engineering.John Wiley and sons . New York Saleam S.N.1988. Geotechnical characteristics of gypseois sandy soil including the effect of contamination with some oil products. MSc. Theisis . University of Technology. Baghdad.Iraq. Nashat I.H. 1990. Engineering characteristics of some gypseous soils in Iraq. PhD. Thesis . University of Baghdad. Iraq Al- Mufty A.A. 1997. Effect of gypsum dissolution on the mechanical behaviour of gypseous soils. PhD. Thesis . University of Baghdad. Iraq Khattab S.A. 1986. Effect of gypsum on strength of cement treated granular soil and untreated soil. MSc. Thesis . university of Mosul. Iraq Al-Dilaimy F.H. 1989. Effect of gypsum content on strength and deformation of remolded clayey soil. MSc.Thesis University of Salahddin. Iraq Subhi H.M. 1978 The properties of salt contaminated soils and their influence on the performance of rocks in Iraq. PhD. Thesis Queen Mart College. University of London Al-Heeti 1990. The engineering properties of compacted gypsified soil. MSc. Thesis. University of Baghdad. Iraq Al- Khashab M.N. 1981 Investigation of foundation soil behaviour of Qadisiah site –Mosul. MSc. Thesis. University of Mosul. Mohammed R.K. 1993 Effect of wetting and drying of engineering characteristics of gypseous soils. MSc. Thesis University of technology. Baghdad. Iraq. Al-Khuzaie H.M.A.1985 The effect of leaching on the engineering properties of Al-Jezirah soil.. MSc. Thesis . university of Mosul. Iraq Majeed A.H. 2000. Data base for gypseous soils. PhD. Thesis University of Baghdad. Iraq. HesseP.R. 1971. A textbook of soil chemical analysis. Chemical publishing co..Inc.New York pp520 Jennings J.E. and Knight K.1957. The additional settlement of foundationsandy subsoil on wetting. Proceeding 4th Int. Conf. Soil mechanics and foundation engineering . vol.1. pp316-319 Al- Khafaji A.N, Al-Mosawi M.J., Khorshid N.S. and Al-Obaid B.M. 2007 Proceeding of the 17th ICSMGE Alexandia .Eygept pp 727729. Al- Khafaji A.N. Densification of gypseous soil by compaction. Symposium on ground improvement geosystems. London Al-Morshedy A.D. 2001 The use of cutback MC-30 for controlling the collapsibility of gypseous soils.MSc.thesis. University of Technology. Baghdad.Iraq.

Site characterization by seismic dilatometer (SDMT): the Justice Court of Chieti Caractérisation du site par dilatomètre sismique (SDMT): la Cour de justice de Chieti Amoroso S., Totani F., Totani G. University of L’Aquila, Italy

ABSTRACT: A detailed investigation of several seismic dilatometer (SDMT) tests was performed in 2011 on Chieti hill to restore the Justice Court, an historical building damaged by the April 6, 2009 L’Aquila earthquake. Moreover, boreholes were carried out to investigate foundation base level and cyclic simple shear tests with double sample were realized to analyze the seismic site response. The paper illustrates the potential of the seismic dilatometer to efficaciously approach a geotechnical problem by the interpretation of SDMT parameters, as the shear wave velocity VS, the constrained modulus M and the horizontal stress index Kd. Fi- nally, the paper combines SDMT results with laboratory data to analyze the site response of the Justice Court. RÉSUMÉ : Une étude détaillée de plusieurs sismiques dilatomètre (SDMT) tests a été réalisée en 2011 sur la colline de Chieti pour restaurer la Cour de justice, un bâtiment historique endommagé par le tremblement de terre qui a eu lieu le Avril 6 2009 à L'Aquila . En outre, des sondages ont été effectués pour étudier le niveau de base de fondation et cycliques essais de cisaillement simple avec échantillonnage double ont été réalisées pour analyser la réponse sismique du site. Cet article montre efficacement le potentiel de la dilatomètre sismique à l'approche d'un problème géotechnique par l'interprétation des paramètres SDMTs, comme la vitesse de l'onde de cisaillement VS, le module M et l'indice de contrainte horizontale Kd. Enfin, le document combine les résultats SDMT aux données de laboratoire pour analyser la réponse du site de la Cour de justice. KEYWORDS: seismic dilatometer, horizontal stress index, shear wave velocity, site response analysis, local site effetcs.

1

INTRODUCTION

The April 6, 2009 L’Aquila (Italy) earthquake (MW = 6.3) caused heavy damages not only in the city of L’Aquila basin but also in few cities, as Chieti, approximately 100 km far from the epicenter. In this respect, a detailed investigation of several seismic dilatometer (SDMT) tests (Marchetti et al., 2008) in virgin soils and inside boreholes backfilled with sand (Totani et al. 2009), foundation boreholes and cyclic laboratory tests were performed in 2011 on Chieti hill to restore the Justice Court, an historical building damaged by the above mentioned earthquake. The geotechnical campaign allowed to characterize the subsoil, to investigate foundation base level and to analyze the seismic site response of this construction. In particular, the paper illustrates the potential of the seismic dilatometer to efficaciously approach a geotechnical problem by the interpretation of SDMT parameters, as the shear wave velocity VS, the constrained modulus M and the horizontal stress index Kd, even combinig SDMT results with laboratory data for the evaluation of the local site effects (e.g. topography, soil conditions) with mododimensional (1D) and bidimensional (2D) seismic site response analyses. 2

GEOTECHNICAL INVESTIGATION ON CHIETI HILL

A detailed investigation of eleven SDMT tests, six in virgin soil, each 10-20 m in depth, and five inside boreholes backfilled

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with sand, each 30-50 m in depth, were performed in 2011 on Chieti hill to restore the Justice Court, an historical building damaged by the April 6, 2009 L’Aquila earthquake. Moreover, boreholes were carried out to investigate foundation base level and cyclic simple shear tests with double sample were realized to analyze the seismic site response. The historical centre was built on sandy and arenaria deposits (45 m in depth), while moving towards the bottom of the slope the colluvial cover start to emerge over the OC silty clay, as shown in Figure 1 together with the shear wave velocity VS profiles. Figure 2 emphasizes the main reason of the damage due to the seismic action on the construction. As shown by the inspection, the structure appears to be considerably fissured in its Southern part, while in the Northern area it seems to be intact (Figure 2a). This aspect can be justified referring to the four SDMTs performed along the perimeter of the building (Figure 2b). SDMT1 and SDMT4, as well as SDMT3 and SDMT2 profiles, can be coupled. In fact, in the Northern part of the Justice Court the constrained modulus M reaches on average values over 100 MPa and the horizontal stress index Kd indicates OC soils. Instead, in the Southern area, until about 8.00 m in depth, M assumes very low values (under 50 MPa) and Kd is about equal to 2 and thus Kd individuates NC layers (TC16, 2001). In both the cases VS appears less sensible to the stress history and the stiffness of the deposits compared to M and Kd. In addition, the boreholes on the foundations illustrate that in the Southern part


the base level is at about 4.90 m in depth. Then, before intercepting sandy and arenaria deposits, it was found a landfill layer from 4.90 to 8.00 m. In terms of stiffness it means that in the Northern portion of the structure the foundations stand on soil with higher mechanical properties compared the soil in the Southern part. A possible solution to restore the historical build-

ing is to improve the stiffness of the soils placed in the Southern portion, without acting on masonry foundations. This achievement could be realized for example, using, from 4.90 to 8.00 m in depth, special injections at low pressure, able to penetrate and mix with the existing soil structure.

Figure 1. Geotechnical cross section with VS profiles.

Figure 2. (a) Justice Court: site investigation by Seismic Dilatometer; (b) SDMT results: M, Kd and VS profiles.

3

SITE RESPONSE ANALYSIS

Numerical analyses of seismic site response were carried out using the computer codes EERA (Bardet et al. 2000), a monodimensional linear equivalent model, and QUAD4M (Hudson et al. 1994), a bidimensional linear equivalent model. that considers a cross section of 3.5 km of width, with 5860 elements and 5844 joints. The evaluation of the local site effects (e.g. topography, soil conditions) plays an important role in the non-uniform amplification response obtained at different sites (Paolucci 2002). In order to compare the 1D and 2D analyses, the 1D elastic response spectrum were multiplied by the topographic amplification factor, assumed equal to 1.2 (CEN 2003). Both the analyses were performed on the top of Chieti hill, in correspondence of Southern portion of the Justice Court. Moreover, a 1D compari-

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son was carried out between the Northern portion (absence of filling material) and the Southern portion (presence of filling material) to evaluate the site effects due to the different mechanical behaviour of the upper 8 meters. 3.1

Input ground motions

For the numerical analyses two natural accelerograms, applied on the bedrock, were selected as input ground motions. Both the accelerograms were chosen from the software REXEL (Smerzini et al. 2012) and the Italian Accelerometric Archive ITACA (Working Group ITACA 2010). The first accelerogram “UM_EW” is the strong motion recorded at the Assisi station (Italy) during the September 26, 1997 Umbria-Marche (UM) earthquake (Mw = 6, on outcrop, normal fault, site-source distance ≈ 20 km), scaled, according to CEN (2003), to a peak ground acceleration of 0.164g, for a return period TR = 475 years and a soil type, for the site of Chieti. The


second accelerogram “VN_NS” is the strong motion recorded at the Cascia station (Italy) during the September 19, 1979 Val Nerina (VN) earthquake (Mw = 5.8, on outcrop, normal fault, site-source distance ≈ 9 km), scaled to the same peak ground acceleration of UM earthquake. 3.2

Geotechnical model

The geotechnical model of Chieti hill, used in the numerical analyses, is illustrated in Figure 1 and 3 and Table 1, by including the soil and dynamics parameters (unit weight γ, Poisson coefficient ν, shear wave velocity VS, stiffness decay curves G/G0 and damping D curves). 18

1

0.6

16

1,2 - Fillin g material 4,5 - San d , San dstone

14

6 - OC Silty clay

12

7 - OC Silty clay

10

3 - Silty clay co lluvial cover 1, 2 - Filling material

0.4 0.2

8

4,5 - San d , San dstone 6 - OC Silty clay

6

7 - OC Silty clay

4

Damping Ratio D (%)

Normalized sher mo dulus G/G0

0.8

2 0 0.0001

0.001

0.01

0.1

1

0

Shear Strain (%)

Figure 3. Stiffness decay curves G/G0 and damping D curves of Chieti hill for numerical analyses. Table 1. Geotechnical model of Chieti hill for numerical analyses.

1

Material

γ

ν

(kN/m 3)

VS (m/s)

Filling material

17.1

0.30

220

2

Filling material

17.1

0.30

440

3

Silty clay colluvial cover

18.7

0.45

280

4

Sand, sandstone

20.6

0.30

580

5

Sand, sandstone

20.6

0.30

870

6

OC silty clay

20.2

0.45

600

7

OC silty clay

20.2

0.45

800

8

Bedrock

21.0

0.30

1000

9

Bedrock

22.0

0.30

1300

Results

3.3

1D numerical analyses of seismic site response were carried out by considering the Northern portion and the Southern portion. The spectral accelerations (Figure 4) highlight the different mechanical behaviour of the upper 8 meters: the Southern portion shows pronounced amplifications for a period of 0.1-0.2 s, that is the fundamental period of the filling material, while the spectral accelerations of Northern portion appear lower. 2D numerical analyses of seismic site response were performed by considering in the Southern portion two point, A and B, 20 m far from each other, shown in Figure 3. The peak ground accelerations (Figure 5) doesn’t appear influenced by spatial position and input ground motion, even thought the analyses consider only two time histories. In addition, the spectral accelerations emphasize the site effect due to the topography: point A, closer than point B to the hillside, shows higher amplifications for a period of 0.2-0.4 s, compared to the ones of point B. The comparison of the average results from 1D and 2D numerical analyses in the Southern portion of the Justice Court (Figure 6) illustrates that the 1D peak ground accelerations are higher than the ones evaluated from 2D analyses, probably due to the higher sensitivity of 1D model to stratigraphic effects. In addition, 2D method shows local site effects mainly due to topography for a period of 0.3-0.4 s, that is the fundamental period of the Justice Court. 1.8 UM_EW Southern portion 1D VN_NS Southern portion 1D

1.6

UM_EW Northern portion 1D

1.4

VN_NS Northern portion 1D Average Southern portion 1D

Spectral accel eration Sa (g)

Layer

an average value interpolated from the experimental relationship Crespellani et al. (1989) and SDMT profiles. The site campaign of the Justice Court had provided only a cyclic simple shear tests with double sample in OC silty clay. In this respect, the following reference laboratory curves were assumed to evaluate the non-linear and dissipative soil behaviour: Anh Dan et al. (2001) for filling material, Marcellini et al. (1995) for sand and sandstone, MS–AQ Working Group (2010) for silty clay colluvial cover. The bedrock has G/G0 - γ and D – γ linear behaviour.

On the top of the hill, in correspondence of the Justice Court, the subsoil was modelled by considering in the upper 8 m filling material in the Southern portion of the Justice Court and sand and sandstone in the Northern portion, sand and sandstone between 8 m and 42 m of depth, OC silty clay between 42 m and 342 m of depth and the bedrock beyond 342 m of depth, while on the hillside the model reflects the silty clay colluvial cover in the upper 15 m up to the OC silty clay layer. In the upper 50 meters the VS profile was defined as an average of SDMT profiles, while in the lower OC silty clay VS was estimated by using

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1.2

Average Northern portion 1D

1.0

0.8

0.6

0.4

0.2

0.0 0.0

0.5

1.0

1.5

2.0

2.5

3.0

Period T (s)

Figure 4. Spectral accelerations form 1D analyses.

3.5

4.0


Bardet J.P., Ichii K., Linn and C.H. 2000. EERA – A Computer Program for Equivalent-linear Earthquake site Response Analyses of Layered Soil Deposits. University of Southern California. CEN, EuropeanCommittee for Standardisation (2003) Eurocode 8: design provisions for earthquake resistance of structures, Part 1.1: general rules, seismic actions and rules for buildings, prEN 1998-1 Crespellani T., Ghinelli A. and Vannucchi G. 1989. An evaluation of the dynamic shear modulus of a cohesive deposit near Florence, Italy. Proc. XII ICSMFE, Rio de Janeiro. Hudson, M., Idriss, I.M., and Beikae, M. 1994. QUAD4M: A Computer Program to Evaluate the Seismic Response of Soil Structures using Finite Element Procedures and Incorporating a Compliant Base. Center for Geotechnical Modeling, Dep. of Civil & Env. Engng, University of California, Davis. Marcellini A., Bard P.Y., Vinale F., Bousquet J.C., Chetrit D., Deschamps A., Marcellini A., Iannaccone G., Romeo R.W., Silvestri F., Bard P.Y., Improta L., Meneroud J.P., Mouroux P., Mancuso C., Rippa F., Simonelli A.L., Soddu P., Tento A. and Vinale F. 1995. The Benevento Seismic Risk Project. I- Seismotectonic and Geotechnical Background. Proc. 5th International Conference on Seismic Zonation, Nice, France 1: 802- 809. Marchetti S., Monaco P., Totani G. and Marchetti D. 2008. In Situ Tests by Seismic Dilatometer (SDMT). In J.E. Laier, D.K. Crapps & M.H. Hussein (eds), From Research to Practice in Geotechnical Engineering, Geotechnical Special Publication No. 180: 292–311. ASCE. MS–AQ Working Group. 2010. Microzonazione sismica per la ricostruzione dell’area aquilana. Regione Abruzzo—Dipartimento della Protezione Civile, L’Aquila, 3 vol. & Cd-rom (in Italian). Paolucci R. (2002). Amplification of earthquake ground motion by steep topographic irregularities. Earthquake Engineering and Structural Dynamics, 31: 1831-1853. Smerzini C., Galasso C., Iervolino I. and Paolucci R. 2012. Engineering ground motion selection based on displacement-spectrum compatibility. Proc. 15th World Conference on Earthquake Engineering, Lisbon, Portugal, September 24-28, 2012. TC16. 2001. The DMT in Soil Investigations. A Report by the ISSMGE Committee TC16. May 2001, 41 pp. Reprint in R.A. Failmezger & J.B. Anderson (eds), Flat Dilatometer Testing, Proc. 2nd Int. Conf. on the Flat Dilatometer, Washington D.C.: 7–48. Totani G., Monaco P., Marchetti S. and Marchetti D. 2009. Vs measurements by Seismic Dilatometer (SDMT) in non-penetrable soils. In M. Hamza et al. (eds), Proc. 17th Int. Conf. on Soil Mechanics and Geotechnical Engineering, Alexandria, 2: 977–980, IOS Press. Working Group ITACA. 2010. Data Base of the Italian strong motion records: http://itaca.mi.ingv.it

1.8 UM_EW Southe rn portion 2D point A VN_NS Southern portion 2D point A

1.6

UM_EW Southe rn portion 2D point

1.4

B VN_NS Southern portion 2D point

Spectral acceleration Sa (g)

B Average Southe rn portion 2D

1.2

point A Average Southe rn portion 2D point B

1.0

0.8

0.6

0.4

0.2

0.0

0.0

0.5

1.0

1.5

2.0

2.5

3.0

3.5

4.0

Period T (s)

Figure 5. Spectral accelerations from 2D analyses. 1.8 Average Southern portion 1D

1.6

Average Southern portion 2D point A Average Southern portion 2D point B

Spectral acceleration Sa (g)

1.4

1.2

1.0

0.8

0.6

0.4

0.2

0.0

0.0

0.5

1.0

1.5

2.0

2.5

3.0

3.5

4.0

Period T (s)

Figure 6. Comparison between 1D and 2D spectral accelerations.

4

CONCLUSION

The paper illustrates the potential of the seismic dilatometer to efficaciously approach a geotechnical problem by means of the results analyses. While VS appears less sensible to both the stress history and the deposits stiffness, M gives precious information on soil stiffness, while Kd provides for important details about the deposits overconsolidation. Combining SDMT results with laboratory data it has been possible to evaluate the the local site effects by means of 1D and 2D seismic site response analyses of the Justice Court. These numerical analyses indicates that in complex stratigraphic and topographic conditions, it appear appropriate to combine 1D and 2D methods. 5

ACKNOWLEDGEMENTS

This study was founded by Provincia di Chieti and Studio Prof. Marchetti s.r.l. 6

REFERENCES

Anh Dan,L.Q., Koseki,J. and Tatsuoka,F. 2001. Viscous deformation in triaxial compression of a dense well-graded gravel and its model simulation. In Tatsuoka et al. (eds) Advanced Laboratory Stress- Strain Testing of Geomaterials, Balkema, pp.187-194.

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th


Détermination du coefficient rhéologique  de Ménard dans le diagramme Pressiorama®. Obtaining the Ménard  Rheological Factor in a Pressiorama® Diagram. Baud J.-P.

Eurogéo, France

Gambin M.

Apagéo, France

RÉSUMÉ: Louis Ménard a défini le coefficient rhéologique  à l’aide du rapport adimensionnel des deux caractéristiques classiques de l’essai pressiométrique EM/p*LM. La connaissance de ce rapport est un élément essentiel de la validité des calculs de déformation du sol au contact de toutes sortes de fondations. Sa valeur dépend simultanément de la qualité de réalisation du forage pressiométrique, et de la proportion entre cohésion et frottement dans la résistance du sol soumis à l’essai, c’est-à-dire de sa nature. Dans le but de compléter par un axe des valeurs de  le diagramme Pressiorama® désormais présent dans certaines normes d’application de l’Eurocode 7, nous nous sommes étalonnés sur des essais pressiométriques autoforés par la technique STAF dans des sols divers allant de sols très mous à des rochers très massifs. En effet ces essais, de très bonne qualité, permettent d’obtenir des valeurs EM/p*LM allant de 4 pour les sols purement granulaires à plus de 100 pour les sols cohérents très consolidés et les roches. Les auteurs proposent ici, à partir de ces données, une expression, à la fois chiffrée et graphique, de la valeur du coefficient rhéologique , basée sur 3 paramètres EM, pLM et p0 - ce dernier estimé ou, mieux, mesuré et donc variable selon la profondeur de l’essai dans le sol. ABSTRACT: In the early years of the development of his “direct method” of design based on PMT results, Louis Ménard introduced a rheological factor  based on the ratio EM/p*LM for each test. The knowledge of this factor is necessary to estimate settlement and horizontal displacement of all sorts of foundations. Its value is a function of both the quality of the borehole and the ratio between soil friction and cohesion, i.e. its nature. In order to complete the Pressiorama® diagram with a  values axis, the authors used a calibration mostly based on so-called self-bored PMT tests performed with the STAF technique, in various soil types from soft clay to rock. These very good quality tests permit obtaining EM/p*LM values from 4 in granular soils to 100 in highly consolidated soils and rock. From these data, the authors propose an evaluation of this rheological factor  only based on the values of 3 parameters, namely, EM, p*LM and the earth pressure at rest p0, either estimated, or, much better, measured during the early part of the test. Results are given under the shape of an equation and graphically on the Pressiorama® diagram.

MOTS CLÉS : Pressiomètre, autoforage, classification des sols, coefficient rhéologique . KEYWORDS: Ménard pressuremeter, self-boring, soil classification,  rheological factor. 1

INTRODUCTION

Le rapport adimensionnel EM/p*LM des deux caractéristiques classiques de l’essai pressiométrique Ménard est un facteur complexe et puissant, qui dépend simultanément de la qualité de réalisation du forage pressiométrique, et de la proportion entre cohésion et frottement dans le comportement du sol soumis à l’essai, c’est-à-dire de sa nature. Son utilisation par Ménard pour définir le coefficient rhéologique  est un élément essentiel de la validité des calculs de déformation du sol par les méthodes pressiométriques. Au cours des années récentes, en raison de l’accroissement des essais pressiométriques produits par des opérateurs manquant de formation et de maîtrise des techniques de forage les mieux adaptées à chaque type de sol, techniques spécifiques à cet essai mais remarque valable aussi pour tous les essais géotechniques, de nombreux utilisateurs ont décelé une distorsion entre les prévisions de tassement par la méthode pressiométrique et les déformations réellement observées sur les ouvrage construits. La quasi-totalité des pratiques de forage entraînant le remaniement des parois de forage pressiométrique allant toujours dans le sens d’une diminution parfois dramatique des modules mesurés, les prévisions de tassement qui en ont été déduites deviennent notoirement pessimistes, ceci alors même que la méthode de calcul, confirmée par les normes et

4871

règlementations nationales et européennes, a montré depuis longtemps sa fiabilité lorsqu’elle est appliquée à des données de terrain acquises dans des conditions de qualité normale. (Baguelin et al. 1978). Avec l’émergence de codes de calcul aux éléments finis, de nombreux ingénieurs ont pensé qu’il était possible, voire nécessaire, de délaisser la célèbre règle souvent nommée T-0 (Ménard & Rousseau, 1962) pour le calcul du tassement à partir d’un profil de modules pressiométriques, au profit d’une modélisation numérique complexe tenant compte de la géométrie de l’interface sol-structure, et de traiter le problème de la déformation en appliquant à ce modèle des lois de comportement basées sur l’élasticité linéaire, à l’aide d’une « corrélation » entre le module pressiométrique et un module d’Young. Cette approche nous semble vouée à l’échec, comme chaque fois que l’ingénieur croit pouvoir fait fi de la méthode expérimentale et la remplacer par des calculs que l’on prend pour rigoureux à raison de leur complexité (Briaud & Gibbens 1994, Gambin 2003, Gambin 2010). Notre approche de ce problème majeur relatif à la crédibilité des prévisions de déformation faites par l’ingénierie géotechnique, se distingue de cette tendance aux modélisations complexes, et vise plutôt à assurer ce qui fait l’originalité et la cause du succès de la méthode pressiométrique, c’est à dire atteindre rapidement un



résultat par un calcul simple et fondé sur une connaissance la plus parfaite possible de la loi contrainte-déformation de la sollicitation pressiométrique du sol vierge. Dans ce sens, les essais autoforés dans la plus large gamme de sols possible sont une bonne réponse (Arsonnet et al., 2005), car ils permettent à la fois la mesure du module initial G0, de la loi de décroissance du module avec la contrainte (Baud & Gambin, 2005 ; Baud & Gambin, 2008 ; Baud et al., 2012 et 2013), enfin l’évaluation du coefficient  (Baud, 2005 ; Baud & Gambin, 2012). 2 PARAMETRES PRESSIOMETRIQUES NECESSAIRES ET SUFFISANTS A LA DETERMINATION DE . 2.1

Module EM, pression limite p*LM et confinement de l’essai.

La première présentation du coefficient rhéologique  en tableau à double entrée, en fonction du rapport EM/p*LM et de l’état de consolidation du sol, qui avait été donnée dans la notice D60 sur les « règles d’exploitation des techniques pressiométriques et d’exploitation des résultats obtenus pur le calcul des fondations » (Ménard, 1968) a été reconduite telle quelle dans les règlementations françaises puis européennes. Ménard prévoyait pourtant de réviser le tableau « en fonction de résultats d’essais expérimentaux », raison principale pour laquelle les valeurs de  étaient données sous forme de fractions unitaires. Mais le principe était établi :  est croissant quand EM/p*LM est croissant, depuis le comportement sableux jusqu’au comportement argileux, et croissant également lorsque le sol passe de la sous-consolidation ou de l’altération à la surconsolidation. Pour suivre cette règle usuelle, des lignes d’isovaleurs des valeurs fractionnaires de  peuvent être tracées directement dans le diagramme bi-logarithmique Pressiorama® (Baud, 2005) en log (p*LM), log (EM/p*LM), tel qu’il a été édité en annexe des normes NF P94-261 et NF P94-262 (Fig. 1), la valeur 1 formant la limite supérieur du diagramme et la valeur ¼, la plus faible du tableau de Ménard, étant affectée au rapports EM/p*LM très faibles, correspondant soit à des sables et graviers, soit à des sols très remaniés, remaniement naturel in situ, ou bien lié à une mauvaise exécution du forage pressiométrique. La principale raison de tracer des droites en coordonnées bilogarithmiques, était que l’on ne dispose absolument pas, dans le référentiel des confrontations entre méthode pressiométrique et fondations instrumentées, de données suffisantes pour proposer des courbes plus sophistiquées. Ces droites sont donc des constructions mathématiques simples, basée sur l’hypothèse :

Figure 1 Valeurs de  dans le diagramme Pressiorama® tel qu’il peut apparaître dans les normes NF-P94-261 et NF-P94-262 (en projet).

carotté au voisinage, la position des lignes d’isovaleurs de  par rapport à p*LM n’est plus compatible dès lors que l’on considère des essais réalisés à des profondeurs importantes : ceci est lié au fait que ce n’est pas la valeurs absolue de p*LM qui doit être prise en compte, mais la valeur relative p*LM/p0, sans dimension, qui tient compte du confinement de l’essai par la pression horizontale des terres au niveau où elle est mesurée. On est ainsi conduit à la formulation suivante, dans laquelle kE, m et n jouent le même rôle. 1



(2) m

p * LM n k E .  p 0  

1

 EM n  p * LM    m k .  p * LM  n

EM n  p * LM  

Le nouveau diagramme [log (p*LM/p0), log (EM/p*LM)] qui en résulte n’est pas tracé ici. Il constitue une présentation peu intuitive, essentiellement parce que le géotechnicien est très habitué à ce que représente la pression limite, directement proportionnelle à la résistance du sol, alors que la grandeur sans dimension p*LM/p0 est difficile à saisir. Elle représente en quelque sorte le degré de résistance du sol soumis à l’essai par rapport à une augmentation « normale » de résistance attendue croissante en fonction de la profondeur. Les coefficients m et n peuvent prendre en théorie une large gamme de valeurs, mais en pratique seule une faible gamme de variations laisse les droites représentatives toutes présentes et étalées dans le champ de vision du diagramme. Nous proposons de les arrêter à m=0,5 et n=2, et avec ce couple de coefficients, kE prendra une valeur comprise entre 3 et 5 pour que  reste compatible avec l’usage. Nous avons retenu ici la valeur entière kE=4, d’où :

(1)

Sous cette forme en effet, les coefficients m et n (en exposants) et le facteur k, tous sans dimensions prennent empiriquement les valeurs nécessaires pour assurer le tracé désiré : m détermine l’angle des droites iso-, n détermine l’écartement entre les valeurs, et k détermine, pour un couple de valeurs (m, n) donné, la position de la ligne maximale  = 1. Les valeurs de  ainsi proposées peuvent être rendues assez conformes au tableau à double entrée de Ménard pour des essais à profondeur moyenne de quelques mètres utilisés pour des fondations superficielles. Mais on voit assez vite que si les valeurs de EM/p*LM décrivent toujours assez bien la nature du sol, sableux, intermédiaire ou argileux, qui peut être connu par ailleurs au moment du forage pressiométrique ou, mieux par un sondage

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013 1

EM 2  p * LM    1 4 p * LM   4.  p 0  

2.2

(3)

3 PROPOSITION D’UN NOUVEAU CADRE POUR LE DIAGRAMME PRESSIORAMA®.

Quelle est la relation possible entre E Young et EM ?

Tout d’abord, il paraît peu judicieux de comparer le module d'Young mesuré par traction sur des solides polycristallins (barres métalliques) où la déformation est linéaire jusqu'à la fin de la phase strictement élastique, et le module de déformation des sols, variable avec l’intensité de la contrainte, cette déformation étant de nature hyperbolique tout au long de l’application de la charge. C’est la raison qu’expose Ménard, dans son article fondateur de Sols-Soils n°1 (Ménard & Rousseau, 1962), pour créer la notion de coefficient rhéologique  pour lequel il propose simultanément la gamme des valeurs fractionnaires dans différents sols, et une expression basée sur le module alterné Ea : 1     

 E  2     Ea 

ici. En effet, dans un modèle (q,) ramenant le comportement du sol à une phase élastique linéaire bornée par un critère de rupture fixe, celui-ci impose bien un rapport unique entre le module E et la valeur choisie pour le déviateur q.

(4)

1     

 E  2     Ea 

E étant aujourd’hui noté EM , et  étant « une valeur faible comparée à ½, dépendante de facteurs secondaires ». Ménard pensait approcher avec Ea la valeur du module de microdéformations, noté alors E ; on s’accorde plutôt aujourd’hui à voir dans Ea une approche d’un module élastique EY. Paraphrasant Ménard on peut écrire ; EY = EM/n

(4b)

avec nO2, soit en négligeant les facteurs secondaires : EY = EM/²

(4c)

Pour des raisons non explicites, l’habitude avait été prise en France par les utilisateurs de résultats pressiométriques, de tronquer cette proposition et de retenir Ey = EM/ pour des estimations du module d’Young qui ne donnent pas satisfaction. Il y a eu sans doute attraction et confusion liée au fait que Ménard a plus tard également indiqué que  = EM/E+, E+ « module de déformation du sol dans un champ quasiisotrope », plutôt assimilable donc à un module de type œdométrique et non à un module d’Young. Des utilisateurs de longue date des méthodes pressiométriques avaient gardé l’usage de cette relation entre Ea et EM pour donner une estimation de  à partir d’essais cycliques (R Heintz, 2012). En identifiant  entre les relations (3) et (4), il vient : 1

EY 2 = 16 .  p * LM  p p * LM 0  

(5)

Cette relation remarquable qui élimine EM et  et établit une relation directe entre Ey et p*LM, découle directement du choix fait ci-dessus pour les coefficients m et n. En effet, quelle que soient les valeurs adoptées pour ces coefficients, il se maintient toujours une forte corrélation entre un paramètre de rupture, p*LM, et un module d’Young définissant une relation linéaire élastique, donc constituant une corde sur la courbe pressiométrique, entre son origine (p = p0) et un point situé vers le milieu de l’intervalle p0 - pLM c’est-à-dire proche de la pression de fluage de l’essai. La mise en évidence de cette relation confirme bien la définition initiale de Ménard et l’expression qui en est proposée

En vue de déterminer la valeur du coefficient rhéologique  pour chaque essai pressiométrique dont on connait normalement et simultanément la pression limite, le module pressiométrique, et la profondeur permettant d’estimer p0, ou mieux la valeur mesurée de p0, il est possible de proposer une façon différente de placer les résultats d’essais, dans un nouvel abaque construit de la façon suivante (Fig. 2): - en abscisse, le coefficient  , en échelle logarithmique et en valeurs décroissant de gauche à droite. - en ordonnée, placé sur  = 1, le module pressiométrique relatif EM/p0, qui est donc un nombre sans dimension, en échelle logarithmique et en valeurs croissantes vers le bas. - l’axe des pressions limites relatives p*LM/p0 vient se placer en oblique des deux axes, avec un angle variable selon les rapports d’échelles. - l’axe des rapports EM/p*LM est alors conjugué et orthogonal à l’axe p*LM/p0. Chaque essai pressiométrique est représenté par un point unique au croisement de ses 4 caractéristiques. Limité vers le haut par la ligne EM/p*LM = 3 au-delà de laquelle on ne doit pas trouver de matériau naturel ou fabriqué, l’abaque est un triangle rectangle englobant tous les types de sols, roches et matériaux fabriqués. La base, que l’on tronque plus ou moins tôt selon que l’on s’intéresse plus à la mécanique des roches, ou à celle des sols, ou au domaine intermédiaire, représente les matériaux cimentés. Les sols très mous, les vases et boues sont dans la pointe également tronquée. Les sols habituels de la géotechnique sont entre ces deux extrêmes, et sont ici qualifiés dans un quadrillage de 3 fois 3 cases, N°1 à N°9, dont les matériaux sont identifiés dans la légende de la figure. L’expérience réduite d’essais pressiométriques que nous avons personnellement dans le rocher franc (Baud & Gambin, 2011 et 2012) nous a permis de confirmer les zones N°10, pour les graves et roches très fracturées, N°11, pour les roches tendres, fracturées ou altérées et N°12 pour les roches très dures. Elles sont également en accord avec des études antérieures sur les roches (Failmezger et al., 2005). 4

CONCLUSIONS

Cette étude a permis de montrer qu’il est possible d’introduire un axe des  comme abscisse de notre diagramme Pressiorama®, l’ordonnée étant EM/p0, et d’en graduer la valeur en fonction des résultats obtenus au pressiomètre, ce qui n’était pas évident a priori. (Fig. 2). Deux autres axes apparaissent transversalement aux coordonnées cartésiennes : la pression limite relative p*LM/p0 et le rapport bien connu EM/p*LM. L’abscisse  est également normée arithmétiquement en définissant le comportement pressiométrique du sol par un indice de granularité g = [2.Ln()]/kE, proportionnel à un angle de frottement déductible directement de l’essai : M = arctan (g). Nous ne considérons pas notre travail comme terminé, car il est nécessaire qu’il soit confronté à de nombreux résultats d’essais pressiométriques dans les roches en particulier. Et il est possible que cette confrontation, ainsi que des études de corrélations entre EY et p*LM, conduisent à une évolution de notre schéma dont les bases paraissent cependant bien acquises.

489


th


Figure 2 Diagramme Pressiorama en coordonnées bilogarithmiques adimensionnelles [, module relatif EM/p0].

5

REFERENCES.

Arsonnet, G., Baud, J.-P., Gambin, M.P. 2005. Réalisation du forage pour essais pressiométriques par un système de tube fendu autoforé (SFAF), Actes du Symp. Intern. ISP5 – PRESSIO 2005, sous la direction de Gambin, M., Magnan, J.-P., & Mestat., Paris, 22-24, Août. 2005, Vol.1, Paris: Presses des Ponts. Baguelin F., Jezequel J.F. Shields D.H. 1978. The Pressuremeter and foundation engineering. Trans Tech Publications, Clausthal, Germany, chap. 3 pp 284-291. Baud, J.-P, Analyse des résultats pressiométriques Ménard dans un diagramme spectral [log (pLM), log (EM/pLM)] et utilisation des regroupements statistiques dans la modélisation d’un site, Actes du Symp. Intern. ISP5 – PRESSIO 2005, Paris, sous la direction de Gambin, M., Magnan, J.-P., & Mestat, P. 22-24 Août. 2005, Vol.1, Paris: Presses des Ponts. Baud, J.-P, Gambin, M. P. 2005. Déduction d’une loi de réponse hyperbolique unique par complilation de courbes pressiométriques dans un sol de lithologie homogène Actes du Symp. Intern. ISP5 – PRESSIO 2005, sous la direction de Gambin, M., Magnan, J.-P., & Mestat, P., Paris, 22-24, Août. 2005, Vol.1, Paris: Presses des Ponts. Baud, J.-P, Gambin, M. P. 2008. Homogenising MPM tests curves by using a hyperbolic model, in Huang, A.-B., & Mayne, P. W. (eds) Geotechnical an Geophysical Site Characterization, Proc. ISC’3 Taipei, Taiwan, 1-4 April 2008, London: Taylor & Francis Baud, J.-P, Gambin, M. P. 2011, Classification des sols et des roches à partir d’essais d’expansion cylindrique en haute pression, C. R. du 15ème Congrès Européen de Mécanique des Sols et de

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Géotechnique, edited by A. Anagnostopoulos, M. Patchakis, C.Tsatsanifos, ISO Press, Amsterdam . Baud, J.-P, Gambin, M. P. 2012. 50 MPa Ménard PMTs help linking Soil and Rock Classifications. in A. Anagnostopoulos, ed., Geotechnical & Geological Engineering Journal, Special Issue on Hard Soils and Weak Rocks, Springer Verlag, Berlin. Baud, J.-P, Gambin, M. Schlosser F. 2012. Stress-strain hyperbolic curves with Ménard PMTs in R.Q. Coutinho ed., Geotechnical and Geophysical Site Characterization, Proc. ISC’4 Porto de Galinhas, Brazil, Sept.18-21, 2012, London: Taylor & Francis Baud, J.-P, Gambin, M. Schlosser F. 2013. La courbe contraintedéformation au pressiomètre Ménard Actes du 18ème CIMSG, Paris, 1-5 sept. Briaud, J.-L., & Gibbens, R. 1994, Test and Prediction Results for Five Spread Footings on Sand ASCE Geotechnical Specification Publication No.41 Failmezger, R., Zdinak, A., Darden, J., Fahs, R. 2005, Use of Rock Pressuremeter for Deep Foundation Design Actes du Symp. Intern. ISP5 – PRESSIO 2005, sous la direction de Gambin, M., Magnan, J.-P., & Mestat, P., Paris, 22-24, Août. 2005, Vol.1, Paris: Presses des Ponts. Gambin, M. 2003. Etude élémentaire d’un mythe. FONSUP 2003, Symposium International sur les fondations superficielles, Paris, 57 novembre 2003, p.251-254, J.P. Magnan & N. Droniuc éd., Presses de l’ENPC, Paris. Gambin M. 2010. Les théories et leur évolution face à la réalité en Géotechnique. VIIème Conférence Coulomb, Comité Français de Mécanique des Sols, Paris. Heintz R. 2011, Communication personnelle. Eurasol, Luxembourg. Ménard L. Rousseau J. 1962. L’évaluation des tassements, tendances nouvelles. Sols-Soils, N°1, Paris. Ménard L. 1968. Règles d’exploitation des techniques pressiométriques et d’exploitation des résultats obtenus pur le calcul des fondations. (en anglais, trad. Française TLM notice D60), Sols-Soils, N°26 Paris.

Courbes hyperboliques contrainte-déformation au pressiomètre Ménard autoforé Stress-Strain Hyperbolic Curves Obtained With a Selfboring Ménard PMT Baud J.-P.

Eurogéo, Avrainville, France

Gambin M. Apagéo, Paris, France Schlosser F.

École des Ponts Paris-Tech, Navier-CERMES, Marne-la-Vallée, France RÉSUMÉ : On présente ici les derniers résultats de nos recherches sur le module de déformation du sol déterminé en assimilant à un arc d’ hyperbole la courbe pression-volume obtenue dans un essai au pressiomètre Ménard. L’essai est réalisé en utilisant une cavité obtenue par un système de tube fendu auto-foré, soit en roto-percussion, le STAF®, soit en rotation seule, le ROTOSTAF®. L’ajustement des points de mesure sur une branche d’hyperbole du second degré permet d’obtenir une expression analytique originale de la déformée  sous la forme  = f (G0, po, pLM, PL) dans laquelle PL représente l’abscisse de la « pression limite vraie » prise comme asymptote. Pour chaque essai, il est alors facile d’obtenir la variation, calculée à partir des points de mesure, du module tangent Gt et du rapport Gt/G0 en fonction de , et pareillement le module sécant Gs. Cette variation peut être comparée avec celle donnée par le modèle hyperbolique courant : la concordance est bonne, sauf pour les très faibles déformations initiales où les courbes obtenues à partir des points d’essais donnent des modules sensiblement plus élevés. Ces résultats sont en bon accord avec les résultats d’essais en laboratoire et en géophysique. Ils mettent en évidence une décroissance typique des modules lorsque la déformation s’accroit.

ABSTRACT : The present stage of our research work on soil E-moduli values are submitted here. These values are obtained by assimilating the pressure-volume plot of a Ménard PMT to a 2nd degree hyperbole arc. The tests were performed using a self-bored steel slotted tube implemented either by the STAF® technique involving a drifter and a full-face bit, or the ROTOSTAF® method with a drag bit protruding from the tube outlet. Getting the hyperbolic best fit of the plotted readings makes it possible to obtain an original equation of the radial borehole expansion  as  = f (G0, po, pLM, PL) in which PL is the true “limit pressure” value of the vertical pressure asymptote. Then, it is easy to derive the tangent modulus Gt for each reading and the corresponding Gt/G0 ratio as a function of ε, and similarly the secant modulus Gs. Their variation can be compared with those given by the usual hyperbolic model: a very good agreement is obtained, except for very small initial strains where the readings plot yields moduli sensibly higher in value. These results are similar to those obtained by laboratory tests and in geophysical surveys. They exemplify the typical decrease of the deformation modulus when the stress or the strain increases. MOTS-CLÉS : Pressiomètre, autoforage, modèle hyperbolique, KEYWORDS: Ménard Pressuremeter, self-boring, hyperbolic soil model. 1 CONDITIONS D’OBTENTION D’UNE EXPANSION PRESSIOMETRIQUE QUASI VIERGE. L’analyse exposée ici est appliquée à des essais pressiométriques réalisés dans les conditions « traditionnelles » de chargement par paliers de l’essai pressiométrique Ménard (NF P91-110-1 et Pr EN-IS0 22476-4), avec des méthodes et matériels de forage et d’essai permettant de réduire autant que possible les effets perturbateurs pour que le sol, avant le départ de l’essai, ne soit ni décomprimé ni comprimé. 1.1. Essais pressiométriques non remaniés : nécessité de l’autoforage. Depuis les années 1970, la mise en place du pressiomètre par autoforage a été utilisé dans les sols mous, suivi par des essais en déformations contrôlées (Baguelin et al., 1978, Mair & Wood, 1987). Progressivement, les essais en auto-forage ont été essentiellement limités pour tester des sols sous-consolidés, supposés dans les conditions idéales de démarrage à po, en vue d’obtenir des relations contrainte-déformation sous cycles de petites déformations, usuellement de moins de 20% d’amplitude. De cette façon, plusieurs modules de sol pouvaient être obtenus, mais aucune pression limite n’était recherchée (Clarke & Gambin, 1998). Plus récemment, une technique d’auto-forage a été proposée (Arsonnet et al., 2005) pour réaliser des essais pressiométriques

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Ménard. Cette méthode, appelée STAF®, consiste à enfoncer un tubage continu, muni d’un tube fendu au voisinage de sa base, à l’aide d’un taillant débordant en roto-percussion à l’extrémité d’un train de tiges centré. Il était ainsi possible d’obtenir un essai en auto-forage de très bonne qualité en petites déformations, en raison de la réduction de la décompression du sol durant la création de la cavité, tout en permettant de réaliser des essais jusqu’à de grandes déformations rendant possible d’atteindre la pression limite conventionnelle (Baud & Gambin, 2005). Avec cette conception, le STAF® fut utilisé avec succès dans les sols cohésifs mous et moyennement compacts, et dans les matériaux granulaires compacts, mais peu d’exemples pouvaient être proposés dans les argiles raides et les sols marneux, la vitesse d’avancement de l’appareillage restant réduit par rapport aux méthodes traditionnelles de pré-forage, telles que la tarière continue ou les outils à dents avec injection de boue. 1.2. Essais autoforés des sols meubles aux roches tendres par la technique Rotostaf. Pour résoudre ce problème du faible avancement du STAF® dans les sols raides ou très compacts, l’utilisation de la rotation simple d’un train de tige a été résolu à l’aide d’une tête de forage hydraulique spécifique combinant une faible vitesse de


rotation du tube extérieur avec une bonne rapidité du vibromarteau agissant sur le taillant débordant. Avec cette conception, appelée ROTOSTAF®, les essais peuvent être réalisés dans des sols tels que les argiles compactes, les marnes et même des couches calcaires. 1.3 Régulation et conduite automatisée d’un essai par le pressiomètre GéoPac® L’usage d’un contrôleur pression-volume (CPV) traditionnel en pression contrôlée est certes possible dans un tube fendu descendu par le STAF®. Mais de meilleurs résultats peuvent être atteints avec un nouveau type de CPV, GeoPac®, intégrant le logiciel de traitement Géovision. Au lieu de travailler au gaz comprimé, ce pressiomètre comprend un piston motorisé extrêmement précis qui permet de réaliser les essais normalisés en paliers de pression, avec une précision volumétrique de 10-3 cm3, soit une précision sur la déformation radiale moyenne de 10-5. Actuellement, dans la mise au point de ce CPV, au moins quatre avancées fondamentales dans la procédure de l’essai ont été recherchées : - une stricte compensation automatique de la résistance propre de la membrane des sondes, ce qui permet une correction de la pression vraie sur la paroi du forage à tout instant - une automatisation complète de la procédure d’essai : le système assisté par ordinateur détermine les modules du sol à partir des trois premiers pas de pression, et ajuste la procédure d’essai de manière à mener à bien l’essai avec un nombre de paliers de pression optimisé. L’opérateur conserve cependant toujours la possibilité de travailler en semi-automatique. - le calibrage automatique du volume initial de la sonde avec le volume réel du trou auto-foré, la pression pour ce premier palier de l’essai étant mis en équilibre avec la pression des terres au repos. Ce procédé dans le système “volume en fonction de la pression” rappelle celui du “lift-off” dans le système « pression en fonction du déplacement » des sondes du pressiomètre auto-foré SBP avec capteurs mécaniques. - enregistrement des lectures (p,V) avec une très grande précision, la courbe ne présentant pas de point d’inflexion. Le logiciel Géovision traite automatiquement les données reçues d’un Géopac, avec sur option de l’utilisateur dérivation des courbes de modules tangent et sécant depuis po comme dans l’exemple donné ci-après. 1.4 Un exemple d’essai Sur l’essai ci-contre (Fig.1), on remarquera que la pression du premier point de lecture est acceptable comme po, avec un volume de contact exactement égal au volume annulaire entre la sonde dilatable et le tube fendu, et que l’essai est régulé jusqu’à un volume de près de 1000 cm3 soit un déplacement de la paroi du forage de l’ordre de 12 mm. L’erreur moyenne répartie entre les points mesurés et le modèle hyperbolique est ici de 1,8 cm3 (soit environ 25µm). D’autres caractéristiques de l’essai sont également remarquables et observées aussi sur de nombreux essais autoforés. La première est que la courbe de fluage présente deux cassures : l’une à la pression pf1 prise ici pour le calcul du module pressiométrique, l’autre à une pression pf2 nettement plus élevée (Monnet et Khlif, 1994). Une autre est que le modèle hyperbolique permet de tracer automatiquement les courbes de décroissance régulière du module sécant Es et de module tangent Et, qui recouvrent bien les mêmes courbes calculées sur les points d’essais, ceux-ci montrant dans les très faibles déformations initiales un module nettement plus élevé, ici presque doublé, que nous chercherons à expliquer.

Figure 1. Résultat d’un essai à 18 m de profondeur dans une argile sableuse raide (Cénomanien de la bordure Ouest du Bassin Parisien), autoforé par Rotostaf, sonde diamètre 44 mm à cellule de mesure de 37 cm dans le tube fendu de 63 mm de diamètre, essai piloté par le CPV GéoPac®. et résultats tracés par Géovision

2. MODÈLE HYPERBOLIQUE DE L’EXPANSION D’UNE CAVITÉ CYLINDRIQUE 2.1. De la double hyperbole au modèle hyperbolique Très tôt dans la pratique de l’essai pressiométrique, Louis Ménard avait défini la pression limite pLM, notion née avec l’essai, au doublement du volume de la cavité de forage initiale, en sachant que cette convention n’impliquait pas que la déformation correspondante soit matériellement atteinte par les sondes. La société Ménard et les concessionnaires pionniers ont dès lors proposé successivement de nombreuses méthodes d’extrapolation de la courbe vers la pression limite : courbe inverse, méthode des volumes relatifs, coordonnées log-log, courbe Lemée, dont la convergence vers une pression limite unique n’était pas évidente (Baguelin et al. 1978). La double hyperbole est l’une de ces méthodes d’extrapolation, dérivant du dessin des essais en (P, 1/V) (d’Hemricourt 2005). La méthode en double hyperbole a été formalisée et programmée (Baud et al., 1992) et constitue un modèle décrivant bien la forme en « S » des essais en préforage, liée aux vicissitudes de diamètre de forage, du temps d’attente et de décompression entre forage et essai et du choix de paliers initiaux inférieurs à la pression des terres po avant forage, par un opérateur qui en principe ignore la valeur de po et de pLM, sujets de la mesure, et à qui il est demandé de prévoir avant la mesure un résultat final divisé en n paliers égaux. Il apparaît rapidement que l’application aux essais autoforés tel que celui de la figure1 simplifie plus ou moins radicalement la modélisation, en réduisant la première hyperbole au rôle de facteur secondaire. Pour un essai débutant par un palier de pression corrigée légèrement supérieure ou idéalement égale à la pression des terres au repos po sans décompression ni refoulement du sol avant l’essai, les points d’essai décrivent une simple hyperbole, de la forme A3 V  A  A2 . p  1 A4 - p (1) où V est le volume mesuré au-delà du volume de la sonde au repos Vo et p la pression d’essai corrigée de l’étalonnage et du calibrage. Pour le pressiomètre Ménard l’usage est de nommer le V mesuré par la simple notation « V ». Il est possible d’exprimer ces données d’essai en fonction de la déformation dite circonférentielle c :

c

492



aa a

0

0 

V p  V V P   V0

1

(2)


où a est le rayon du forage en expansion, a0 le rayon initial au moment du contact sonde-sol à la pression po, Vp le volume de la sonde au repos, V0 volume de contact entre la sonde et le sol vierge correspondant à a0. L’essai pressiométrique ne mesurant qu’un déplacement à la paroi de la cavité, c est la seule déformation relative qui peut être ainsi déduite de l’essai et dans la suite du texte nous désignerons simplement - la déformation pressiométrique à la paroi par , sans indice ; - les pressions nettes à la paroi, après déduction de la pression horizontal du sol au repos p* = p – p o L’hyperbole ainsi ajustée sur les données d’essai est de la forme   C1  C 2 . p

*



C3 C4 - p *

(3) Nous rappelons rapidement les étapes conduisant de cet ajustement mathématique sur les points de mesure aux paramètres du modèle hyperbolique présenté ici : C1 est homogène à une déformation, C2 à une unité de contrainte ayant le rôle la dimension d’un module, C3 et C4 à des contraintes (pressions). On montre facilement (Baud & Gambin, 2005, 2008) que C4 est l’ordonnée de l’asymptote verticale pour  = , notée pL par les Anglo-Saxons, et ici p*L, et que les 3 autres paramètres ne sont pas indépendants et se réduisent à 2 : une déformation 0 et un module E0 qui est la pente de l’asymptote oblique :  ε0 

-

p* ε  p*L  0* E0 pL - p*

(4)

Le module de cisaillement du sol pour les déformations infinitésimales à partir de po, noté G0 est un des paramètres du modèle (Baud et al., 2012), comme on peut le voir en construisant hors de toute référence à une base expérimentale une hyperbole passant par le point (po, 0), ayant une pente 2G0 à ce point initial et admettant une asymptote verticale p*L :

ε 

p * 1  p * pc* . 2.G0 1  p * p *L

(5)

Cette expression nécessite pour être déterminée de connaître un point quelconque de la courbe (px, ex) définissant comme paramètre complémentaire une pression nette p*c telle que :

p c*  -



p L* . p *x

2



2 . x .G 0 p L*  p *x  p L* . p *x

En complément de G0 et p*L, un seul autre paramètre est donc nécessaire à la définition complète du modèle, soit pc, soit 0 :Erreur ! Signet non défini.

 1  0 . p*    0*  p *  * L *  pL  p  2.G 0 p L 

   0  -

(6)

(7)

Les expressions (5) et (7) sont équivalentes dès que l’on détermine le même point complémentaire sur la courbe.

2.2. Rôle de la limite conventionnelle de l’essai Le choix d’une valeur de rupture conventionnelle pour l’essai, p*LM, a été dicté par la nécessité pragmatique de déduire de cette caractéristique globale du sol au niveau de l’essai des règles de dimensionnement à la rupture réalistes, et indépendantes de la recherche d’une pression limite « vraie » p*L correspondant à une déformation infinie. On peut remarquer qu’elle signifie, pour les essais en forage calibrés de diamètre 60 à 63 mm (2 pouces ½) qui sont devenus la pratique et la norme de l’essai, un déplacement absolu de la paroi de 13 mm environ (½ pouce, ou 20% du diamètre).

493

L’expérience de l’utilisation de sondes de diamètres très différents démontre que cette convention n’est équivalente au doublement du volume de la cavité que par le hasard historique du choix pa r Ménard de sondes de 55 à 63 mm, comme les plus pratiques et les plus répandues (en réalité, jusqu’à 76 mm ou 3 pouces si on considère la pratique du pressiomètre Ménard au niveau mondial). On constate que des sondes de diamètres très différents conduisent à une pression limite équivalente à p*LM pour des taux d’expansion qui ne correspondent pas au doublement de volume de leur cavité, mais à un même déplacement absolu de la paroi du forage de l’ordre de 13 mm : Dans les petits diamètres, les sondes de diamètres 22 mm et 32 mm dites « minipressiomètre », de volume 240 cm3 au repos, restent dans le domaine pseudoélastique ou proche de la pression de fluage lorsqu’elles ont doublé de volume, et nécessitent une expansion jusqu’à 350 à 400 cm3 pour montrer une rupture franche du sol ; ce volume, atteint sans difficulté grâce à l’élancement important de ces sondes, correspond à un déplacement de la paroi de 11 à 13 mm. Dans les diamètres plus importants, les utilisateurs des sondes autoforeuses de type PAF76, de diamètre 140 mm, ont montré que la rupture était amorcée dès le début de l’essai, et ont fixé expérimentalement une équivalence avec la pression limite Ménard pour une pression p20 déterminée par une déformation diamétrale de 20%, soit un déplacement absolu de 14 mm. (Baguelin et al. 1978). Dès lors, nous proposons d’utiliser comme convention pour le calcul de la pression limite pLM pour tout essai de chargement radial la déformation conventionnelle  = r/Rref et non plus  = r/r0, ce qui rend  dépendant d’une longueur absolue Rref=13 mm dont la signification reste à rechercher, mais indépendant de la sonde utilisée et du diamètre du forage. 2.3. Expression de la pression limite conventionnelle Le modèle de comportement de sol hyperbolique de type élasto-plastique avec écrouissage dit « Hardening soil model » de Plaxis, B.V. est bien connu. Ce modèle utilise dans le repère (1, q) où 1 est la déformation axiale des essais triaxiaux, et le déviateur q = 1 - 3, une courbe hyperbolique passant par l’origine et d’asymptote horizontale qa. Par analogie avec ce modèle définissant un module E50 correspondant à la déformation acquise pour la moitié du déviateur de contrainte de rupture, soit qf, nous définissons sur la courbe pressiométrique le module de cisaillement sécant GM atteint à la moitié de la pression limite p*L ( 1 ). Soit : p* (8) GM  L 4. M En portant cette valeur M dans (7), on obtient p*  1 1  (9)   ε 0  L  2  G0 G M  D’où une expression du modèle pressiométrique : *2 *   1 1  p (10)  p  *   1   *  p  p  2.G 2 GM G 0  0   L  Pour obtenir la présentation de l’expression (10), nous avons choisi un module à la moitié de la rupture « vraie », asymptotique, et non la moitié de la rupture conventionnelle. Ce second choix aboutit également à une expression  = ƒ(p) déterminée par 3 paramètres physiques (G0, p*LM et un module sécant G’M), mais moins simple. En pratique, ces deux modules GM et G’M sont évidemment très peu différents, puisque l’on



1



N.B. Les modèles hyperboliques « hardening soil » et pressiométrique n’étant pas dans les mêmes coordonnées de contraintes et déformations, il ne s’agit ici que d’une analogie.


constate sur les essais autoforés que le rapport p*LM/p*L, analogue au rapport Rf = qf/qa, est de l’ordre de 0,7 à 0,9. La pression limite conventionnelle p*LM est celle qui correspond à  = 1, ce qui conduit pour p*LM à une simple équation du second degré : *   1 1  * 2  p L   p LM    1  p *LM  p *L  0    2.GM G0   2.G0 

(11)

dont p*LM est l’unique solution positive.

2.4. Expression des modules réduits G/G0.

Figure 2. Module tangent réduit Gt/G0 en fonction de la déformation pressiométrique radiale pour la gamme des valeurs du rapport EM/p*LM.

Les expressions (10) et (11) permettent de calculer les modules de cisaillement tangent Gt et sécant Gs, normalisés par leur rapport au module tangent initial G0, et de visualiser l’évolution de ces modules en fonction de la déformation, pour la gamme des valeurs possibles pour les rapports G0/p*LM ou p*L/p*LM, rapports caractéristiques du type de comportement du sol. En figure 2 est présentée la même évolution du module tangent réduit Gt/G0 en fonction des valeurs de EM/p*LM, plus familières aux utilisateurs du pressiomètre. 3. CONCLUSION La possibilité existe avec la méthode du tube fendu autoforé (forage STAF®) et l’utilisation d’un pressiomètre qui pilote et régule l’essai automatiquement (pressiomètre Géopac®), de réaliser dans des sols variés et offrant la plus large gamme de résistances, des essais pressiométriques autoforés, fournissant instantanément (logiciel industriel Géovision®) non seulement la courbe pressiométrique "classique", mais aussi ses dérivées en module tangent et sécant depuis une origine po. Elle ouvre des perspectives de développements nouveaux dont nous n’avons pu donner ici qu’un aperçu. En particulier nous pensons possible de traduire très facilement les paramètres des courbes d’essais nécessaires au dimensionnement direct de fondations par la méthode de J.-L Briaud (2003, 2006).

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4. RÉFÉRENCES. AFNOR, 2012, Pr EN ISO 22476-4,. Reconnaissance et essais géotechniques Essais en place - Partie 4: Essai au pressiomètre Ménard. Arsonnet, G., Baud, J.-P., Gambin, M. P. 2005. Réalisation du forage pour essais pressiométriques par un système de tube fendu autoforé (STAF), in ISP5 – PRESSIO 2005, Actes Symp. Intern. Paris, Gambin, M. P., Magnan, J.-P., Mestat, P. (eds) 22-24 août, 2005, Paris: Presses des Ponts. Vol.1 pp 31-45.. Baguelin F., Jezequel J.F., Shields D.H. 1978. The Pressuremeter and Foundation Engineering. Trans Tech Publications, Clausthal, Germany, pp 425-439. Baud, J.-P, Gambin, M. P. 2005. Déduction d’une loi de réponse hyperbolique unique par compilation de courbes pressiométriques dans un sol de lithologie homogène, in ISP5 – PRESSIO 2005, Actes Symp. Intern., Gambin, M. P., Magnan, J.-P., Mestat, P. (eds) Paris, 22-24 août, 2005, Vol.1 Paris: Presses des Ponts pp 175-186. Baud, J.-P, Gambin, M. P. 2008. Homogenising MPM Tests Curves by Using a Hyperbolic Model, in Huang, A.-B., & Mayne, P. W. (eds) Geotechnical and Geophysical Site Characterization, Proc. ISC’3 Taiwan, 1-4 April 2008, London: Taylor & Francis Baud, J.-P, Gambin, M. P., Schlosser F. 2012. Stress-strain Hyperbolic Curves with Ménard PMTs, in R.Q. Coutinho (ed.), 4th Int. Conf. on Geotechnical and Geophysical Site Characterization (ISC'4), P. de Galinhas, Brazil, 18-21 sept. 2012. London: Taylor & Francis Baud J.-P., Gambin M., Uprichard S. 1992. Modeling and automatic analysis of a Ménard pressuremeter test. Géotechnique et Informatique, Presses des Ponts, Paris.pp 25-32. Briaud, J.-L., Hossein K. et BarfknechtJ, .2003. Méthode de détermination de la courbe charge-tassemennt pour les fondations superficielles dans les sables. Presses des Ponts, Rev. Fr. de Géotechnique N°105, pp 29-39. Briaud, J.-L., 2006. The preboring pressuremeter, some contributions, in ISP5 – PRESSIO 2005, Actes Symp. Intern., Gambin, M. P., Magnan, J.-P., Mestat, P. (eds) Paris, 22-24 août, 2005, Vol.2 Paris: Presses des Ponts pp 103-124. Clarke B. G., Gambin, M. 1998 Pressuremeter Testing in Onshore Ground Investigations: A report by the ISSMGE Committee TC16, Geotechnical Site Investigation, Proc. First Int. Conf. on Site Characterization (ISC’1), P. K. Roberts on & P. W. Mayne eds., Vol.2, A. A. Balkema, Rotterdam, pp 1429-1468. d’Hemricourt J. 2005. L’interprétation de l’essai pressiométrique : de la courbe inverse à la double hyperbole. in ISP5 – PRESSIO 2005, Actes Symp. Intern. Paris, Gambin, M. P., Magnan, J.-P., Mestat, P. (eds) 22-24, 2005, Paris: Presses des Ponts. Vol.1 pp 319-328. Mair, R. J., Wood, D. M. 1987. Pressuremeter Testing, CIRIA Series, London: Butterworths Monnet J. et Khlif J. 1994 Etude théorique de l’équilibre élastoplastique d’un sol pulvérulent autour du pressiomètre. Presses des Ponts, Rev. Fr. de Géotechnique N°65.

Quality control of Cutter Soil Mixing (CSM) technology – a case study Contrôle de la qualité des la technologie Cutter Soil Mixing (CSM) – une étude de cas Bellato D., Simonini P.

University of Padua - Department of Civil, Environmental and Architectural Engineering

Grisolia M., Leder E., Marzano I.P.

Sapienza University of Rome - Department of Civil, Environmental and Architectural Engineering

ABSTRACT: The Cutter Soil Mixing (CSM) is a relatively new Deep Mixing (DM) method that offers versatile construction solutions suitable for various types of ground improvement. Besides the many advantages compared to the most common DM methods, CSM has a high level of process control. Quality control and quality assurance (QC/QA) procedures are essential aspects of each DM project, and a successful treatment is related closely to the professional ability to control and verify the DM construction. This paper presents the results of laboratory tests carried out on wet grab samples collected from a CSM construction site characterized by the presence of sandy soil. Similar soil-binder mixture were then produced and tested in the laboratory accordingly, using the same binder adopted for the in situ panel construction and the sandy soil taken directly from the jobsite. A comparison between the results obtained by UC tests carried out on the wet grab and the laboratory mixed samples is also presented. The results obtained using an innovative experimental apparatus underline the influence of the physical and chemical characteristics of the natural soil on the strength gain of the stabilized material. RÉSUMÉ : Le Cutter Soil Mixing (CSM), appartenant à des méthodes Deep Mixing, est une technique récente qui offre des solutions constructives adaptés à différents types d'amélioration du sol. En plus des nombreux avantages sur les méthodes les plus courantes, le CSM a un niveau élevé de contrôle de processus. Les procédures de contrôle et d'assurance de la qualité (QC/QA) sont des aspects essentiels du projet, et le succès du traitement est étroitement liée à la capacité de contrôler la phase d'exécution. Cet article présente les résultats de tests de laboratoire effectués sur des échantillons prélevés “wet grab” d'un site à CSM caractérisé par la présence d'un sol sableux. Semblables sol-liant mélanges ont ensuite été fabriqués et testés dans le laboratoire avec le même liant adopté pour la construction des panneaux in situ et le sol sablonneux prises directement à partir du site. En particulier, il est présenté une comparaison entre les résultats obtenus avec des essais de compression simple, effectuée sur des échantillons prélevés “wet grab” sur le site et éprouvettes réalisés en laboratoire. Les résultats obtenus par l'utilisation d'un appareil expérimental innovateur ont souligné l'influence des caractéristiques physiques et chimiques du sol naturel sur l'augmentation de la résistance du matériau stabilisé. KEYWORDS: deep mixing, cutter soil mixing, sandy soil, unconfined compressive strength. 1

INTRODUCTION

The Cutter Soil Mixing (CSM) offers numerous advantages over the more traditional methods of mixing soils using standard rotary tools (Fiorotto et al. 2005), being equipped with two sets of cutting wheels rotating around horizontal axes producing treated soil panels of rectangular shape. Several successful applications in different geotechnical contexts for various engineering purposes have been recently documented by Gerressen and Vohs (2012). The Quality Control/Quality Assurance (QC/QA) programs have the objective to ensure the compliance between the actual field performance and the design requirements, therefore special attention is required. Due to the significant uncertainties related to the site activity, most of the mix design and mixing procedure calibration is performed in the laboratory. In order to develop a tool for an effective comparison between laboratory and field values, a specific CSM jobsite located in the city of Zandvoort (NED) has been selected. The subsoil condition is characterized by the presence of sandy soil. Despite the fact that higher performance are usually obtained in the laboratory (Porbaha et al. 2000), the comparison between strength tests on wet-grab samples and laboratory specimens have shown sometimes opposite outcomes (Bellato et al. 2012). The mechanical properties of in-situ improved soil may be found larger than that of laboratory specimen when using cement slurry (wet method) to stabilize loose sandy ground due to water drainage (Yoshimura et al., 2009).

495

Three types of water drainage may occur during soil mixing operations: potential expulsion of part of pore-water contained in the original soil by the injection of the cement slurry; bleeding of the soil-binder mixture, i.e. drainage of water due to sedimentation processes; possible drainage towards the surrounding soil layer of part of the water in the mixture due to consolidation under the effective overburden pressure. In this paper the effect of water drainage was investigated trough an original laboratory experimental apparatus. To assess the influence of the granular soil type on test results, the analysis were replicated on a different marine sand. Moreover, important considerations regarding the significant influence of the physical and chemical characteristics of the natural soil on the strength gain of the stabilized sands are presented and discussed. 2

SITE DESCRIPTION

A requalification activity was planned in Zandvoort, a small village next to the North Sea coast at about 30 km west of Amsterdam. Preliminary geotechnical ground investigations were performed in the jobsite area. The results show a relatively uniform sand profile characterized by the prevalence of a medium to fine sand, generally of medium density, whose grain size distribution is reported in Figure 1. The groundwater level ranges around 2.5 m below the ground surface.


Figure 1. Grain size distribution of the Zandvoort and Serapo sands .

To construct the 11.5 m deep CSM panels (2.4 x 0.55 = 1.32 m2 sectional area) the 1-Phase system was chosen, therefore the grout was injected on both downward and upward stroke. The grout composition adopted for the panel production was characterized by a water-to-cement ratio w/c = 1.12 and a binder factor α = 509 kg/m3 of natural soil. The cement used was a special composite cement especially produced for ground improvement applications. After mixing, several wet-grab samples were collected from the fresh panels at about 2,0 m from the ground level and immediately sealed into watertight tins (inner diameter of 98 mm and height of 113 mm). 3

MATERIALS AND TESTING PROGRAMME

The wet-grab samples collected from the site (in the following referred as “SWGS”) were cured under controlled condition (room temperature of about 20°C and at a relative humidity > 95%.) for 40 and 125 days in order to measure also the time influence on the unconfined compressive strength (UCS) of the treated soil. Before testing, the wet-grab samples were cored to provide specimens of 37 and 54 mm in diameter with an aspect ratio of 2. Finally, the specimens were trimmed to regularize the bases and wrapped with plastic film to prevent moisture loss. Laboratory soil-binder mixtures were prepared at the same grout/sand ratio used on site, according to the treatment parameters evaluated from the elaborated machine production data. The grout and the soil were first prepared separately and then mixed together for 10 minutes using a high power mixer to produce the stabilized soil, according to the recommendations for laboratory mixed specimens provided by the Japanese Geotechnical Society (JGS0821-2000). The stabilized soil was then poured into plastic moulds 50 mm in diameter and 100 mm in height using the No Compaction technique (simply consisted in filling the mold) to realize the laboratory mixed specimens (referred as “LS”). Past experiences of sandy soil stabilization (Yoshimura et al, 2009, Grisolia et al, 2010, Bellato et al., 2012) showed the following occurrences related to water drainage conditions:  The physical properties (water content and wet density) of sandy soil collected from the site, especially when taken below the groundwater table, typically are different from the initial in-situ conditions, due to the loss of fine particles and water during sampling and transportation to the laboratory;  Bleeding, i.e. separation of water from the soil-binder mixture, generally occurs immediately after the mixing process in the bowl and causes the sedimentation of some amount of cement at the surface;  Every molded sample usually shows the occurrence of bleeding phenomena, that inevitably leads to a reduction in the specimen’s height;  In addition, when the mixture is taken from the bowl for molding operations, separation among constituent materials

496

may be observed. This further increases the variability in terms of amounts of binder, water and sand of the samples.  Moreover, during in-situ soil treatments, some water drainage may also arise depending on the type of mixing procedure adopted and the specific subsoil conditions. In particular, sedimentation mechanisms in the liquid soil-binder slurry mixture may develop just after the passage of the mixing tools and some amount of water can be radially drained away into the surrounding permeable sandy layers (Yoshimura et al, 2009). To simulate the effects of water drainage on the mechanical properties of stabilized soils in the laboratory an original experimental set up was designed and used (Figure 2). The apparatus was essentially composed of a watertight container in which a cylindrical sand core, reproducing the site conditions, is placed and surrounded by a gravel filter, with installed an open pipe for water level control (Figure 2a). A cylindrical cavity was then prepared and filled with the stabilized soil just after the mixing operations (Figure 2b). After a time span equal to that adopted on site before sampling, a laboratory wet-grab specimen was retrieved (Figure 2c and 2d). The two types of specimens, i.e. laboratory (“LS”) and laboratory wet-grab (“LWGS”), were cured at 20°C and at 95% relative humidity in curing tanks and removed from the moulds just before the test. In order to investigate the influence of the sand type and mineralogy on the performance of the stabilized material, a marine soil namely Serapo Sand (Figure 1) was also used to prepare laboratory and laboratory wet-grab specimens. The experimental investigation mainly consisted of unconfined compression tests. The specimens were tested at different curing times, ranging from 7 to 125 days. To evaluate the influence of the physical and chemical characteristics of the natural soils (Zandvoort and Serapo sands) SEM (Scanning Electron Microscope) and EDS (Energy Dispersive Spectroscopy) analysis were carried out. A CamScan MX2500 electron microscope, equipped with a EDAX EDS (energy dispersive X-ray spectrometer) system was used to determine both the morphology and chemical composition of the grains. Two small samples for both sands were first oven dried at 40°C for 24 h and then coated with a layer of carbon using an high-vacuum evaporative coater to prevent the accumulation of electrostatic charges at the surface during irradiation. 4

RESULTS AND DISCUSSION

The results of the unconfined compression tests performed on the three series of samples (SWGS, LS, LWGS) are presented in

Figure 2. Experimental set-up for laboratory wet grab specimens: a) cavity preparation, b) mixture pouring, c,d) specimen retrieval.


Unconfined compressive strength, UCS (Mpa)

25 SWGS Zandvoort LS Zandvoort 20

LWGS Zandvoort

15

10

5

0

0

25

50 75 100 Curing time, tcur (days)

125

150

Figure test results onFourier Zandvoort sand specimens Figure 3. 7. UC Lower order Descriptors obtained from the two types of sands.

Figure 3. From this figure it clearly appears that higher strength was provided by the tests carried out on the SWGS, for which UCS has been found to range between 12 and 16 MPa at 40 curing days. UCS obtained from the LS is lower at any curing time investigated. In particular the UCS was found to be about 7.0 MPa at 40 curing days. From the same figure, it also appears that UCS of LWGS approaches the field values. These results underlines the effectiveness of the experimental set up in simulating the real field conditions, and emphasizes the significant effect of drainage conditions, which increase the UCS of about 1,9 times at 40 curing day. To evaluate the influence of the type of sand, and, therefore, of the related drainage effect on strength properties, the same experimental procedure for sample preparation was replicated on Serapo sand. The results of Figure 4 confirm also for this kind of sand an increment, even though less significant, of the UCS due to the drainage effect. The increment was about 40% at 40 curing days for the LWGS specimens with respect to the classical LS. It is important to note (Figure 4) that similar UCS at 40 curing days was obtained from the LS of both Zandvoort and Serapo sands (prepared according to JGS0821-2000). This was expected since the two sands presents similar grain size

Unconfined compressive strength, UCS (Mpa)

25

LS Zandvoort

distributions. The results obtained from the newly developed experimental apparatus show that the type of sand and the corresponding water drainage effect may greatly influence the mechanical properties of the stabilized sandy soils. To investigate in more details the reason of this particular outcome, mineralogical and microstructural tests were performed on the two types of sand. The SEM and EDS analyses results are shown in Figure 5 and 6. Figure 5 presents two backscattered electron (BE) images of two different sand grains: the grain on the left referring to Zandvoort sand, whereas that on the right to Serapo sand. Generally, both sands are predominantly composed of quartz minerals, but in the Serapo sand a significant portion of carbonate particles is present (Figure 6). In addition, it is easily detectable the more irregular and angular morphology of quartz grains of Zandvoort sand with respect to the more rounded, sub-angular carbonate grain of Serapo sand. To quantify the degree of angularity different methods have been proposed in the literature (de Santiago et al., 2008). Among them, the procedure based on the Fourier descriptors (Bowman et al, 2001) is one of the most diffuse recent approaches. The boundary of the particle is circumnavigated in the complex plane at a constant speed. The step size is selected so that the circumnavigation takes 2π and the number of steps is 2k. The complex function presented in Eq. (1) allows to determine the aforementioned Fourier descriptors N /2   i 2 n m  xm  i y m  Z n exp    M    n  N / 2 1

(1)

where x, y are the coordinates of the particle boundary, N is the

Figure 5. Comparison between SE images of a Zandvoort (on the left) and Serapo sand grain (on the right)

LWGS Zandvoort LS Serapo

20

LWGS Serapo

15

10

5

0

0

25

50 75 100 Curing time, tcur (days)

125

150

Figure 6. Chemical compositions from EDS analysis performed on a Zandvoort (on the left) and a Serapo sand grains (on the right).

Figure 4. UC test results on Zandvoort and Serapo specimens

497


total number of descriptors, n is the descriptor number, M is the total number of points describing the boundary, m is the index number of a point on the boundary, Zn is the Fourier descriptor and i is the imaginary number. Each Fourier descriptor, especially those of the lower order, are associated to specific and morphological features of the particle shape. The average shape descriptors obtained for a reasonable number of grains taken from each sand sample are shown in Figure 7. A clear more unevenness in the boundary of the Zandvoort grains can be recognized due to the higher contribution of higher order Fourier descriptors to the shape morphology. 5

CONCLUSIONS

The calibration of relationships between real and laboratory scale treatment may support soil mixing QC/QA procedures. In sandy soil, laboratory specimens tests results may be lower than that obtained by wet grab samples due to water loss during in situ mixing operations. The results show that quite a good match may be achieved by simulating in laboratory the in situ water drainage. The results obtained from the newly developed experimental apparatus show that the kind of sand may greatly influence the water drainage effect on the mechanical properties of the stabilized soil. The different degree of angularity of the grains and the different nature of the minerals composing the two sands considered in this study should be considered as relevant factors affecting the performance of the stabilized soil, as well as the grain size distribution. Further study are needed to validate and extend the results and findings described in this case history. To simulate in situ condition it is also necessary to carefully take into account other possible factors such as: mixing energy, use of compressed air, molding technique and curing conditions. 6

AKNOWLEDGEMENT

The authors would like to thank the Hoffman Group and the Bauer Group, especially Mr. Franz Werner Gerressen & Mr. Thomas Vohs for kindly providing the construction data for this study. The authors wish to thank Mr. Fabrizio Tocci for his help in conducting the laboratory test at the Department of Civil, Environmental and Architectural Engineering of Sapienza University of Rome. 7

REFERENCES

Bellato D., Simonini P., Marzano I.P., Leder E., Grisolia M.,Vohs T., Gerrresen F.W., 2012. Mechanical and physical properties of a CSM cut-off/retaining wall. International Conference on Ground Improvement and Ground Control (ICGI 2012), University of Wollongong, Australia. ISBN 978-981-07-3561-6. Bowman, E.T., Soga, K., and Drummond, T.W., 2001. Particle shape characterization using Fourier descriptor analysis. Géotechnique, 51 (6), pp. 545-554. De Santiago, C., Santana, M., and Manzanas, J., 2008. Digital Image processing and Fourier descriptors analysis of the porosity in various volcanic rocks. Proc. Of the International Geotechal Conference - Development of urban areas and geotechnical engineering, S. Petersburg (Russia), Vol. 2, pp. 449-454. EN 14679, 2005. Execution of special geotechnical works - Deep mixing, CEN - Technical Committee CEN/TC 288. Fiorotto, R., Schöpf, M., and Stötzer,E., 2005. Cutter Soil Mixing (CSM) - An innovation in Soil mixing for creating Cut-off and Retaining walls. Proceedings 16th ICSMGE, 15 sept. 2005, Osaka (Japan), pp. 1185-1188. Gerressen, F.-W. and Vohs, T., 2012. CSM - Cutter Soil Mixing Worldwide experiences of a young soil mixing method in soft soils.

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Proceedings 4th International Conference on Grouting and Deep Mixing, New Orleans (USA). Grisolia M., Marzano I.P., De Lentinis D., Leder E. 2010. “Performance of CSM stabilised soils in geotechnically complex formations”. Proc. Geotechnical Challenges in Megacities, Moscow (Russia). ISBN 978-5-9902005-2-4. Grisolia M., Kitazume, M., Leder E., Marzano I.P., Morikawa Y. 2012. Laboratory study on the applicability of molding procedures for the preparation of cement stabilised specimens” International Symposium & short courses on Recent Research, Advances & Execution Aspects of ground improvement works, Brussels. JGS 0821-00 2000. Practice for Making and Curing Stabilised Soil Specimens Without Compaction (Translated version). Geotechnical Test Procedure and Commentary, Japanese Geotechnical Society. Kitazume, M., 2005. State of Practice Report: Field and laboratory investigation, properties of binders and stabilised soils. Procceedings International Conference on Deep Mixing – Best Practice and Recent Advances. Swedish Deep Stabilization Research Centre, Stockholm (Sweden). Vol. 2, pp. 660-684. Larsson, S., 2001. Binder distribution in lime-cement columns. Proceedings of the ICE - Ground Improvement. Vol. 5, No. 3, pp. 111-122. Marzano I.P., Leder E., Grisolia M., Danisi C. 2012. Laboratory study on the molding techniques for QC/QA process of a Deep Mixing work. 3rd International Conference on New Developments in Soil Mechanics and Geotechnical Engineering, Near East University, Nicosia, North Cyprus. ISBN 975-8359-28-2. Porbaha, A., Shibuya, S., and Kishida, T., 2000. State of the art in deep mixing technology. Part 3: geomaterial charachterization. Proceedings of the ICE - Ground Improvement. 4 (3), pp. 91-110. Yoshimura K., Mochizuki H., Kon N., Saito S., Suzuki Y., Sugiyama T., Takahashi, S., 2009. New Procedure for Making Specimens in Laboratory Mix Test for Sandy Soil Improved by Cement Slurry. International Symposium on Deep Mixing & Admixture Stabilization, Okinawa, Japan.

Mesures dynamiques lors du battage pénétromètrique – Détermination de la courbe charge-enfoncement dynamique en pointe Dynamic measurements of the penetration test – Determination of the tip’s dynamic loadpenetration curve Benz M.A., Escobar E., Gourvès R., Haddani Y. Sol-Solution Géotechnique Réseaux, Riom, France

Breul P., Bacconnet C.

Institut Pascal-Polytech’Clermont-Ferrand Université Blaise Pascal, Clermont-Ferrand, France RÉSUMÉ : Dans cet article, nous présentons les résultats des récents développements réalisés sur l’essai de pénétration Panda 3® en vue de permettre d’enrichir son exploitation. A partir de la mesure et du découplage des ondes créées suite à l’impact sur l’appareil, nous pouvons obtenir pour chaque coup la courbe charge enfoncement p-sp permettant de déterminer des paramètres de résistance et de déformation mis en jeu pendant l’enfoncement de la pointe. Une campagne d’essais au laboratoire dans une chambre de calibration pour deux sols a été menée afin de valider les résultats obtenus par l’exploitation de la courbe p-sp. Les résultats obtenus sont confrontés aux résultats obtenus à l’aide d’essais triaxiaux et œdométriques. ABSTRACT: In this paper, we present the results of the recent developments done on the Panda 3® dynamic penetrometer aiming at improving its use will be presented. From measurement and decoupling of waves created by the impact on the penetrometer, we can obtain for each blow the load-settlement p-sp curve allowing determination of the strength and deformation parameters brought into play during the cone penetration. A series of tests in a calibration chamber for two soils were conducted to validate the results obtained by the exploitation of the p-sp curve. The obtained results are compared with the results obtained using the triaxial and oedometer test. MOTS-CLÉS : caractérisation des sols, pénétromètre dynamique, Panda 3®, propagation d’ondes, courbe charge-enfoncement. KEYWORDS : soil characterization, dynamic penetrometer, Panda 3®, wave propagation, load-penetration curve. 1

INTRODUCTION

En reconnaissance de sols, du fait des contraintes des essais de laboratoire (coût, échantillonnage, transport…), l’utilisation d’essais in-situ est une pratique très répandue. Parmi ceux-ci, les pénétromètres dynamiques sont les plus utilisés dans le monde et sont intéressants pour l’étude du comportement dynamique des sols (Tokimatsu 1988) et ce bien que pour la plupart des ingénieurs, leur caractère dynamique soit considéré comme un désavantage. Toutefois, les pénétromètres dynamiques ne permettent d’obtenir qu’une seule information sur le sol : la résistance de pointe ; et au contraire des pénétromètres statiques qui sont devenus des outils très sophistiqués grâce à l’incorporation de différents capteurs dans les pointes, les pénétromètres dynamiques sont restés éloignés de ces avancées et demeurent d’une technicité ancienne. Par ailleurs, le battage pénétromètrique a longtemps été expliqué par la théorie des chocs de Newton, bien que l’on sache que ce problème ne peut être résolu avec la seule application de cette théorie. A l’heure actuelle, on sait que le battage pénétromètrique est mieux représenté par la théorie de transport des ondes où le transfert d’énergie se fait sous forme d’une onde de compression qui parcourt le pénétromètre après chaque impact (Smith 1962, Aussedat 1970). C’est dans ce cadre et sur la base du pénétromètre PANDA® développé depuis plus de vingt ans (Gourvès 1995) que nous avons conçu et développé un pénétromètre et un procédé de mesure permettant d’améliorer l’information obtenue lors d’un sondage : Le PANDA 3® (Benz et al. 2010). Le principe consiste à mesurer et à découpler les ondes crées par l’impact du marteau sur la tête de l’appareil et à calculer ensuite la force, l’accélération et la vitesse subis en pointe pour nous permettre de tracer la courbe charge-enfoncement p-sp pour chaque coup fourni lors du battage (figure 1).

Figure 1. Principe de l’essai Panda 3® (c.f. Benz, 2009).

L’exploitation de la courbe permet de déterminer des paramètres de résistance et de déformation du sol mis en jeu lors de la pénétration de la pointe tels que la célérité des ondes, le module pénétromètrique et l’amortissement de Smith. 2

PRINCIPE DU PANDA 3®

Le principe de l’essai est simple : au cours du battage on vient mesurer dans les tiges, au voisinage de l’enclume, les variations de déformation ε(x,t) et/ou d’accélération a(x,t) entraînées par l’onde de compression créée par l’impact. En effet, quand le marteau de masse M animé d’une vitesse vm heurte la tête du pénétromètre, une onde de compression u(x,t) est engendrée dans celui-ci et se propage à une vitesse constante ct vers le cône. Lorsque u(x,t) arrive à l’interface cône/sol, une partie de celle-ci est utilisée pour déformer le sol et une autre partie est réfléchie vers le haut. La propagation de u(x,t) dans les tiges est décrite par la équation (1) et sa solution générale correspond à la superposition de deux ondes, ud et ur, descendante et remontante (équation 2). Lors de son parcours u(x,t) entraîne dans tout point x des tiges des variations de déformation ε(x,t) et

499


de vitesse particulaire v(x,t) représentées par la superposition des ondes élémentaires. 2  2u ( x, t ) 2  u ( x, t )  ct 2 t x 2

(1)

 u ( x, t ) u d (t  x / c t )  u r (t  x / c t )

(2)

  ( x, t )  d (t  x / c t )   r (t  x / c t )

(3)

 v ( x , t ) v d (t  x / c t )  v r (t  x / c t )

(4)

Il est possible de montrer que l’expression (4) peut être exprimée en fonction des ondes de déformation εd et εr.  v ( x, t ) ct  d (t  x / ct )   r (t  x / ct )

(5)

La connaissance des ondes εd(t) et εr(t) permet ainsi de décrire entièrement le phénomène dynamique du battage pénétromètrique en tout point x le long des tiges. Dans la pratique, cette connaissance peut se faire à l’aide de mesures réalisées par le biais de jauges de déformation et/ou d’accéléromètres. Toutefois, dans les enregistrements réalisés lors du battage, ces ondes se trouvent souvent imbriquées les unes avec les autres et il devient nécessaire de les découpler. 2.1

Découplage d’ondes et construction de la courbe chargeenfoncement en pointe

Figure 2. Exemple de courbes charge-enfoncement pénétromètriques obtenues avec le Panda 3 pour deux types de sol (c.f. Benz 2009).

Pour chaque impact, on calcule des paramètres de résistance, d’amortissement, de déformation ainsi que de célérité d’ondes dans le sol. 2.2.1 Paramètres de résistance En supposant que la contrainte en pointe qd(t) est la résultante des composantes statiques Rs (obéissant à une loi élastoplastique parfaite) et dynamiques Rd(t) (proportionnelle à la vitesse d’enfoncement vp(t)); on détermine la valeur de Rs en admettant que lorsque vp(t) est nulle la composante dynamique Rd(t) s’annule et Rs est donc égale à qd(t).

Différentes méthodes peuvent être employées pour découpler les ondes εd(t) et εr(t) à partir des enregistrements réalisés. Celles-ci diffèrent suivant le type de mesures (déformation, accélération…), suivant la quantité (1… n) ainsi que suivants les conditions aux limites imposées. Toutefois, il a été montré que la méthode proposée par (Casem et al. 2003) est celle qui s’adapte le mieux au cas du battage pénétromètrique. A partir des enregistrements εA(t) et vA(t) réalisés dans un point A, les ondes εd(t) et εr(t) sont découplées d’après : v A (t )  v A (t )  1 1 et  r (t )  A (t )    A (t )   2 2 ct  ct 

  d (t )

Figure 3. (a) Modèle de Smith et (b) exploitation de la courbep-sp

(6)

En supposant les efforts externes nuls le long de tiges, la connaissance de εd(t) et εr(t) permet de calculer les signaux de force FN(t) et de vitesse vN(t) pour tout point N situé en dessous du point de mesure A, notamment dans la pointe, selon la solution proposée par (Karlsson et al. 1989). FN (t ) 





1 FN 1 (t  t n ( n1) )  FN 1 (t  t n ( n1) ) 2 Z  n v N 1 (t  t n( n1) )  v N 1 (t  t n ( n1) ) 2





(7)





1 v N 1 (t  t n( n1) )  v N 1 (t  t n( n1) ) 2 (8) 1  FN 1 (t  t n( n1) )  FN 1 (t  t n( n1) ) 2 Zn avec Δtn-(n-1)=(xn-1-xn)/cn et Zn=EnAn/cn l’impédance mécanique de la section n définie par son module de Young En, sa section An et la célérité de l’onde cn. Ainsi, à partir des enregistrements εA(t) et aA(t) il est possible de calculer les signaux de force Fp(t), vitesse vp(t) et enfoncement sp(t) en pointe et donc de tracer la courbe p-sp pour chaque coup de marteau fourni lors du battage. De nombreux tests ont été réalisés pour valider la faisabilité d’un tel essai. Dans la figure 2 on présente un exemple de courbes obtenues pour deux types de sol. On peut remarquer que ces courbes sont répétitives pour un même matériau et varient selon la nature du milieu ausculté permettant d’identifier des comportements différents.

 v N (t )



2.2



Exploitation de la courbe charge-enfoncement p-sp

Pour interpréter la courbe p-sp il a été proposé une méthodologie fondée sur les travaux de (Smith 1962).

500

Les valeurs de Rd(t) et du coefficient d’amortissement de Smith Js sont déterminés dans l’intervalle d’enfoncement [se; smax], avec se et smax les enfoncements élastique et maximal, en écrivant que Rd(t) = qd(t)-Rs et Js=Rd(t)/(Rsvp(t))(figure 3.b). 2.2.2

Paramètres de déformation

Une fois l’enfoncement maximal smax atteint, on admet que sol et pénétromètre se mettent à vibrer ensemble dans un régime pseudo-élastique. Dans cette partie de la courbe p-sp, deux modules sont ainsi définis : un module de déchargement EdP3 (droite AB) et un module de rechargement ErP3 (droite BC) (figure 3.b). En assimilant la pointe à une petite plaque encastrée à l’intérieur d’un massif élastique semi-infini, on calcule la valeur de Ed,rp3 en appliquant la équation de Boussinesq (9) proposée par (Arbaoui 2006). q d  d p 1 (9) s p 4 k M avec ν supposé égal à 0,33, dp le diamètre de la pointe et kM le coefficient d’encastrement de mindlin. d 'r E (1   2 ) p3

2.2.3

Célérité des ondes cp et cs

La célérité des ondes de compression cp et de cisaillement cs dans le sol est calculée par le biais des polaires de choc préconisée par (Aussedat 1970). Pour chaque impact on mesure les pics des ondes descendantes et remontantes dans un espace de temps to+2Lt/ct nous permettant de calculer la valeur de cp (Benz 2009). La valeur de cs est calculée d’après l’expression (10) en supposant la valeur de ν égale à 0,33.


Tableau 1 – Caractéristiques des matériaux et des éprouvettes. Caractéristiques Tmax/2mm/80µm/IP OPN – WOPN éprouvettes

-

w% s kN/m3 %OPN –D.R% qd(MPa) Pda2 Etriaxial MPa Eoed MPa (*)

1

Sable d’Allier (GTR : B1 - USCS : SP) 2,73mm / 90,3% / 1,14% / NP 17,07 kN/m3 – 7%

0,8 16,1 94 - 28 2,1 2,9 28,1

2

0,8 16,83 99 - 57 4,5 11,2 97,6

3

14,6 16,26 95 - 35 2,4 4,1 37

 cs c p (1  2 ) 2(1   )

4

1

14,6 16,52 97 - 44 4,6 6,3 57,6

0,6 15,73 87 9,8 19,6

(10)

Enfin, à l’issue d’un sondage Panda 3®, on trace en fonction de la profondeur z les pénétrogrammes de : résistance de pointe qd, célérité des ondes cp et cs, module pénétromètrique Edp3 et du coefficient d’amortissement Js (i.e. figure 6) 3

3

0,6 17,43 96 43 63,5

4

15,38 16,65 92 1,9 24

5

15,61 17,43 96 2,8 36

6

19,15 16,72 93 1,2 17,6

pour chaque impact réalisé lors du battage, la courbe p-sp a été déterminée. Un exemple des courbes obtenues lors d’un essai pour quatre éprouvettes différentes est présenté dans la figure 5. Dans l’exemple, les échelles de charge p et d’enfoncement sp sont normalisées par rapport à la charge p-max et à l’enfoncement sp-max maximal mesurés pour chaque coup tracé.

(a) sable d’allier moyennement dense

(b) sable d’allier dense

Réalisation des éprouvettes et des essais

Diverses éprouvettes ont été réalisées en faisant varier la densité s et la teneur en eau w pour chaque sol étudié (tableau 1). Le compactage des éprouvettes est réalisé à l’aide d’un vérin équipé de différents capteurs suivant la procédure décrite par (Chaigneau 2001). Lors du compactage, des essais de chargement œdométriques étaient réalisés selon la procédure préconisée par (Gorena 2007). L’exploitation des courbes œdométriques a permis de déterminer les valeurs Eoed présentés dans le tableau 1. Une fois les éprouvettes réalisées, trois essais de pénétration étaient réalisés avec le Panda 2 et le Panda 3 et dont au moins 1 avec ce dernier (figure 4.c).

Figure 4. (a) essais de chargement œdométriques lors du compactage, (b) éprouvette compactée et (c) réalisation des essais Panda 3®.

Des essais triaxiaux ont été également réalisés pour le sable d’Allier. Les échantillons, conçus à même densité que les éprouvettes auscultées, ont été cisaillés dans un chemin triaxial à pression de confinement égale à 50kPa et les résultats obtenus (module tangent initial Etriaxial) sont présentés dans le tableau 1. 3.2

2

0,6 16,69 92 33,9 38,1

ESSAI EN CHAMBRE DE CALIBRATION

Une série d’essais a été réalisée dans une chambre de calibration composée d’un cylindre en acier de 400mm de diamètre et 810mm de hauteur (figure 4). L’objectif était d’une part de valider les résultats obtenus par l’exploitation de la courbe p-sp du Panda 3®, d’autre part de vérifier leur sensibilité à l’état du sol et enfin de les confronter aux résultats obtenus par le biais d’essais classiques (œdomètre, triaxial…). Deux sols ont été employés : un sable d’Allier et une argile de Laschamps. 3.1

Argile de Laschamps (GTR : A2 - USCS : ML) 0,08mm / 99,3% / 96,3% / 15,1% (WL :42,7% ) 18,08 kN/m3 – 15,8%

Résultats

Au total, une dizaine d’essais Panda 3® ont été réalisés, un pour chaque éprouvette (tableau 1). Pour chaque essai et donc

501

(c) argile de Laschamps moyennement dense

(d) argile de Laschamps dense

Figure 5. Courbes p-sp obtenues pour du Sable d’allier (a) s : 16,26kN/m3, w% :14,6, (b) s : 16,83kN/m3, w% :0,8 et pour de l’Argile de Laschamps (c) s :16,72kN/m3, w% :19,15 et (d) s :17,43kN/m3, w% :0,6

A partir des courbes p-sp exposées, on peut remarquer que celles-ci sont caractéristiques et répétitives pour chaque sol ausculté. De même, l’allure des courbes est sensible à l’état du sol. Dans le cas du sable d’Allier, on constate que la courbe psp devient presque asymptotique avec l’augmentation de la densité et que le retour élastique augmente aussi (figure 5.a,b). Quant à l’argile, on peut remarquer que les courbes p-sp sont sensibles à l’état du sol. Lorsque le sol est très humide (figure 5.c) la contrainte augmente très rapidement jusqu’au pic p-max, puis elle chute à même vitesse vers la valeur résiduelle. Cela n’est pas le cas lorsque le sol est sec et plus dense (figure 5.d) Pour chaque éprouvette, l’ensemble de courbes p-max est exploitée automatiquement selon la procédure citée dans §2.2. Les paramètres calculés sont tracés sous forme de pénétrogrammes, tel que montré dans la figure 6. Dans l’exemple présenté, on compare les pénétrogrammes obtenus pour deux éprouvettes de sable d’Allier à différents états de densité (D1 et D2). De manière générale, on peut constater que l’ensemble des paramètres est sensible à l’évolution de la densité du milieu. A partir des pénétrogrammes obtenus nous avons calculé la valeur moyenne pour chaque paramètre issu de l’exploitation des courbes p-sp (tableau 2). On peut remarquer que ceux-ci varient en fonction de la nature et de l’état du sol ausculté. De même, l’ordre de grandeur des paramètres calculés, tel que la


célérité des ondes cp et cs, a une bonne correspondance avec Résistance de pointe, (MPa)

Profondeur, z (m)

0.1

1

10

100

10

ceux proposés dans la littérature (Sharour et Gourvès, 2005).

Célérité des ondes (m/s) 100

10

1000

Module, Edp3 (MPa)

Coefficient d'amortissement Js

100

0.0

0.0

0.0

0.0

0.0

0.1

0.1

0.1

0.1

0.2

0.2

0.2

0.2

0.3

0.3

0.3

0.3

0.4

0.4

0.4

0.4

0.5

0.5

0.5

0.6

0.6

0.6

0.7

0.7

0.7

*Pda2 : valeurs obtenus avec la formule des Hollandais selon (NF 94‐105).

0.5 0.6 0.7

Pda 2 ‐ D1 Pda 2 ‐ D2

0.8

Pda 3 ‐ D1 Pda 3 ‐ D2

0.9

0.8

Cp ‐ D1

Cs ‐ D1

Cp ‐ D2

Cs ‐ D2

0.8 0.9

0.9

0.2

0.4

0.6

0.8

1.0

0.8

Ed P3 ‐ D1

Ed p3 ‐ D2

Js ‐ D1

Js ‐ D2

0.9

Figure 6. Résultats Panda 3® obtenus en chambre de calibration pour un sable d’Allier sec à 2 états de densité différents D1, D2 (éprouvettes 1 et 2). Tableau 2 – Synthèse des résultats obtenus à l’aide du Panda 3® dans chambre de calibration éprouvettes w% s kN/m3

-

qd P3 (MPa) cp (m/s) cs (m/s) EdP3 (MPa) Js (Ns/m)

Argile de Laschamps (GTR : A2 - USCS : ML)

Sable d’Allier (GTR : B1 - USCS : SP)

1 0,8 16,1

2 0,8 16,83

2,3 620 298 37 0,26

11 1504 708 120 0,09

3 4 1 2 3 14,6 14,6 0,6 0,6 0,6 16,26 16,52 15,73 16,69 17,43 Résultats Panda 3® (valeurs moyennes calculées pour chaque éprouvette)

1,6 383 184 41 0,15

2,6 504 242 65 0,15

8 2380 1145 60 0,63

Par ailleurs, nous avons confronté les valeurs moyennes des modules EdP3 obtenus pour chaque éprouvette avec ceux obtenus par le biais des essais triaxiaux (cas du sable) et œdométriques (figure 7). Dans le cas du sable, on peut constater qu’il existe une très bonne corrélation entre les modules EdP3 et les modules triaxiaux Etriaxial et œdométriques Eoed (figure 7.a). Toutefois, la corrélation EdP3-Eoed est moins bonne pour le cas de l’argile (figure 7.b). Cela peut s’expliquer par le caractère dynamique du chargement pénétromètrique et par l’augmentation des pressions interstitielles au sein du milieu pendant l’enfoncement du cône. De même, les valeurs du module Eoed obtenues lors du compactage des éprouvettes peuvent être perturbées par la création des surpressions interstitielles lors du chargement.

(a) Sable d’Allier

(b) Argile de Laschamps

Figure 7. Corrélation entre les modules EdP3 et les modules triaxiaux Etriaxial et œdométriques Eoed pour (a) Sable d’Allier et (b) Argile de Laschamps.

4

CONCLUSION

L’essai au pénétromètre dynamique bien que largement utilisé à travers le monde souffrait du peu de développements réalisés pour permettre d’améliorer la qualité des mesures effectuées et enrichir son exploitation. Cet article a présenté les développements récents réalisés sur le pénétromètre Panda 3®, qui permettent à partir de la mesure et du découplage des ondes créées par l’impact sur l’appareil, d’obtenir pour chaque coup une courbe charge-enfoncement p-sp du sol testé. L’exploitation de cette courbe permet de déterminer des paramètres de résistance (résistance de pointe), de déformation (module dynamique), des caractéristiques d’amortissement et de célérité des sols auscultés en fonction de la profondeur tout au long du sondage. Les tests réalisés en chambre de calibration ont montré la bonne répétabilité des mesures ainsi que leur sensibilité aux conditions du sol (état de serrage et état

502

32 6151 2955 130 0,65

65 5775 2773 221 0,70

4 15,38 16,65

5 15,61 17,43

6 19,15 16,72

1,5 597 286 31 0,43

3 989 475 62 0,56

4 1081 519 64 0,75

hydrique) et leur bonne adéquation avec les valeurs de la littérature. Des études comparatives entre les modules obtenus au pénétromètre et ceux obtenus à partir d’essais de chargement œdométrique ou triaxial ont montré une bonne corrélation pour le sable. Cet outil est maintenant opérationnel in situ et des travaux complémentaires sont actuellement menés en vue d’obtenir une meilleure interprétation des paramètres extraits à partir de la courbe charge/enfoncement. 5

REFERENCES

Tokimatsu, K., (1988). Penetration tests for dynamic problems, Proc., ISOPT1, 1, pp. 177-136. Gourvès R, Barjot R (1995). Le pénétromètre dynamique PANDA, Proc. of ECSMFE, Copenhagen, Denmark, 1995, p 83- 88. Benz, M.A. (2009). Mesures dynamiques lors du battage du pénétromètre Panda 2®. Thèse de l’Université Blaise Pascal, Clermont-Fd, (2009). Casem, D., Fourney, W. et Chang, P. (2003), Wave separation in viscoelastic pressure bars using single-point measurements of strain and velocity, Polymer testing 22, 2003, pp 155-164. Chaigneau, L. (2001). Caractérisation des milieux granulaires de surface à l’aide d’un pénétromètre. Thèse de l’Université Blaise Pascal, Clermont-Fd (2001). Karlsson L.G., Lundberg B, Sundin K.G. (1989), Experimental study of a percussive process for rock fragmentation, Int J Rock Mech Min Sci Geomech, 1989, pp.45-50. Smith, E.A.L. (1962), Pile-Driving Analysis by the Wave Equation, ASCE. Paper No. 3306, Volume 127, Partie I, 1962, pp 1145-1193. Arbaoui, H., Gourvès, R., Bressolette, Ph., Bodé, L. (2006), Mesure de la déformabilité des sols in situ à l’aide d’un essai de chargement statique d’une pointe pénétromètrique, Canadian geotechnical journal, vol. 43, 2006, pp. 355-369. Sharour, I et Gourvès R (2005) Réconnaissance des terrains in situ. Ed. Hermes Lavoisier.191pp. 2005 Aussedat G. (1970). Sollicitations rapides des sols, Thèse de doctorat, Faculté de sciences de l’Université de Grenoble. Gorena A. (2007). Mesure des propriétés de déformabilité de sols de référence, Mémoire d’Ingénieur CUST, Juin 2007. Benz, M.A., Gourvès, R. et Haddani, Y. (2010). Détermination de la courbe charge enfoncement dynamique en pointe pénétromètrique par découplage des ondes. JNGG 2010, Grenoble 7-9 Juillet 2010, France Tome 1, pp17-24.

th


Performance of a deep excavation in downtown Toronto Performance d'une excavation profonde au centre-ville de Toronto Cao L.F., Peaker S.M., Ahmad S.

SPL Consultants Limited, Ontario, Canada

ABSTRACT: This paper presents field measurements of soldier pile walls installed in the clayey soils and shaly rock in downtown Toronto. The method of deducing wall bending moments from the inclinometer measurements was evaluated and discussed. Backanalysis using a finite element program has been carried out to evaluate the shoring wall performance as well as the creep behaviour of the shaly rock. Recommendation for the design of soldier pile walls in the similar soils and bedrock conditions were provided. RÉSUMÉ : Cet article présente des mesures de terrain pour des murs de pieux soldats installés dans les sols argileux et le roc shaleux du centre-ville de Toronto. La méthode donnant les moments de flexion du mur à partir des mesures inclinométriques a été évaluée et discutée. Une analyse a été effectuée avec un programme d'éléments finis pour évaluer la performance du mur étayé ainsi que le comportement en fluage du shale. Des recommandations, pour la conception des murs de pieux soldats dans des conditions semblables de sols et de roc, ont été fournies. KEYWORDS: deep excavation, field measurement, inclinometer, bending moment, finite element, time-dependent deformation 1

The Georgian Bay formation is generally massive shaly rock with widely spaced jointing and sub-horizontal bedding planes. The influence of sedimentary shaly bedrock formations on the engineering performance of underground structures in Southern Ontario was summarized by Lo (1989). The shaly bedrock formations are subjected to high in-situ horizontal stresses with typical coefficient of lateral earth pressure Ko of 4 or greater. Upon relief of the high residual horizontal stresses, timedependent, creep-like deformations take place. These timedependent deformations that are highly stress dependent, persist well beyond the initial elastic deformations and generally exceed the magnitude of the elastic movements. Soldier piles of steel H-beam W610x82 at 3.05 m spacing with wood lagging were employed to support an approximately 14 m deep excavation in which 9.3 m excavation was inside overburden soils and 4.7 m excavation inside the bedrock. The soldier piles were installed typically 16 m below the existing ground surface in 910 mm diameter drilling holes. The drilling holes were backfilled by 0.4 MPa concrete with the exception at the pile toe, where 20 MPa concrete was used to support the pile toe. Two layers of tiebacks were installed at approximately 3.3 and 8.3 m below the existing ground surface, respectively to support the soldier pile walls during excavation. The tiebacks were installed within 150 mm dia. cased boreholes and bonded in bedrock. Each tieback was made up of 6 to 7 numbers of 15 mm strand tendons. The upper and lower tiebacks were installed at 45o and 25o to the horizontal direction, respectively. The bond length of the upper tiebacks was typically 5 m and the free length 9.4 m. The bond length of the lower tiebacks was typically 3 m and the free length 3.9 m. The tiebacks were generally post-grouted the day after they were installed. The typical design loads for the upper and lower tiebacks were 1000 and 800 kN respectively. Figure 1 shows outlook of soldier piles with wood lagging supported by tiebacks. Two performance tests for the tiebacks were conducted up to 138% and 200% of the design load, respectively. The test loads were maintained for 0.5 to 1 hour and the tests met the PTI criteria (PTI, 1996). Proof tests were carried out for all

INTRODUCTION

Underground structures such as basements and subway have to go deeper today than in the past due to limited space in densely populated urban environments. As deep excavations induce large stress and strain, underground structures and the adjacent structures/utilities will confront risks of being damaged. As the soil/rock stress-strain behaviour is non-linear and affected by many factors, it is difficult to predict the ground movement induced by excavation. In practices, field measurements are widely used to monitor soil/rock behaviour and to control ground movement. This paper presents a case study of a deep excavation in downtown Toronto. Soldier piles with tiebacks were used to support the excavation. Two inclinometers and one hundred and twenty seven reflective targets were installed to monitor the movements of the shoring walls during and after excavation. The inclinometer measurements have been used to deduce the wall bending moments. A finite element program has been carried out to evaluate the performance of the shoring walls. It is found that the total stress analysis leads a good prediction of wall deflections during the excavation, whereas the effective stress analysis is required to model the behaviour of shoring walls after excavation. The back-analysis also shows the evidence of the creep movement of the shaly rock. 2 GROUND CONDITION AND TEMPORARY SUPPORT SYSTEM The site is located at 352 Front Street West in Toronto, Ontario. Field investigation with drilled boreholes revealed that the site stratigraphy was made up of about 1 m thick, compact sand to gravel fill with asphalt surface overlying 3 to 4 m thick, firm to hard clayey silt fill over 2 to 5 m thick, stiff to very stiff clayey silt till. Both clayey fill and till are low plasticity soils. Georgian Bay formation of shale and limestone/siltstone was encountered at 9 to 9.5 m below existing ground surface. The groundwater table was about 5 m below grade.

5031


th


tiebacks. The test load was 133% of the design load and maintained for 10 minutes. All tiebacks except three tiebacks met the PTI criteria. The three tiebacks could not reach the test load due to the broken wires. A lower design load was used for the three tiebacks. Detailed discussions on the tiebacks are presented by Cao and Peaker (2011).

0

-15

-10

-5

Lateral Deflection (mm) 0 5 10

15

5 Bedrock Surface

Depth (m)

10 T ieback

Excavation Level

15

Inclinometer after upper tieback installed Reflective target after upper tieback installed 20

Inclinometer after lower tieback installed Reflective target after lower tieback installed

Figure 1. Outlook of soldier piles with wood lagging supported by tiebacks

Two inclinometers were installed inside the soldier pile walls during the pile installation. The inclinometers were monitored during and after the excavation. Figure 2 shows the monitoring results of one inclinometer including the reading taken after upper and lower tieback installations, 1 day after the excavation to bottom, and 11 months after the excavation. The lateral deflections measured by reflective targets installed at the top of soldier piles are also shown in Figure 2. The measurements of reflective targets are consistent with the inclinometer measurements.

M = I = KI(d2y/dx2)

Reflective target 1 d after excavatio to bottom Inclinometer 11 m after excavation to bottom

Reflective target 11 m after excavation to bottom 30 Figure 2. Lateral deflection of soldier pile wall

Bending Moment (kN/m) -600 0

3 BENDING MOMENT FROM WALL INCLINOMETER MEASUREMENTS

-400

-200

0

200

400

600

5 Bedrock Surface Depth (m)

The inclinometer measurements have been used to estimate wall bending moments by some researchers (Poh et al. 1999). The inwall inclinometers provide a direct measurement of the rotation. These measurements can be subsequently converted into wall deflections along the wall. The wall curvatures  can be derived from the wall deflection data. The second differential equations of the wall deflection will give the  along the wall. The bending moment M can be computed from  using the following equation (West, 1993)

Inclinometer 1 d after excavation to bottom

25

10 Excavation Level 15

(1)

Inclinometer after upper tieback installed

where E is the elastic modulus of the wall, I is the inertia moment of the wall, y is the lateral deflection of the wall and x is the distance along the wall. Using Microsoft Excel spreadsheet, the inclinometer measurements were fitted with a sixth- degree polynomial and double differentiation of this polynomial gave . The coefficient of determination value obtained during the curve fitting ranged from 0.98 to 0.99, indicating minimal error during the process of curve fitting. The Young’s moduli of 0.4 MPa concrete and H-beam W610x82 were taken as 2.8 GPa and 200 GPa, respectively. The sum of concrete EI and H-beam EI was used in the calculation of the bending moment. Figure 3 shows the bending moments deduced from the wall inclinometer measurements. Higher bending moments were observed at the locations of tiebacks. However, significant high values of bending moments were obtained near the ground surface, which is against the typical distribution of bending moment along a cantilever beam. This could be an error inducted in the double differentiation of the wall deflection. Further study using a

504

20

1 d after excavation to bottom 11 m after excavation to bottom

25 Figure 3. Wall bending moments deduced from inclinometer measurements

higher degree polynomial and a defined boundary condition is required. 4

FINITE ELEMENT BACK-ANALYSIS

The finite element program Phase 2 (version 8.0) was used in the back-analysis. The program can be used to simulate excavation in soil and rock under plane strain condition. Sixnode triangle elements were used to model the soil and bedrock media. The soldier pile wall and tiebacks were modelled by



structural beam elements. The analysis modelled a half width of the excavation where the right-hand boundary of the mesh represented the line of symmetry at the centre line of excavation. The finite mesh was 140 m long and 84 m deep. The half width of the excavation was 20m. The bottom boundary was strained from both vertical and horizontal movements. The left-hand and right-hand boundaries were free to move in the vertical direction. The soil and bedrock profiles used in the analysis were based on borehole logs. The groundwater level was taken at 5 m below existing ground surface at the initial stage. During the excavation, the groundwater level was assumed to be drawn down to the excavation level at the excavated side. For the surface sandy fill, the Young’s modulus E of 25 MPa and the friction angle  of 30o were assumed. For the clayey soils, the undrained shear strength su was estimated from 6N, where N is the blow counts of the standard penetration testing. The residual su was taken as 50% of the initial su. The undrained E of clayey soil was estimated from 1500su for the native low plasticity clayey silt till and 500su for the clayey silt fill, respectively. The soil Poisson’s ratio was taken as 0.3. The unit weight  was obtained from available laboratory testing data. Mohr Coulomb failure criterion was used for soils. The soil properties used in the analysis are shown in Table 1.

The soldier pile wall was modelled as reinforced concrete with W610x82 at spacing of 3.05 m. The equivalent thickness of 0.4 MPa concrete was taken as 0.2 m and the Young’s modulus was 2.8 GPa. The concrete compressive and tensile strengths were taken as 400 kPa and 40 kPa, respectively. The compressive and tensile strengths of W610x82 were taken as 345 MPa. The Poisson’s ratio for steel and concrete was taken as 0.2. The equivalent bolt diameters for the upper and lower tiebacks were taken as 32 mm and 24.5 mm, respectively. The Young’s modulus of tiebacks was taken as 200 GPa. The bond shear stiffness was taken as 6000 kN/m/m based on the tieback proof test results. The bond lengths of the upper and lower tiebacks were taken as 5 m and 3 m, respectively. The spacing of tiebacks was taken as 3.05m. The measured and computed wall deflections after the installation of upper tiebacks and the excavation just to the bottom are shown in Figure 4. The computed wall deflections are in a good agreement with the inclinometer measurements, indicating that the in-put parameters used in the analysis are reasonable. The computed bending moments for the excavation just to bottom are compared with those deduced from the inclinometer measurements as shown in Figure 5. The bending moments deduced from the inclinometer measurements are comparable with the computed except near the ground surface where significant high values deduced from the inclinometer measurements. Ignoring the high bending moments near the ground surface, the bending moments deduced from the inclinometer measurements can be used for the checking of the capacity of the soldier piles. The inclinometer measurements show that up to 7 mm lateral movement was developed after the excavation to bottom as shown in Figure 2. This could be due to three possible reasons: (1) the consolidation of clayey soil; (2) de-stressing of tiebacks; and (3) time-dependent deformation of the shaly rock upon relief of the initial high horizontal stresses. The first two possible reasons have been studied in the finite element analysis using the effective parameters and reduced modulus for the

Table 1. Soil parameters used in the finite element analysis (kN/m3)



E (MPa)

su (kPa)



Ko

Sandy fill

20

25

0

30

0.5

1– 4.5

Clayey fill

20

30

60

-

0.75

4.5 – 9.3

Clayey till

21

225

150

-

0.75

Depth (m)

Type

0–1

Note: Ko is the coefficient of lateral earth pressure (total stress)

0



E (MPa)

mb

s

a

Ko

Weathered bedrock

25

244

0.3

0.004

0.52

2

Sound bedrock

26

3072

1.3

0.004

0.51

4

9.3 – 11.3 >11.3

-5

Lateral Deflection (mm) 0 5 10

15

Bedrock Surface

Excavation Level 15

20

(kN/m3)

Type

-10

10

Table 2. Rock parameters used in the finite element analysis Depth (m)

-15

5

Depth (m)

For the jointed shaly bedrock, the generalized Hoek-Brown constitutive model was used. The following parameters were used to generate the generalized Hoek-Brown rock-mass strength criterion: (1) The geological strength index was taken as 60 for sound bedrock, respecting blocky to very blocky, good to fair joint surface, and 30 for weathered bedrock, respecting blocky/disturbed/seamy joint surface; (2) The intact rock constant was taken as 8 for sound bedrock (highest value for shale) and 4 for weather bedrock (lowest value for shale); (3) The disturbance factor was taken as 0 for excellent quality controlled excavation; and (4) The modulus ratio was taken as 250 for bedrock (highest value for shale) and 150 for weather bedrock (lowest value for shale). The intact compressive strength was obtained from available results of rock point load testing and unconfined compressive testing. The rock Poisson’s ratio was taken as 0.15. Based on the above assumptions, the obtained strength parameters for the generalized Hoek-Brown’s model are summarized in Table 2.

Inclinometer readings after upper tieback installed Computered deflections after upper tieback installed

25

Inclinometer readings after excavation to bottom Computered deflections after excavation to bottom

30 Figure 4. Measured and computed wall deflections

Note: mb, s and a are parameters used in generalized Hoek-Brown’s model

505



5

Bending Moment (kN/m) -600 0

-400

-200

0

200

400

In-wall inclinometer has an importation role in the monitoring of shoring walls during and after excavation. Wall bending moments can be estimated from the inclinometer measurements except near the ground surface where the bending moments are overestimated probably due to the error in double differentiation of the wall deflection. Both total and effective stress analyses should be used for the design of shoring walls installed in the clayey soils. As supported by the finite element back-analysis, the clayey soils behave as undrained during excavation and as drained after excavation. The finite element analysis confirmed that the timedependent deformations of shale rock during 11 months after excavation could be up to 5 mm for 4.5 m excavation in the shaly rock. The time-dependent deformation should be considered in the shoring wall design.

600

5

Depth (m)

Bedrock surface 10 Excavation Level 15

6 Deducted from inclinometer readings (1 d after excavation to bottom)

20

Computed (1 d after excavation to bottom) 25 Figure 5. Deduced and computed wall bending moments

clayey soils, and reduced prestressing for tiebacks. Figure 6 show the comparison of the computed and measured lateral deflections. In this analysis, the E was taken as 80% of the initial E; the effective  and cohesion for clayey fill were taken as 30o and 3 kPa, respectively; the effective  and cohesion for clayey till were taken as 32o and 7.5 kPa, respectively; and the lower tiebacks were assumed to be de-stressed to 50% of the initial prestressing. The computed lateral deflections within the upper portion of the overburden are in a good agreement with the inclinometer reading. However, there is a difference of up to 5 mm between the computed and measured deflections in the lower portion of the overburden and the bedrock, which should belong to the time-dependent deformation of the shaly rock as the de-stressing of tiebacks or the consolidation of clayey soils could not lead such deformation.

0

-15

-10

Lateral Deflection (mm) -5 0 5

10

15

5 Bedrock Surface Depth (m)

10

15

20

CONCLUSIONS

T ieback Excavation Level

Inclinometer reading (11 m after excavation to bottom) Computed deflections (11 m after excavation to bottom)

25 Figure 6. Measured and computed wall deflections (11 months after excavation)

506

REFERENCES

Cao L.F. and Peaker S.M. 2011. Use of tieback in Southern Ontario. 64th Canadian Geotechnical Conference, Toronto, Paper 175 Lo K.Y. 1989. Recent advances in design and evaluation of performance of underground structures in rocks. Tunnelling and Underground Technology 27, 461-477. Poh T.Y., Goh A.T., Wong K.S., Wong I.H. and Poh K.B. 1999. Determination of bending moments in diaphragm wall. 5th International Symposium on Field Measurements in Geomechanics, Singapore, 229-234. PTI 1996. Recommendations for prestressed rock and soil anchors. Post-Tension Institute, Phoenix. West H.H. 1993. Fundamentals of structural analysis. John Wiley & Sons, Inc.

Permeability scale effect in sandy aquifers: a few case studies Effet d’échelle et perméabilité des aquifères sableux : quelques études de cas Chapuis R.P.

École Polytechnique, Montréal, QC, Canada

ABSTRACT: In sandy aquifers, stratification results in a range of values for the hydraulic conductivity K, which can be evaluated at three scales. Since large-scale tests are more likely to meet preferential flow paths, they are also likely to yield larger K values than small-scale tests, which may be viewed as a scale effect. The small scale is that of soil samples: their quality must be assessed and their grain size distribution analyzed to check for mixes of sub-layers, before using reliable methods to predict the K values. The middle scale is that of field permeability tests for which it is important to respect the standards and perform verifications. The large scale is that of pumping tests. The paper presents a few case studies of sandy aquifers. Their stratification led to unimodal or multimodal grain size distributions. For all cases, the K distributions provided consistent images of the sandy aquifers. It was then concluded that, after a quality control of data and interpretations, there was no scale effect in the aquifers. RÉSUMÉ : Dans les aquifères sableux, la stratification donne une gamme de valeurs pour la conductivité hydraulique K qui peut être évaluée à trois échelles. Les essais à grande échelle ayant plus de chances de tester des zones d’écoulement préférentiel, ils ont aussi plus de chances de donner des valeurs élevées de K que les essais à petite échelle, ce qui peut être vu comme un effet d’échelle. La petite échelle est celle des échantillons : leur qualité doit être évaluée et leur granulométrie analysée pour détecter les mélanges de strates, avant d’utiliser des méthodes fiables de prédiction de K. L’échelle moyenne est celle des essais de perméabilité in situ pour lesquels on doit respecter les normes et faire des vérifications. La grande échelle est celle des essais de pompage. L’article présente des études de cas d’aquifères sableux. Leur stratification a donné des granulométries unimodales ou multimodales. Pour tous les cas, les distributions de K ont fourni des images cohérentes des aquifères sableux. On a conclu, après un contrôle de qualité des données et des interprétations, qu’il n’y avait pas d’effet d’échelle dans ces aquifères. KEYWORDS: aquifer, grain size distribution, monitoring well, permeability test, pumping test, scale effect 1

INTRODUCTION

In sandy aquifers, groundwater seepage is controlled by stratification, with coarse size sediments deposited at high water velocities and small size sediments settling at low water velocities, or in temporary ponds. Many methods can be used to assess the hydraulic conductivity, K, which can vary over orders of magnitude. It is often believed that since large-scale tests involve large volumes, which are more likely to meet preferential flow paths, they are likely to yield larger K values than small-scale tests (Bradbury and Muldoon 1990; Rovey 1998; Rovey and Niemman 1998). Thus, there should be a scale effect for the K value, some increase with the tested volume. There is no consensus about this scale effect. Many studies tried to check or challenge theoretical opinions. They differed about testing techniques, investigated scales, and geologic media. Alas, the quality of each K value usually was not questioned even if poor quality data and interpretation are known to yield an artificial scale effect. Regrettably, the quality control of groundwater parameters, which must be methodically completed for engineered facilities, is not always done (Chapuis 1995). This paper examines quality control issues with data and interpretation, in order to exclude artificial scale effects. The idea of scale effect was rejected by Butler and Healey (1998). They argued that scale effect results from artifacts linked to incomplete well development and low-K skins around well screens, but they did not study what produce a positive or negative skin. These skin phenomena and their effects on the apparent K value being related to safety issues, they are more studied in geotechnique (Chapuis and Chenaf 2010) than in geosciences. Moreover, many studies have not examined how incorrect interpretation methods for slug tests and pumping tests can yield artificial scale effects. However, the quality control of slug test

507

methods has been largely investigated in geotechnique (Chapuis et al. 1981; Chapuis 1988, 1998, 1999, 2001; Chapuis and Chenaf 2002, 2003). For pumping tests in unconfined aquifers, the large-scale K values obtained were shown to be incorrect if the interpretation was performed using current methods for unsteady-state (Akindunni and Gilham 1992). Therefore, when studying scale effect, some caution must be observed to avoid using scale effect as a final excuse, or as a fudge factor, when the heterogeneity of the tested material could have been more thoroughly investigated and when errors involved in sampling, testing and interpretation methods could have been taken into account. Note that properly taking into account scale effect is important for numerical analyses, since an aquifer numerical model cannot be as detailed as the physical reality. Most often, the grids of numerical models cannot contain enough elements to model the detail of real features. This is why up-scaling techniques are needed to define some equivalent K value for grid elements (Renard and de Marsily 1997; Zhang et al. 2011). In this paper, the results of three sites are briefly examined. The small scale, about 10-3 m3, is that of samples recovered in boreholes for which the K value was evaluated using predictive methods. The middle scale, about 1 m3, is that of field permeability tests in monitoring wells. The large scale, about 103 m3, is that of pumping tests. Now, the problems linked to the collected data at three scales in sandy aquifers are examined in detail, starting with the soil samples taken in boreholes. 2

SMALL-SCALE K VALUES (SAMPLES)

Many soil samples can be taken in boreholes, usually with a split spoon. Quality issues relative to soil sampling have been the topic of many geotechnical researches. Five sample classes are defined by considering the relationships between sampling


THE SITES

5.1

The Lachenaie site

The site is located 50 km north-east of Montreal. The sand unconfined aquifer has been used for field training and research. The GSDCs could be correctly fitted using a unimodal lognormal distribution. The little variability for the mean and the standard deviation indicate homogeneity (Fig. 1). For the pumping test, the steady-state drawdown data were used, the interpretation methods being proven to be reliable (Chapuis et al. 2005a, b). In this aquifer, the average K values at the three scales are very close, and thus there is no scale effect (Fig. 2). 0 mean 50

sdt dev.

100 150 200 250 300 350 -2.0

-1.0

0.0

1.0

2.0

log (mean size) and log (std dev.)

Figure 1. Lachenaie: modal decomposition of the sand GSDCs.

MEDIUM-SCALE K VALUES (SLUG TESTS IN MWS)

The middle scale, about 1 m3, is that of permeability tests (slug tests) performed in monitoring wells. It is important to use the standard methods to interpret the slug test data. In Canada, CAN/BNQ 2501-135 is the standard for an overdamped response (CAN/BNQ 1988, 2008), but there is no standard for an underdamped response. ASTM, however, has standards for the underdamped response (ASTM 2012a) and for the critically damped response (ASTM 2012b). For overdamped slug tests, the velocity graph method helps to establish the correct piezometric level (PL) and K value for the test. It also helps to detect several phenomena during the test. Even if the aquifer is unconfined, and even if the MW is correctly installed, there are several reasons why the test data must be corrected by a systematic error on the assumed PL, of a few centimetres (Chapuis 2009a, b). The velocity graph gets rid of any systematic error, which may be due to incorrect calibration of a pressure transducer (PT), waiting time, PT line slippage, piezometric modification, faulty MW installation, and unknown PL. However, it cannot make a distinction between these six errors. For underdamped slug tests, it is preferable to fit the test data using a least squares method, instead of a visual fit, and the verification of three physical conditions must be done for each tests, otherwise large errors can be made (Chapuis 2012c). 4

5

LARGE-SCALE K VALUES (PUMPING TESTS)

For the large scale of pumping tests, about 103 m3, precautions must be taken when installing the pumping well and MWs, and also when interpreting the pumping test data. The common theories for unsteady-state are based on some wishful thinking

508

100% 90%

% of cases lower than x

3

about drainage, unsaturated seepage and a misleading concept of specific yield (e.g., Akindunni and Gilham 1992; Chapuis et al. 2005a). For MWs, it is commonly admitted that two thirds of them are improperly installed (Nielsen and Schalla 2005).

mean sampling depth (cm)

methods, quality of sample and quality of laboratory tests. All borehole samples in sandy aquifers are of class-3 or class-4 quality. For information, the class-4 quality is obtained with the hollow stem auger, rotary, percussion, cable tool and sonic drilling methods (Baldwin and Gosling 2009). These methods strongly influence not only the quality of samples, but also the quality of permeability tests, and the quality of the MW installation (Chapuis and Sabourin 1989; Chesnaux et al. 2006; Chesnaux and Chapuis 2007). In sandy aquifers, a tube sampler with a clear plastic liner can be used. This tool does not provide class-1 or -2 samples. It roughly preserves the grain size distribution curve (GSDC), with major margin disturbance (thick-walled sampling) plus some mixing between adjacent sub-layers. It does not preserve the water content w, void ratio e, and K in situ values. For that reason, this sampler provides class-4 samples, and not intact ones as claimed in a few papers. Several methods can be used to predict the K value of a soil sample. Chapuis (2012a) listed 45 methods and assessed their capacity against large data sets for laboratory permeability tests performed on homogenized fully saturated specimens. All tests were not plagued by one of the 14 most frequent mistakes when performing such tests. For sandy aquifers, the in situ porosity n can be assessed using the method of Chapuis (2012b) and the K values can be predicted with the method of Chapuis (2004), which yields good predictions for natural soils in the ranges 0.003 ≤ d10 ≤ 3 mm and 0.3 ≤ e ≤ 1. The range for the effective diameter d10 was recently extended up to 150 mm (Côté et al. 2011; Chapuis et al. 2012). If the soil sample is homogenous, its GSDC is smooth. This is not the case for most borehole samples in sandy aquifers. Therefore, when studying the GSDCs, caution must be taken to avoid confusing homogenous samples (single layer) with those made by mixing 2 or 3 small layers. The analysis proceeds with a modal decomposition (Chapuis 2010; Chapuis et al. 2013), which provides the GSDC and percentage of each layer in the composite sample. The equivalent horizontal K value (stratified sample) is then obtained using the composition rule.

80%

pred. K (small scale) K tests in MWs lognormal best fit MWs data pumping

70% 60% 50% 40% 30% 20% 10% 0% 1.E-03

1.E-02

1.E-01

1.E+00

x = K (cm/s)

Figure 2. Lachenaie: comparison of the K values obtained at the small, medium, and large scales.

5.2

The Shannon site

Shannon is a small town about 30 km north-west of Quebec City. For the TCE-contamination case, a lot of information was given in the defendants' expert reports, but without a quality control, which led to contradictions. The quality control and a synthesis were done in Chapuis (2009c, 2010, 2013a, b). There were about 1000 MWs for this huge contamination case. The aquifer stratification could be considered or not when analyzing the GSDCs. When it was not, the distribution of


predicted K values could not explain the high large-scale K values of pumping tests (Fig. 3). When it was, after using a modal decomposition of each GSDC, the distribution of predicted K values yielded a large-scale K value very close to that of pumping tests (Chapuis 2010, 2013b). For the slug tests in MWs, Chapuis (2010) showed that the defendants’ expert reports gave K values that were obtained without following the standards and without making the required verifications. They were about three times smaller than the K values obtained when following the standards and making the verifications. When the standards were not respected, the distribution of the slug test K values could not explain the largescale K values of pumping tests (Fig. 4). When the standards were followed, the slug test K distribution yielded a large-scale K value very close to that of pumping tests (Fig. 4). 100%


90% 80%

usual range for pumping

70% 60% 50% 40%

homogeneous

#

stratified

30% homogeneous stratified pumping wells

20% 10% 0% 1.E-05

1.E-04

1.E-03

1.E-02

x = estimated K (m/s)

K distributions provided a coherent image of the hydraulic properties in the aquifer. Therefore, there was no scale effect. 5.3

The Sorel site

The Sorel site, 100 km north-east of Montreal, has been used for many years for field training of students in groundwater engineering and geophysics. The site is part of the floodplain at the confluence of the Richelieu River and the St-Lawrence River. Down to about 5 m deep, the stratigraphy includes many layers of fine sand (deposited in low velocity water) and silty clay (deposited in ponds). Over 300 soil samples were recovered in over 40 boreholes. The soil samples provided clearly bimodal GSDCs and K values (Chapuis et al. 2013). The split-spoon sampler could recover 30 or more individual layers of silty clay and fine sand, which were uniform in color. The GSDC modal decomposition provided results such as those of Fig. 5 for a few boreholes in the vicinity of the pumping well. The fine sand and silty clay were fairly homogeneous (Fig. 5). According to the modal decompositions, the portion between 1.9 and 3.1 m deep had more clayey silt than the upper and lower portions. The screens of the pumping well and nearby MWs were installed in the portion between 3.1 and 4.4 m this confined aquifer. The horizontal K distribution curve was obtained from the modal decomposition of GSDCs and the K composition rule. The predicted K distribution was in good agreement with the pumping test K values, whereas the slug test K values were somewhat below the pumping test K values (Chapuis et al. 2013). Due to the fine stratification of fine sand and silty clay sub-layers, the development of monitoring wells was not effective. Therefore, the slug tests have slightly underestimated the horizontal medium-scale K value due to smearing between layers during drilling and MW installation. Therefore, there was no scale effect for the Sorel highly stratified aquifer.

Figure 3. Shannon: K values predicted using the GSDCs, assuming either homogeneous or stratified samples (modal decomposition) and large scale pumping tests.

fine sand, mean clayey silt, mean 0

100%

1

80%

2

usual range for pumping wells

depth (m)


90%

70% 60% 50% 40%

NFS

20%

not following standards following standards pumping wells

10% 1.E-05

1.E-04

1.E-03

3

4

FS

30%

0% 1.E-06

fine sand, std dev clayey silt, std dev

5

6

1.E-02

-9

x = slug test K (m/s)

-8

-7

-6

-5

-4

-3

-2

-1

0

1

2

3

4

log of the means and standard deviations

Figure 4. Shannon: K values obtained with slug tests in monitoring wells (following or not the standards), and large scale pumping tests.

How to perform the modal decomposition of a GSDC, and that of predicted or measured K values, is explained elsewhere (Chapuis 2013b; Chapuis et al. 2013). These papers also explain how to predict, for a K distribution, the large-scale K value which would be given by a pumping test, in order to logically compare the data at the three scales. A closed-form equation is also provided for the soil specific surface, more general than that of Chapuis and Légaré (1992). According to the detailed study following the quality control for the Shannon aquifer, all

509

Figure 5. GSDCs modal decomposition for stratified samples of Sorel, showing fairly homogeneous layers of fine sand and clayey silt.

6

CONCLUSIONS

This paper studies the permeability of sandy aquifers at three scales. The aquifers are stratified or not, which leads to multimodal or unimodal distributions for grain size distribution curves. The small scale is that of soil samples: their quality must be assessed and their GSDC analyzed to check for mixes of sub-layers before using reliable methods to predict the K values. The middle scale is that of field permeability tests for


which it is important to respect the standards and perform verifications. The large scale is that of pumping tests, which must be interpreted for steady-state. The results presented here have shown that, when stratification is adequately considered, slug tests are interpreted according to standards, and the resulting K distributions are taken into account, the conclusion is that there is no scale effect. Therefore, using a quality control approach for analyzing the GSDCs and interpreting field test data is essential for crosschecks, and for avoiding the creation of artificial scale effects. 7

ACKNOWLEDGEMENTS

The author thanks the National Research Council of Canada, BFI Ltd (Lachenaie), the Régie de l’eau de Sorel-Tracy and Aquatech (Sorel), the Shannon Citizen’s Committee, Charles Veilleux and Associates, and the FARC (Shannon). 8

REFERENCES

Akindunni F.F. and Gillham R.W. 1992. Unsaturated and saturated flow in response to pumping of an unconfined aquifer: Numerical investigation of delayed drainage. Ground Water 30, 873–884. ASTM 2012a. D5785: Standard test method (analytical procedure) for determining transmissivity of confined nonleaky aquifers by underdamped well response to instantaneous change in head (slug test). Annual CDs of standards, vol. 04.09, ASTM, West Conshohocken, Penn. ASTM 2012b. D5881: Standard test method (analytical procedure) for determining transmissivity of confined nonleaky aquifers by critically damped well response to instantaneous change in head (slug test). Annual CDs of standards, vol. 04.09, ASTM, West Conshohocken, Penn. Baldwin M. and Gosling D. 2009. BS EN ISO 22475-1: Implications for geotechnical sampling in the UK. Ground Engineering, August 2009, 28–31. Bradbury K.R. and Muldoon M.A. 1990. Hydraulic conductivity determinations in unlithified glacial and fluvial materials. ASTM STP 1053, 138–151. Butler J.J. and Healey J.M. 1998. Relationship between pumping test and slug-test parameters: scale effect or artefacts. Ground Water 36(2), 305–313. CAN/BNQ 1988. Canadian Standard CAN/BNQ 2501-135/1988: Soils – determination of permeability by the Lefranc method. CAN/BNQ 2008. Canadian Standard CAN/BNQ 2501-135/2008: Soils – determination of permeability by the Lefranc method. Chapuis R.P. 1995. Controlling the quality of ground water parameters: some examples. Can Geotech J 32(1), 172–177. Chapuis R.P. 1988. Determining whether wells and piezometers give water levels or piezometric levels. In Ground Water Contamination: Field Methods, ASTM STP 963, 162–171 Chapuis R.P. 1998. Overdamped slug test in monitoring wells: Review of interpretation methods with mathematical, physical, and numerical analysis of storativity influence. Can Geotech J 35(5), 697–719. Chapuis R.P. 1999. Borehole variable-head permeability tests in compacted clay liners and covers. Can Geotech J 36(1), 39–51. Chapuis R.P. 2001. Extracting the local piezometric level and hydraulic conductivity from tests in driven flush-joint casings. Geotech Testing J 24(2), 209–219. Chapuis R.P. 2004. Predicting the saturated hydraulic conductivity of sand and gravel using effective diameter and void ratio. Can Geotech J 41(5), 787–795. Chapuis R.P. 2009a. Permeability or hydraulic conductivity tests in a monitoring well: Why are piezometric level corrections required? Geotech News 27(2), 46–49. Chapuis R.P. 2009b. Interpreting slug tests with large data sets. Geotech Testing J 32(2), 139–146. Chapuis R.P. 2009c. Recours collectif - Résidents de Shannon Expertise sur les conditions hydrogéologiques - Rapport préliminaire en 3 volumes, 970 p. Chapuis R.P. 2010. Recours collectif - Résidents de Shannon – Expertise sur les conditions hydrogéologiques - Rapport d’expertise, 156 p. Chapuis R.P. 2012a. Predicting the saturated hydraulic conductivity of soils: A review. Bull Eng Geology Envir 71(3), 401–434.

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Chapuis R.P. 2012b. Estimating the in situ porosity of sandy soils sampled in boreholes. Engng Geology 141–142, 57–64. Chapuis R.P. 2012c. Improved curve fitting methods for underdamped slug tests. Geotech Testing J 35(5), 752–761. Chapuis R.P. 2013a. TCE-contaminated groundwater in Shannon, Quebec: 2. Evaluating the hydraulic conductivity with permeability tests in observation wells. Bull Eng Geol Env, submitted Chapuis R.P. 2013b. TCE-contaminated groundwater in Shannon, Quebec: 3. Evaluating the hydraulic conductivity at three scales. Bull Eng Geol Env, submitted Chapuis R.P. and Chenaf D. 2002. Slug tests in a confined aquifer: Experimental results in a large soil tank and numerical modeling. Can Geotech J 39(1), 14–21. Chapuis R.P. and Chenaf D. 2003. Variable-head field permeability tests in driven casings: Physical and numerical modeling. Geotech Testing J 26(3), 245–256. Chapuis R.P. and Chenaf D. 2010. Driven field permeameters: Reinventing the wheel? Geotechnical News 28(1), 37–42. Chapuis R.P. and Légaré P.P. 1992. A simple method for determining the surface area of fine aggregates and fillers in bituminous mixtures. In Effects of Aggregates and Mineral Fillers on Asphalt Mixture Performance, ASTM STP 1147, 177–186. Chapuis R.P. and Sabourin L. 1989. Effects of installation of piezometers and wells on groundwater characteristics and measurements. Can Geotech J 26(4), 604–613. Chapuis R.P., Paré J.J., and Lavallée J.G. 1981. Essais de perméabilité à niveau variable. Proc. 10th ICSMFE, Stockholm, Balkema, Vol. 1, 401–406. Chapuis R.P., Chenaf D., Acevedo N., Marcotte D. and Chouteau M. 2005a. Unusual drawdown curves for a pumping test in an unconfined aquifer at Lachenaie, Quebec: Field data and numerical modeling. Can Geotech J 42, 1133–1144. Chapuis R.P., Dallaire V., Marcotte D., Chouteau M., Acevedo N. and Gagnon F. 2005b. Evaluating the hydraulic conductivity at three different scales within an unconfined aquifer at Lachenaie, Quebec. Can Geotech J 42, 1212–1220. Chapuis R.P., Weber S. and Duhaime F. 2012. Intrinsic permeability of materials ranging from sand to rock-fill using natural air convection tests: Discussion. Can Geotech J 49(11), 1319–1322. Chapuis R.P., Dallaire V. and Saucier A. 2013. Getting information from modal decomposition of grain size distribution curves. Geotech Testing J, submitted. Chesnaux R. and Chapuis R.P. 2007. Detecting and quantifying leakage through defective borehole seals: A new methodology and laboratory verification. Geotech Test J 30(1), 17–24. Chesnaux R., Chapuis R.P. and Molson J.W. 2006. A new method to characterize hydraulic short-circuits in defective borehole seals. Ground Water 44(5), 676–681. Côté J., Fillion M.H. and Konrad J.M. 2011. Intrinsic permeability of materials ranging from sand to rock-fill using natural air convection tests. Can Geotech J 48, 679–690. Nielsen D.M. and Schalla R. 2005. Design and installation of groundwater monitoring wells. Chapter 10, Practical Handbook of Environmental Site Characterization and Ground–Water Monitoring, 2nd edition, CRC Taylor & Francis. Renard P. and de Marsily G. 1997. Calculating equivalent permeability: a review. Adv Water Resources 20(5-6), 253–278. Rovey C.W. II 1998. Digital simulation of the scale effect in hydraulic conductivity. Hydrogeology, 6(2), 216–225. Rovey C.W. II and Niemann W.L. 1998. Wellskins and slug tests: where’s the bias? J Hydrology 243(1-2), 120–132. Zhang Y., Liu B.Z. and Gable C.W. 2011. Homogenization of hydraulic conductivity for hierarchical sedimentary deposits at multiple scales. Transp Porous Med 87, 717–737.

A Study of Cuttability Indices for Tunnel Penetration Étude sur les indices d’aptitude à la coupe pour la pénétration de tunnels Chen L.-H.

Dept. of Civil Engineering, National Taipei University of Technology, Taipei, Taiwan

Chen Y.-C., Chen W.-C., Liu H.-W.

Dept. of Construction Engineering, National Taiwan University of Science & Technology, Taipei, Taiwan

ABSTRACT: To speed up construction of mass transit subway and the popularity rate of sewage, Taiwan's underground excavation works, especially for mechanical cutting cases, show an ascendant tendency. This study presents a generalized solution for underground geological-mechanical interaction. By using dimensional analysis, this model generalizes geological characteristics grouped into three categories: (1) brittle (rock-like), (2) the ductile (soil-like), and (3) brittle-ductile (gravel-like) type with respect to two cutting forces: (1) thrust and (2) torque to evaluate their excavation/penetration rate. Furthermore, the leading cuttability indices can be obtained to enable to assess the underground excavation. Meanwhile, in-situ experimental results from shield tunneling and pipe jacking construction were used to examine this model and it showed a nice agreement between both. From this analytical approach, a proposed “oval-shaped cutting ellipsoid”, including its center (O), area (A), and long/short axis (ax/by, or ay/bx), can be used not only to estimate the functionality and efficiency of cutting machine adopted for tunnel project, but also to offer a warning information for inadequate cutting strategy. RÉSUMÉ: Pour accélérer la construction du système de transport en commun souterrain et des eaux usées, les travaux d'excavation souterrains de Taïwan, en particulier les coupes mécaniques, montrent une tendance ascendante. Cette étude présente une solution généralisée pour l’interaction géologique-mécanique souterraine. En utilisant l'analyse dimensionnelle, ce modèle généralise les caractéristiques géologiques regroupées en trois catégories: (1) fragile (comme la roche), (2) l'ductile (comme le sol), et (3) fragileductile (comme le gravier) en respectant deux types de coupe: (1) la poussée et (2) torsion pour évaluer leur taux d’excavation / pénétration. Par ailleurs, les indexes d’aptitude à la coupe peuvent être obtenue pour permettre d'évaluer l'excavation souterraine. Pendant ce temps, les résultats expérimentaux in-situ de bouclier tunnel et de la construction de tuyau de fonçage ont été utilisés pour examiner ce modèle et ceux-ci concordent. A partir de cette approche analytique, une proposition de «ellipsoïde de coupe de forme ovale " (comprenant son centre (O), sa surface (A) et ses axes longs et courts (ax/by, ou ay/ bx) ) peut être utilisé non seulement pour estimer le bon fonctionnement et l'efficacité de la machine de découpage adopté pour le projet de tunnel, mais aussi pour fournir une alerte à propos d’une stratégie de coupe inadaptée. KEYWORDS: Generalized cutting mechanism , Thrust, Pipe jacking, Cuttability indices 1 1.1

INTRODUCTION

開挖機具

DT = 11740 mm

Multi-scale underground cutting project

Recently, the construction projects increase the cases of underground tunneling by mechanical cutting such as tunnel in the mountain, mass rapid transportation system in the city and sewer system, etc. There are different types of cutting methods including TBM, shield tunnel (ST), as well as pipe jacking (PJ) with various sizes corresponding to different geological conditions (see Figure 1). This study presents a normalized evaluation to meet the multi-scale underground cutting projects so that all of the in-situ data can be collected and compared with each other.

DS = 2920 mm

DP = 1500 mm

Figure 1. Different size of cutting machines ranged from 11740 to 1500 mm in diameter.

2 2.1

CONCEPTUAL MODEL Indentation-typed fracture mechanism

Based upon normal indentation fracture in a MohrCoulomb material, Huang （ 2000 ） proposed a conceptual model as follows:

511


(1   ) *( kd 1)/ kd  

*( k p 1)/ k p

(1)  where γ is a key dimensionless factor, which is a function of wedge angle of cutter , elastic constants and plastic strength parameters. (2)  * = * E,   or G  , qu ,  ,  * ,  i ,  c





and ξ is defined as dimensionless elasto-plastic (E-P) radius while ξ* reach critical value where brittle fracture occurs on this E-P interface. Thrust force, therefore, can be estimated using the indentation pressure P and indentation force F as seen in Figure 2 schematically.  P 1  (n  1)  K p n (K p 1) / K p (3)   *  1  q K p  1  K p  n 

Fi (3  n)  

n 1

d )  P( tanβ

(4)

Figure 3b. Trust force system with inclined angle

Therefore, Ff (kN) is estimated from different types (nj) of cutters fj and water/earth pressure Ps (kN/m2): m

Ff   n j  f j  Ps  A

(6)

j1

F

where A is cross area of cutter head (m2). This paper presents an analytical estimation to deal with different mechanical cutting methods (tunnel boring machine, shield tunnel and pipe jacking), construction types (earth pressure balance, slurry pressure balance, thick-mud), and geological conditions (soil, gravel and rock) by generalizing their total thrust system. The straight-line thrust is calculated for either wedge- or conical-typed cutters of tunneling machine. In this generalized work, the upper bound and lower bound of trust are highlighted for the warning situations for risk assessment.

2a In denter





r*

X

d

Core

E-P interface

Plastic zone

E lastic zone Intrinsic flaw

 x , max

3

y

3.1

Figure 2. Schematic normal indentation fracture

2.2

Generalized trust system

This study presents a generalized trust system of cutter head globally by taking each different types of individual cutters into account locally with respect to different methods (TBM, ST, and PJ) and geological conditions (rock, soil, & gravel). Figure 3a and 3b show the total trust force F, which is consists of front resistance Ff and lateral resistances Fp including both Fp,m for machine itself and Fp,p for pipes. (5) F Ff  Fp,m  Fp,p

CASE STUDY Case I: Taoyuan tunneling project in Taiwan

In addition, the in-situ data of trust in shield machine (Taoyuan tunneling project) is presented to confirm with. It depicts a favorable agreement for the estimation of thrust in this study as shown in Figure 4 (cutter head), and Figure 5 (results) with respect to normal cutting as well as abnormal conditions (point a and b shown in Fig.5) once the in-situ data out of the theoretical boundaries.

Fp

Ff

F Fp,m

Fp ,p

Figure 4. Cutting head in field for shield tunnel project

Figure 3a. Trust force system

512


4

CONCLUSIONS

The results shows that the total thrust for upper bounds and lower bounds are: (401%, 37.8%) and (258.2%, 31.7%) compared with normal condition in gravel and weathered sandstone cases respectively, which normalized boundary values are (13×10-4 ~ 82×10-4) and (0.97 ~ 4.98) for cuttinghead resistance respectively. It also found that the cutting-head resistance take about 28% of the total trust resistance (F=4773kN) in the gravel case by taking cutters’ forces into account. 5

Figure 5. Theoretical upper and lower bounds associated with data curve in field for shield tunnel project (vertical axis: trust in kN and horizontal axis: rate of penetration in m)

3.2

Case II: Pipe jacking project in Taiwan

Another case study is presented for pipe jecking tunnel shown in Fig. 6.

Figure 6. Cutting head in field for pipe jacking project

Unlike a flat data curve in field for the case of shield tunnel, the in-situ data curve for pipe jacking method in Fig. 7 increases in trust (vertical axis) with the increase of rate of penetration (horizontal axis) due to the lateral resistance is proportional to the pipe length. In this cutting case of sewer system, there is no abnormal excavation situation such that the data curve does not reach the theoretical boundaries.

Figure 7. Theoretical upper and lower bounds with in-situ data curve for pipe jacking project of sewer system (vertical axis: trust in kN and horizontal axis: rate of penetration in m)

513

REFERENCES

Balci, C., "Correlation of rock cutting tests with eld performance of a TBM in a highly fractured rock formation: A case study in Kozyatagi-Kadikoy metro tunnel, Turkey", Tunnelling and Underground Space Technology, Vol.24, 2009, pp. 423 - 435. Chen, L. H., Failure of Rock under Normal Wedge Indentation, Ph. D. Thesis, University of Minnesota, U.S.A., 2001. Huang, H., Detournay, E., and Alehossein, H., "Analytical Model for the Indentation of Rocks by Blunt Tools," Rock Mechanics and Rock Engineering, Vol.33, No.4, 2000, pp. 267 - 284. Farrokh, E. and Rostami, J., "Correlation of Tunnel Convergence with TBM Operational Parameters and Chip Size in the Ghomroud Tunnel, Iran," Tunnelling and Underground Space Technology, Vol.23, 2008, pp.700 - 710.

th


Survey results of damaged areas in flood disaster of typhoon Morakot and suggestions for restoration projects Résultats des investigations sur les zones ravagées et inondées par le typhon Morakot, propositions de projets de restauration Chou J.C., Huang C.R.

Sinotech Engineering Consultants, Taipei, Taiwan

Shou K.J.

National Chung Hsing University, Taichung, Taiwan

ABSTRACT: Typhoon Morakot attacked Taiwan on August 8th, 2009 and caused heavy rainfall in Taiwan. The precipitation from August 5th to 10th, 2009 in the survey station of Ali Mountain was about 3049.5mm and the precipitation of 24-hour and 48-hour period in the south of Taiwan was close to the maximum observed precipitation in the world. This heavy rainfall caused many natural disasters including landslides, floods and debris flows. These natural disasters severely damaged the middle, south and east of Taiwan and cost huge property and life losses. Sinotech Engineering Consultants conducted a comprehensive survey after the flood disaster of Typhoon Morakot (1) to collect information of damaged areas, (2) to analyze the causes and mechanisms of failures and (3) to draft strategies and suggestions for restoration projects and future projects. Survey areas included the drainage basins of Chenyulan Creek, Laonong River, Cishan River and Ailiao River, the watershed of Nahua reservoir and Taitung area. Depending on different geological environments, the survey team concluded ten mechanisms causing failures. This article presents survey observations, discussions of failures in flood disaster and provides strategies and suggestions for restoration projects and future projects. RÉSUMÉ : Le typhoon Morakot a atteint Taiwan le 8 août 2009 accompagné de fortes pluies. Les précipitations du 5 août au 10 août 2009, mesurées à la station d’Ali Mountain, ont été de 3 049,5 mm, entraînant dans le sud de Taiwan des niveaux de précipitation en 24 heures et 48 heures proches des maximums de précipitations connues dans le monde. Les fortes pluies ont causé de nombreux cataclysmes naturels tels que glissements de terrains, inondations et charriage de débris, qui ont ravagé le Centre, le Sud et l’Est de Taiwan entraînant des pertes en vie humaines. Par la suite, le bureau d’ingénierie Sinotech a réalisé des études détaillées afin de dresser un bilan exhaustif des dégâts des inondations causées par le typhon Morakot : (1) recensement et investigation des zones ravagées ; (2) analyse des causes et des mécanismes des désordres, (3) ébauches de stratégies et solutions pour les travaux de restaurations et les projets futurs. Les zones d’investigations comprenaient les bassins versants de Chenyulan Creek, Laonong River, Cishan River et Ailiao River, le bassin hydrographique du réservoir Nahua et de la zone de Taitung. Suivant les différents environnements géologiques identifiés, les investigations ont conclu à dix mécanismes à l’origine des désordres. Cet article présente les résultats des investigations sur sites, l’établissement des mécanismes des désordres liés aux inondations qui en suivirent ainsi que les solutions de restaurations et les projets correspondants. KEYWORDS: Typhoon Morakot, Landslides, Debris Flows 1

failures and (3) to draft strategies and suggestions to restoration projects and future projects.

INTRODUCTION

Typhoon Morakot attacked Taiwan on August 8th, 2009 causing heavy rainfall in Taiwan. The precipitation from August 5th to 10th, 2009 in the survey station of Ali Mountain was about 3049.5mm and the precipitation of 24-hour and 48-hour period in the south of Taiwan was close to the maximum observed precipitation in the world. Figure 1 shows the maximum accumulated precipitation in 24 hours in the south of Taiwan. This heavy rainfall caused floods, many natural disasters and severely damaged mountain areas of the middle, south and east of Taiwan. All these floods and disasters are called “88” Flood. “88” Flood caused 643 deaths, 60 missing, 2,555 injured and 16.5 billion NT dollars economy lost which makes “88” Flood the worst natural disaster since Chi-Chi Earthquake happened on September 21, 1999. The most severe disaster is the disaster in Xiaolin village. The village was destroyed by a catastrophic debris flow during Typhoon Morakot. Over 500 residents were buried alive and 350 houses were damaged. Right after “88” Flood, Sinotech Engineering Consultants conducted a comprehensive survey on slopelands, roads, bridges and hydraulic facilities in the drainage basins of Chenyulan Creek, Laonong River, Cishan River and Ailiao River, the watershed of Nahua reservoir and Taitung area (See Figure 2). Purposes of this survey are (1) to collect information of damaged areas, (2) to analyze the causes and mechanisms of

—Rainfall Isoline

Figure 1. Maximum accumulated precipitation in 24 hours in the south of Taiwan (Precipitation in mm).

515

1


th

Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013 Drainage basins of Chenyulan Creek

County Line Drainage Basin Survey Area

Drainage basins of Cishan River Watershed of Nahua reservoir Drainage basins of Laonong River

Drainage basins of Ailiao River

Figure 3. Failure caused by Erosion of Bottom Roadside Slope Mechanism (Provincial Highway 21 (228K+500)).

Taitung Area

Figure 2. Survey area conducted by Sinotech Engineering Consultants survey team.

2

Figure 4. Failure caused by Debris Flow (County Road 179 in Nantou County).

CAUSES AND MECHANISMS OF FAILURES

Flood and heavy rain are two main reasons causing slopeland and road failures. Depending on different geological environments, the survey team concluded ten mechanisms of slopeland and road failures. Table 1 listed these mechanisms and numbers of failures caused by each mechanism. Figure 3 to Figure 12 show failures caused by Mechanism No.1 to No.10. One mechanism should be noticed is Complex Failure Mechanism. Complex Failure Mechanism means that slopeland or road failures were caused by more than one mechanism simultaneously. Damages caused by Complex Failure Mechanism were usually severe and massive. Figure 12 shows one failure site (County Road 64 in Taitung County) damaged by Complex Failure Mechanism which included Erosion of Bottom Roadside Slope and River Channel Erosion of Roadbed mechanisms.

Figure 5. Failure caused by Erosion of Top Roadside Slope (County Road 60 in Nantou County).

Table 1. Failure Mechanisms and Number of Failures No. 1 2 3 4 5 6 7 8 9 10

Failure Mechanism Erosion of Bottom Roadside Slope Debris Flow Erosion of Top Roadside Slope Shallow Slope Failure River Channel Erosion of Roadbed Dip slope Circular Failure of Slope Barrier Lake Deep Sliding Failure of Colluvium Complex Failure Mechanism

# of Failures 35 33 27 22 21 4 4 2 2 15 Figure 6. Failure caused by Shallow Slope Failure (River in Tao Yuan District in Kaohsiung City).

516



Colluvium Failure Area

Chenyulen Creek

Figure 7. Failure caused by River Channel Erosion of Roadbed (Provincial Highway 21).

Figure 11. Failure caused by Deep Sliding Failure of Colluvium (Xin Fa Village in Kaohsiung City).

Dip Slope

Erosion of Top Roadside Slope River Channel Erosion of Roadbed Figure 8. Failure caused by Dip slope (County Road 179-1 in Nantou County).

≒ 50m

≒ 205m

Figure 12. Failure caused by Complex Failure Mechanism (County Road 64 in Taitung County).

In general, Mechanism No.1, No.3 and No.5 were main mechanisms of road failures and Mechanism No.2, No.4 and No.5 were main mechanisms of slopeland failures. These five mechanisms caused most of failures in “88” Flood.

Circular Failure Figure 9. Failure caused by Circular Failure of Slope (At 3K of County Road 179 in Nantou County).

Figure 10. Failure caused by Barrier Lake (County Road 179-1 in Nantou County).

3 STRATEGIES AND SUGGESTIONS TO RESTORATION PROJECTS AND FUTURE PROJECTS From survey observations and analyses, strategies to prevent future slopeland and road failures can be outlined starting from following aspects: (1) debris flow control and river remediation, (2) proper treatment of landslides, (3) soil and water conservation of slopeland and (4) soil and water conservation of road. Furthermore, improvements of road designs can be done to avoid slopeland and road failures in the design phase: (1) consider and prevent all possible failure mechanisms, (2) use rock shed in potential rock fall area (see Figure 13), (3) from survey observations, the tunnel and bridge are the best solution for roads in potential debris flow area (see Figure 14), (4) use deep foundation (e.g. pile foundation) as retaining wall at bottom roadside slope retaining wall to avoid erosion. If the roadbed can be protected from erosion, it is easier and faster for future road restorations. (5) avoid area where river channel erosion occurs and (6) install proper water drainage system for slopelands (see Figure 15).

517



4

CONCLUSIONS

Because of the global climate change, natural disasters caused by the extreme weather become more frequently in last few years. The magnitude of these disasters and the damage caused by these disasters become greater and greater and are greater than what we experienced in the past. Therefore, in future engineering projects, engineers should consider uncertainties caused by the extreme weather in their designs and understand limitations of engineering techniques. In addition, engineers should try to avoid unnecessary development and construction in environmental sensitive area. 5

ACKNOWLEDGEMENTS

Authors would like to thank Taiwan Geotechnical Society (TGS) for giving this great opportunity to present our work in the 18th International Conference on Soil Mechanics and Geotechnical Engineering. Authors would also like to thank engineers who involved in this survey project and supports from Sinotech Engineering Consultants.

Figure 13. Rock shed used to prevent landslides and rock fall area.

6

REFERENCES

SINOTECH Engineering Consultants, 2010. Survey results of flood disaster of typhoon Morakot and suggestions to restoration projects, SINOTECH, Taipei, 334p

Figure 14. Bridge used to avoid damages from Debris Flow in Wanrung Township, Hualien County (Provincial Highway No. 16).

Building Ditch Road

Water Table Level

Dewatered Water Table Level

Drainage Pipe

Drainage Well

Slip Surface Drainage Gallery

Creek or River Figure 15. Different types of water drainage systems used in slopelands.

518

Stability of chilean’s tailings dams with the Panda® penetrometer. Experiences of the last 10th Dix ans d’études de la stabilité des barrages de résidus miniers chiliens à l’aide du pénétromètre Panda® Espinace R., Villavicencio G., Palma J.

Grupo de Geotecnia. Escuela de Ingeniería en Construcción. Pontificia Universidad Católica de Valparaíso, Chile. Geotecnia Ambiental, Chile.

Breul P., Bacconnet C.

Institut Pascal – Polytech’Clermont-Ferrand. Université Blaise Pascal, Clermont-Ferrand, France.

Benz M.A., Gourvès R.

Sol-Solution Géotechnique Réseaux, Riom, France. ABSTRACT: In Chile, since the beginning of the 20th century, about 40 cases of mechanical instability of the tailing dams have been reported mainly due to liquefaction, slipping of banks or settlement. In order to solve this problem, a scientific and technological cooperation has been established in 2001 between the geotechnical of groups at the Catholic University of Valparaíso (Chile) and Blaise Pascal University Clermont-Ferrand (France) whit the support of two companies, Sol-Solution in France and GeotecniaAmbiental in Chile. This article presents the main results that have allowed to propose a methodology for control and diagnosing of tailing dams and its application in the medium mining sector. It is based on in-situ determination of geomechanical parameters (internal friction angle and density index) using the Panda® lightweigh penetrometer in order to characterize the constituent materials, the variability of these materials and their implementation in the works. Finally, this methodology allows taking into account this variability in the study of stability and the risk of liquefaction of these structures in a probabilistic approach. RÉSUMÉ: Au Chili, depuis le début du XXe siècle, environ 40 cas d'instabilité mécanique de ces dépôts, principalement par liquéfaction, glissement des talus et tassements, ont été rapportés. C’est dans ce contexte et pour apporter une réponse à ce problème, qu’une coopération scientifique et technologique a été établie en 2001 entre les groupes de géotechnique de l’Université Catholique de Valparaiso (Chili) et de l’univeristé Blaise Pascal Clermont-Ferrand (France), avec le soutien des entreprises Sol-Solution (France) et Geotecnia Ambiental (Chili). Cet article présente les principaux résultats qui ont permis de proposer une méthodologie pour le contrôle et le diagnostic des barrages de résidus miniers de relave ainsi que son application au secteur de l’industrie minière moyenne. Elle s’appuie sur la détermination in situ des paramètres géomécaniques (angle de frottement interne et densité relative) à l’aide du pénétromètre Panda® en vue de caractériser les matériaux constituants, de la variabilité de ces matériaux et de leur mise en œuvre au sein des ouvrages. Finalement, cette méthodologie permet de prendre en compte cette variabilité pour l’étude de la stabilité et du risque de liquéfaction de ces ouvrages dans une approche probabiliste. KEYWORDS: soils and site investigation, structures in seismic areas. 1

the stability of these dams, and its variability from dynamic penetration tests. Then models are proposed for all dams composed of the same mine tailings types, making it possible to link a probability law to the calculation parameters ’ and ID%. This method, applied to Chilean dams constructed from copper mine tailings, proposes a single model for all tailings dams so as to associate a probability law to the ’ and ID%.

INTRODUCTION

Mine tailings are frequently stored in dams. This is the case for copper for which the coarse fraction (fine sands) of the tailings form the body of the dams, while the fine saturated fraction (sludge and silts) is poured by cycloning into the reservoirs of the dams thus formed. Chile has a very large number of tailings dams built in this way. Due to the construction methods and materials used, these dams comprise failure mechanisms such as loss of stability, liquefaction, and internal and external erosion leading to major risks for the populations and their environments. Such risks are highlighted by the accidents that have occurred around the world and recently in the case of failures occurring during the earthquake of 27 february 2010 in Chile, with fatal consequences (Dobry and Alvarez 1967, ICOLD 2001, GEER 2010). In order to manage these risks, it appears necessary to employ a probabilistic approach to predict their behaviour during construction and after closing. However, applying such an approach in practice at present is limited by the difficulty of managing the data (random variables and stochastic fields) to be introduced in the reliability calculations for the limit conditions involved and conditioned by the relevance of the probability models chosen to represent the variability of tailings dam properties (Villavicencio et al. 2011). This is the reason why, this article presents an approach of estimating calculation parameters (friction angle ’ and density index ID%) governing

2 ESTIMATION OF THE DENSITY INDEX (ID%) AND THE FRICTION ANGLE (’) 1.1

The objective

In mine tailings with non plastic fine particles (size < 80 µm) ID% and ’ are very important parameters, related to the in situ penetration strength (N, qd, qc, etc), the input parameter of static and dynamic stability models and for the evaluation of the liquefaction (Troncoso 1986). These parameters are greatly influenced by the origin and mineralogy of the particles, by the physical characteristics and state of arrangement of the grains determined by the state of compacting and by the extent of stresses in-situ (Bolton 1986). The methods used to implement mine tailings lead to the prevalence of stratified internal structures that can be heterogeneous. This can result in variations of resistance properties, especially ’ and ID%, as a function of depth. Thus it is important to estimate the values and variability of these

519


b) Performing dynamic cone resistance tests in a calibration mould for different states of density to obtain the relation d/qd (calibration curve). A logarithmic relation can be observed, in agreement with previous results (Chaigneau et al. 2000) for this type of material. Figure 1 gives the calibration curves d/qd obtained for dams No. 1, No. 2 and No. 3. c) Normalisation of qd at atmospheric pressure (equation 1).

parameters. To do this, we propose an estimation method based on measuring the dynamic cone resistance (qd) that can be relatively easily measured on this type of structure. 1.2

Normalisation of qd

Estimating ID% and ’ by using empirical and semi-empirical relations, first implies normalising qd at a reference stress corresponding to atmospheric pressure (pa), using the following equation 1.



c

(1)

where: qdN1 is the dimensionless normalised dynamic cone resistance, qd is the dynamic cone resistance, pa is the atmospheric pressure, ’v is the effective vertical stress, “c” is the normalisation coefficient (0.5 to 0.75).

d = 1,086ln(qd) + 15,543 R² = 0,9883

d = 0,8693ln(qd) + 15,552 R² = 0,9755

16,5 16,0

Tailings Dam No 1

15,5

Tailings Dam No 2

15,0

Tailings Dam No 3

2,0

4,0 6,0 8,0 10,0 12,0 14,0 Dynamic cone resistance, qd (MPa)

16,0

18,0

Figure 1. Relation d/qd for tailings dams No. 1, No. 2 and No. 3 in the study.

1.3.1 Relation ID% = f (qdN1) The equivalence between the state of density (% Optimum Proctor Normal) and ID% was estimated for each calibration test. On the basis of the normalised cone resistance (qdN1), and by considering the classification modified by Skempton (1986) and adapted by Villavicencio (2009), we estimated ID% associated with each degree of compaction (table 2). Table 2. Estimation of the state of compaction and associated mechanical behaviour for silty sands. Villavicencio (2009).

Table 1. Geotechnical properties of mine tailings. Values and statistical analyses of experimental data from three representative tailings dams. No. 2

17,0

0,0

Our study is based on the use of cone penetration resistances (qd) obtained by using the Panda test. The Panda device is a manual light dynamic penetrometer with variable energy and a small cone section (2.0 or 4.0 cm2) (Gourvès et al. 1997, Benz 2009). The Panda provides the cone resistance qd of the soil as a function of depth, and is capable of performing a large number of in situ tests thanks to its small size and its quick implementation. This device can operate until 6.0 (m) in depth and for materials having particles size lower than 50.0 (mm).

No. 1

17,5

14,0

Experimental approach

Geo.

18,0

14,5

According to Moss et al. (2006), this reference stress value is considered as reasonable if the depth/stress relation is taken into account. According to Salgado et al. (1997) and Moss et al. (2006), the normalisation coefficient is not only linked to the intrinsic properties of the soil such as the type of grain and the physical characteristics of the material (mineralogy, granulometry, particle shape and texture characteristics), lateral pressure (Ko), compressibility, cementation, resistance to crushing of the particles, etc. 1.3

d = 1,0811ln(qd) + 15,983 R² = 0,9948

18,5

Dry density (KN/m3)



with C q p a  'v

qd N1 qd  Cq

19,0

No. 3

State of

Mechanical

Liquefaction

compaction

behaviour

potential

0 – 15

Very low

Contractant

High

17 – 69

15 – 55

Low

Contractant

High

69 – 82

55 – 60

Average

qdN1

ID%

0 – 17

Contractant /Limit

Limit

Prop

Av.

CV

Av.

CV

Av.

CV

82 – 162

60 – 80

Dense

Dilatant

Null

s

3.09

4.6

3.36

8.0

3.1

2.2

162 – 326

80 – 100

Very dense

Dilatant

Null

D50

0.13

19.0

0.11

15.2

0.25

8.7

F.C

28.0

28.7

33

26.3

17

10.0

IP

0

0

0

0

0

0

dmax

18.2

6.2

20.8

8.0

18.5

2.3

d

17.5

6.6

20.1

8.2

18.1

2.9

wnat

11.0

22.3

3.3

43.1

7.5

27.3

qd

4.8

50.6

2.87

45.9

1.95

52.8

N60

22

62.5

12

58.8

-

-

3

s: specific weight (kN/m ), D50: median diameter (mm), F.C: percentage of fines less than 80 (µm), IP: plasticity index (%), dmax: Proctor dry density (kN/m3), d: dry density in situ (kN/m3), wnat: water content in-situ (%), qd: cone resistance PANDA test (Mpa), N60: corrected penetration resistance index, Av: average, CV: coefficient of variation (%).

A serie of Panda tests have been performed on the mine tailings coming from three dams studied, under controlled laboratory conditions in a calibration chamber. The following procedure was used: a) Determination of the physical characteristics of 3 samples of mine tailings of copper sulphates (Table 1).

Studies conducted by Troncoso (1986) have concluded that for mine tailings with a percentage of fines around 15% , with confining stresses between 50 kPa and 350 kPa, ID% below 50%-60% is an indicator of contractancy. Under this condition, if the material is saturated or partially saturated, under seismic conditions, the risk of liquefaction is real. On the other hand, the material will tend to a dilatant behaviour for a relative density over these values. Verdugo (1997) have conducted an analysis of the variation of the minimum and maximum densities (Vibratory and Proctor compaction) both with mine tailings and similar soils (sands and silts) with different percentage of fines. They conclude that in situ ID% of 60% is a very reasonable compaction value with a satisfactory mechanical behaviour (dilatancy) in structures that allow certain degree of deformation such as the tailing dams. An empirical model was adapted by using a simple regression on all the pairs of experimental data (qdN1, ID%) for the three samples of mine tailings. Since we consider that mine tailings can be globally classified in a single geotechnical class, it is possible to estimate ID% as a function of the resistance qdN1 by a single relation. The model used is given by the following equation: with 10.0 ≤ qdN1 ≤ 326.0 (2)  ID% 28.5  lnqd N1   65.4

520


Figure 2 shows that the results of the model are very close to the experimental results. In addition, the relation proposed by Tatsuoka et al. (1990) was used by replacing qcN1 by qdN1.

1.4

Application case: density index (ID%) and associated mechanical behavior

On the basis of equation 2, it is possible to estimate the profiles of the density index (ID%) as a function of depth from the penetrometric tests performed in situ. The adaptation of the correlation presented in table 2, allows estimating the mechanical behaviour of mine tailings as a function of ID%. At global scale (measurements processed at the scale of the tailings dam by using the ID% distribution obtained from all the penetration tests performed), the distribution of all these ID% values for each dam can be adjusted by a normal law (Figure 4).

Figure 2. The experimental points, relations proposed and references for estimating the ID% of mine tailings as a function of qdN1.

1.3.2 Relation’ = f (qdN1) Since we considered that mine tailings can be globally classified within one geotechnical class, it is possible to estimate ’ as a function of the resistance of qdN1 by a single relation. To do this, a regression analysis was performed on all the pairs of experimental data (qdN1, ’) obtained during the calibration tests, for the three samples of mine tailings (figure 3). The model used is given by the following equation:   ' 14.79  5.54  lnqd N1  with

10.0 ≤ qdN1 ≤ 280.0

(3)

As it can be seen on figure 3, the results of the model are very close to the experimental results. In addition, the relation proposed by Díaz and Rodríguez-Roa (2007) was used by replacing qcN1 by qdN1.

Figure 4. Density function of ID%. Proposed relation for tailings dams No. 1, No. 2 and No. 3 in the study.

At global scale, the density function makes it possible to obtain a global idea of the mechanical behaviour of the mine tailings stored, by considering the limit value of ID%, which permits classifying contractant or dilatant behaviour and associate in a qualitative way the liquefaction potential. As an example, table 3 presents a probabilistic analysis in global scale of the variability of ID% and the mechanical behaviour for the tailing dam No.1. Table 3. Density index (ID%) and associated mechanical behaviour. Analysis at global scale. Tailings dam No. 1. Analysis of the mechanical behavior

ID% Av.

52 Figure 3. Experimental points, proposed and bibliographic relations for estimating ’ of mine tailings as a function of the qdN1.

This result is in full agreement with the works already carried out on the correlation between qc and qd obtained with a Panda penetrometer. Indeed, it has been proven (Chaigneau et al. 2000, Lepetit 2002) that in the case of sands and silty sands, the average value obtained for the ratio qd/qc is equal to 1.03. More recent research performed by Rahim et al. (2004) confirmed the relation between qd and qc. Their results obtained for granular soils have been demonstrated experimentally and analytically on the basis of the cylindrical cavity expansion theory and that of cavitation collapse. The resistance qd obtained with a light Panda penetrometer can therefore be assimilated with qc. In conclusion, in the case of mine tailings: (1) density index (ID%) and effective friction angle (’) can be deduced very precisely from the normalised cone penetration resistance qdN1 by a two single relations, (2) relation qdN1 = qcN1 is very well validated which allows using either static or dynamic penetrometers according to need.

521

C.V %

28.3

ID%

State of Mechanical Liquefaction % of values compaction behaviour potential

< 55

58

55 – 60

13

60 – 100

29

Low

Contractant

Average

Contractant

density

/Limit

Dense to very dense

Dilatant

High Limit Null

At a local scale (measurements processed at the scale of each penetration test, by using the ID% distribution), the distribution of all these ID% values can then also be adjusted by a normal law (figures 5a, 5b). The so-obtained results are consistent with the compaction test performed during the construction of the three tailings dams. The results are similar for the three dams, they show that a local test can be used to estimate ID% for each penetration test, with sufficient precision provided that the calibration tests have been carried out on the material characteristics of the dam at the scale of the structure concerned. The variability of ID% and the soil mechanical behaviour associated, allows to estimate in a first stage, the liquefaction potential of tailings dams in both scales, global and local, and identify in a local scale the zones with lower strengths through a layer by layer penetration test (Figure 6). The evaluation of the risk of liquefaction has been expressed in an equation formulated by Seed and Idriss (1981). The classical method compares locally the ratio of the cyclic resistance of the soil (CRR) with the ratio of the cyclic shearing stress ratio (CSR) stemming from seismic stress. The notion of liquefaction potential is therefore linked to the fact that ratio


collaboration from the Professor, Mr Pierre Foray, Laboratory 3S-R, Institut National Polytechnique de Grenoble, France.

Density Function

CRR/CSR is lower than unity. It is widely accepted that estimating the cyclic resistance ratio (CRR) can be estimated on the basis of dynamic and static penetration tests (Robertson and Wride 1998, Boulanger 2004 and Idriss, etc.).

a)

2

Scale Global Local

0,04 0,03 0,02 0,01 0

0

10

20

30

40

50

60

70

80

Index Density (ID%)

90

100

b)

Figure 5. a) The breakdown into layers and density index (ID%). b) The distribution of Density Index (ID%). Test No. 1. Tailings dam No. 1.

Figure 6. Example of the factor of safety (F.S) profile. Test No. 1. Tailings dam No. 1.

3

CONCLUSIONS

To predict the behaviour of mine tailings dams in view to managing the risks inherent to them, it appears necessary to carry out a probabilistic approach However, in practice implementing this type of approach is limited by the difficulty of managing the data to be used in reliability calculations for the limit conditions concerned. This article proposed a method for estimating in situ the density index (ID%) and the effective friction angle (’) and its variability, making it possible to carry out a probabilistic study of these structures. A single model was proposed for all the mine tailings dams in Chile, in view to linking a probability law to ID% and the ’. A method was proposed that takes into account the spatial variability of data for performing a reliability calculation of liquefaction potential, which is the main cause for the failure of this type of structure. On the basis of the results obtained, we showed that the method proposed for estimating liquefaction potential permits evaluating the probability of triggering this phenomenon. Estimating the reliability of a dam in relation to the limit states of static and dynamic stability demonstrates the advantages and applicability of the approach, by using the variability of the geotechnical characteristics of mine tailings and resistance to penetration (qdN1) in particular. 4

ACKNOWLEDGEMENTS

Fundings for the work described in this paper was provided by the research department of the Pontifical Catholic University of Valparaiso Chile. This article was developed with the important

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REFERENCES

Benz M.A. 2009. Mesures dynamiques lors du battage du pénétromètre Panda 2. Ph. D. Thesis, Blaise Pascal-Clermont II Univ, France. Bolton M. 1986. The strength and dilatancy of sands. Géotechnique 36 (1), 65-78. Boulanger R. and Idriss I.M. 2004. State normalization of penetration resistance and the effect of overburden stress on liquefaction resistance. Proceedings 11th SDEE and 3rd ICEGE, Berkeley, CA, 484- 491. Chaigneau L. Bacconnet C. and Gourvès R. 2000. Penetration test coupled with geotechnical classification for compacting control. An International Conference on Geotechnical & Geological Engineering, GeoEng2000, Melbourne, Australia Díaz E. and Rodríguez-Roa F. 2007. Ensayos in-situ en Arenas. VI Chilean Congress of Geotechnical Engineering. Chilean Society of Geotechnics. Univeridad Católica de Santiago. Chile, November, 28-30. Dobry R. and Alvarez L. 1967. Seismic failures in chilean tailings dams. J. Soil Mech. & Foundation Eng. ASCE, SM6 (93), 237-260. ICOLD. 2001. Tailings dams. Risk of dangerous occurrences. Lessons learnt from practical experiences. Bulletin Nº 121. UNEP, DTIE and ICOLD, Paris. GEER (Geo-Engineering Extreme Events Reconnaissance Association) 2010. Dams, levees, and mine tailings dams. Turning disaster in knowledge: geo-engineering reconnaissance of the 2010 Maule, Chile Earthquake. J. Bray and D.Frost, Eds., 204-226. Gourvès R. Oudjehane F. and Zhou S. 1997. The in situ characterization of the mechanical properties of granular media with the help of penetrometer. Proceedings of 3rd International Conference on Micromechanics of Granular Media, Powders and Grains, Duram, USA, 57-60. Lepetit, L. 2002. Etude d’une méthode de diagnostic de digues avec prise en compte du risque de liquéfaction. Thesis, Blaise PascalClermont II Univ, France Moss R.E. Seed R.B. Kayen R.E. Stewart J.P. and Der Kiureghian A 2006. CPT-Based probabilistic assessment of seismic soil liquefaction initiation. PEER Report 2005/15. Rahim A. Prasad SN, and George K.P. 2004. Dynamic cone penetration resistance of soils-theory and evaluation. Proceedings of the GeoTrans 2004 Conference, Los Angeles, California. Robertson P.K. and Wride C.E. 1998. Evaluating Cyclic Liquefaction Potential Using The Cone Penetration Test. Canadian Geotechnical Journal, 35 (3). 442-459. Salgado R. Boulanger R. and Mitchell J. 1997. Lateral effects on CPT liquefaction resistance correlations. J. of Geotechnical and Geoenvironmental Engineering, ASCE, 123 (8). 726-735. Seed H.B. and Idriss I.M. 1981. Evaluation of liquefaction potential of sand deposits based on observations and performance in previous earthquakes. In Situ Testing to Evaluate Liquefaction Susceptibility, ASCE Annual Convention, St. Louis. Skempton S.M. 1986. Standard penetration test procedures and the effects in sands of overburden pressure, relative density, particle size, aging and overconsolidation. Geotechnique 36 (3). 425-447. Tatsuoka F. Zhou S. Sato T. and Shibuya S. 1990. Evaluation method of liquefaction potencial and its application. Report on Seismic Hazards on the Ground in Urban Areas. Tokyo. 75-109. Troncoso J. 1986. Envejecimiento y estabilidad sísmica de un depósito de residuos minerales en condición de abandono. ISSN-0716-0348. (22), 147-158. Verdugo R. 1997. Compactación de Relaves. IV Chilean Congress of Geotechnical Engineering. Chilean Society of Geotechnics, Santiago. Universidad Federico Santa María. Chile. October, 29-4. Villavicencio G. 2009. Méthodologie pour evaluer la stabilite des barrages de résidus miniers. Ph. D. Thesis, Blaise Pascal-Clermont II Univ, France. Villavicencio G. Bacconnet C. Breul P. Boissier D. and Espinace R. 2011. Estimation of the Variability of Tailings Dams Properties in Order to Perform Probabilistic Assessment. Geotechnical and Geological Engineering. 29 (6). 1073-1084.

Site Sampling: Assessing Residual Uncertainty Échantillonnage du site : évaluation de l'incertitude résiduelle Fenton G.A.

Faculty of Civil Engineering and Geosciences, Delft University of Technology, Delft, The Netherlands Department of Engineering Mathematics, Dalhousie University, Halifax, Nova Scotia, Canada

Hicks M.A.

Faculty of Civil Engineering and Geosciences, Delft University of Technology, Delft, The Netherlands

ABSTRACT: Geotechnical design is plagued by the uncertainty associated with site characterization. Common questions are “How many samples should be taken?” and “How do these samples reduce my uncertainty?” Of considerable interest is the question “What site sampling plan will give the best cost to effectiveness ratio?” This papers looks specifically at the effect of the number of samples on residual uncertainty. The results can be used to quantitatively select the required number of samples needed to achieve a target maximum residual uncertainty level. To study this problem, a square domain is selected (the site) and a stationary Gaussian random field is simulated within the domain (the random soil properties). The random field is sampled at a series of locations and a trend is estimated from the samples. The trend is then removed from the random field and the residual random field is statistically analyzed to determine various measures of the effectiveness of the sampling scheme. These measures include: 1) the variance of the residual field average (i.e. does the estimate represent the average?), 2) the residual standard deviation (i.e. how much residual uncertainty remains?), and 3) the residual correlation length (i.e. how does trend removal affect the perceived correlation lengths?). RÉSUMÉ : Le design géotechnique est traditionnellement affecté par des incertitudes associées à la caractérisation du site. Les questions les plus courantes sont : combien d’échantillons devraient être prélevés ? Comment ces échantillons peuvent réduire mon incertitude ? Un des intérêts les plus importants vient de cette question. Quel plan d’échantillonnage du site donnera le meilleur coefficient d’efficacité? Cet article examine spécifiquement l’effet du nombre d’échantillons sur des incertitudes résiduelles. Les résultats peuvent être utilisés pour quantifier et sélectionner le nombre demandé d’échantillons nécessaires pour atteindre un objectif d’incertitude maximal avec le niveau résiduel. Pour étudier ce problème, un domaine carré est sélectionné (le site) et un champ gaussien aléatoire stationnaire est simulé dans le domaine (les propriétés du sol aléatoires). Le champ aléatoire est échantillonné à une série d’emplacements et une tendance a été estimée à partir de l’échantillon. La tendance retirée du champ aléatoire et le champ résiduel aléatoire est statistiquement analysées afin de déterminer les mesures diverses de l’efficacité du plan d’échantillonnage. Ces mesures comprennent : 1) la variance de la moyenne de champ résiduel, c’est à dire comment la tendance estimée représentent la moyenne réelle sur le terrain ? 2) l’écart type résiduel, c’est-à-dire à quel degré d’incertitude résiduelle demeure, et 3) la valeur longueur résiduelle de corrélation, c’est-à-dire comment la suppression tendance affecte les longueurs de corrélation ?. KEYWORDS: geotechnical design, site characterization, residual uncertainty, sampling, required number of samples, sampling plans. 1

INTRODUCTION

Site characterization is clearly an essential component of any geotechnical design and a great deal of effort has been devoted over recent decades on how to best perform such a characterization. How many samples should be taken? How should these samples be used in the design process? The ground is one of the most complex of engineering materials, and yet is the most fundamental, in all senses of the word. While steel, concrete, and wood, for example, have fairly well established and relatively small uncertainties, the ground can vary by orders of magnitude from site to site, and even within a site. As a result of the large uncertainty in the ground, all geotechnical designs must start with a geotechnical investigation so that the best “nominal” or “characteristic” ground parameters can be used in the design process. Traditionally, the intensity of the site investigation has not been particularly important, so long as a reasonable estimate of the characteristic design values can be estimated. However, recent impetus has been towards providing reasonable estimates of the reliability of designed geotechnical systems. In order to do so the ground used to provide the geotechnical resistance needs to be properly evaluated, in both the mean and the covariance. In this paper, the ability of a soil sampling scheme to predict the actual mean, variance, and correlation length of the soil at a site is investigated. A key question is how does the number of samples affect the accuracy of the estimate? Or, put another

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way, how many samples are required to achieve a certain desired accuracy? The answer is found by considering a square site and using random field simulation to generate realizations of the soil properties over the site, sampling each realization, and then comparing the estimated mean, variance, and correlation length to the ‘true’ values. The goal here is to investigate the discrepancies between the estimated statistics and the true ‘local’ statistics, the latter obtained by sampling the field at all locations. Note that the ‘local’ statistics will differ from the population parameters,  (mean),  (standard deviation), and  (correlation length), which are used by the random field generator, due to the fact that the local statistics are derived from a single realization. In detail, the soil is represented by a stationary Gaussian random field, X  x  , at spatial position x , which is simulated within the domain and sampled at ns locations. The samples are then used to estimate a mean trend, ˆ  x  , which can then be compared to the field realization to assess its ability to represent the actual mean trend. Defining the residual to be (1) X r ( x )  X ( x )  ˆ ( x ) then ˆ  x  is a good estimate of the mean trend if X r is generally small. If the site is sampled at all locations, then ˆ  x  can be taken to be equal to X  x  , in the event that a pointwise trend is assumed (as in Kriging), in which case X r  x   0 everywhere. Sampling at all locations is the best case since there is then minimum residual uncertainty (zero in the case of Kriging).


Sampling at all locations is, of course, prohibitively expensive and would also change the resulting field properties while measuring them (see, e.g., Heisenberg, 1927). In practice, soil properties are estimated from a relatively small number of samples so that ˆ  x  will only ever approximate X  x  in some way (i.e., via a trend). In assessing the ability of ˆ  x  to represent X  x  , it will also be useful to consider the average residual over the domain, 1 1 n  r  X r  x  dx   X  xi   ˆ  xi  (2)  D  D D D n i 1  where D is the edge dimension of the D  D square domain. The domain is broken up into n cells in the simulation, resulting in the summation form on the right, in which x i is the location of the center of the i ’th cell. The agreement between ˆ  x  and X  x  will be determined here by considering three measures; 1) the standard deviation of the residual field average,  r (i.e., how well does the estimated trend represent the actual field average?), 2) the standard deviation of the residual, X r (i.e. how much residual uncertainty remains?), and 3) the residual correlation length (i.e. how does the trend removal affect the perceived correlation lengths?). Five sampling schemes are considered in the paper, ranging from a single sample taken at the field midpoint to nine samples taken over a 3 x 3 array at the quarter points of the field. In some cases a further ‘maximum' sampling scheme is performed, where every point in the field is sampled, to see what the maximum attainable uncertainty reduction is. For each sampling scheme, three types of trend removal are performed; a) removing the constant sample mean, b) removing a bilinear trend surface which is fit to the sample, and c) removing a Kriged surface fit to the sample. The residual statistics are determined by Monte Carlo simulation, with 2000 realizations for each case, where the field is discretized into 128 x 128 cells and the random fields generated using the Local Average Subdivision method (Fenton and Vanmarcke, 1990). 2

RESULTS

Consider first the average of the residual,  r , given by Eq. 2. It can be shown that the mean of  r is zero, so that a measure of how accurately ˆ  x  represents X  x  can be obtained by looking at the standard deviation of  r – small values of this standard deviation imply that ˆ  x  remains close to the field average. Figure 1 illustrates how the standard deviation of  r , normalized by dividing by the standard deviation of the random field value, X  xi  , in the i ’th cell (referred to as  cell ), varies as a function of the number of samples taken from the domain, ns , and the normalized correlation length,  / D . Note that if only one sample is taken at the midpoint of the domain, ns  1 , then a bilinear trend cannot be fit to the sample, nor is a Kriged surface removal attempted. Thus, parts b and c in Figure 1 do not have a curve corresponding to ns  1 . In all plots it is apparent that as the number of samples increases, the accuracy improves (in agreement with the findings of Lloret-Cabot, et al., 2012). It can be seen, however, that for ns  3 to 9, there is very little difference between the detrending methods, so far as the field average is concerned. It is to be noted that the field average is a constant, not a trend, so it is not expected that the bilinear and Kriged surface trends will do any better than the sample mean, when compared to the field average.

Figure 1. Standard deviation of the field average residual (eq. 2), normalized by the standard deviation of X, versus normalized correlation length.

In all cases in Figure 1, the agreement between ˆ  x  and X  x  improves as the correlation length increases. This is because the field becomes increasingly smooth, or flat, as the correlation length increases, so that all trends considered become closer to the flatter X  x  .

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Figure 2. Standard deviation of the residual (eq. 1), normalized by the standard deviation of X, versus normalized correlation length.

A possibly better measure of how well ˆ  x  represents the field is obtained by considering the standard deviation of the residual, X r  x  (see eq. 1), directly. This measure will include the effects of trend removal and is illustrated in Figure 2, again with the standard deviation of the residual,  r , divided by the standard deviation of X ,  cell . In detail, the standard deviation of the residual is estimated as the square root of the variance, 2 1 n  r2  X  x i   ˆ  xi   (3)  n  1 i 1

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for each realization. The value of  r used in Figure 2 is averaged over all realizations. As in Figure 1, the n s  1 case only appears in Figure 2a, since bilinear trend and Kriging surfaces are not well defined for only one sample point. However, Figures 2a and b now include a limiting case where the entire simulation has been sampled ( ns  all), representing the best site knowledge possible. This case was not included in Figure 1 since, when all values are sampled,  r  0 , that is, the average residual is zero. In Figure 1, this would have corresponded to a horizontal line at zero standard deviation. In Figure 2, the ‘ ns  all’ case corresponds to the classical case where both the estimated mean (trend) and the variance are computed from the same set of observations. As the correlation length decreases, these observations become increasingly independent, and the estimated standard deviation approaches the true standard deviation, so that  r /  cell  1.0 as seen in Figures 2 a and b when ns  all. In Figure 2 c, the case ‘ ns  all’ is not included in the Kriging surface case since, when the entire field is sampled, the residual is zero with zero variability, and so the curve corresponding to this case lies at zero. As in Figure 1, Figure 2 also shows that the ability of ˆ  x  to represent X  x  improves as the correlation length increases, for all of the trends considered. In the limit, as  / D   , all random fields become uniform (under the assumed finite variance correlation structure), random from realization to realization, but constant within each realization. In this limiting case, the sample perfectly predicts the uniform field, and the residual becomes zero everywhere so that  r  0 . It is apparent in Figure 2 that all curves are heading towards 0, as  / D   . One of the perhaps surprising results of Figure 2 is that the removal of a bilinear trend is not generally as good as the removal of the constant sample mean at smaller correlation lengths, and especially at a lower number of samples. The reason for this becomes apparent when, for example, the case where ns  3 is considered. If the correlation length is small, then the three samples will be largely independent, and the resulting fitted bilinear plane could (and often does) end up with quite an unrepresentative slope, leading to a high variability in the residual. Even when ns  9 the residual variability is higher at low correlation lengths than seen using the constant sample mean. At low correlation lengths, the Kriging surface performs about the same as the constant sample mean. At large correlation lengths, e.g.  / D  10 , the bilinear trend performs better than the constant sample mean for all ns except ns  3 , where the relative standard deviation is 0.35 versus 0.32 for the constant sample mean. For higher number of samples, the relative standard deviation using the bilinear trend is 0.25, versus 0.31 for the constant sample mean. The Kriged surface performs the best out of the three methods (relative standard deviation of 0.30) when the number of samples is 3, and about the same as the bilinear trend for higher numbers of samples. The last measure of the quality of the trend type used considered in this paper is how well the estimated correlation length agrees with the actual correlation length, Figure 3. Once ˆ  x  has been established from the soil samples, the correlation length is estimated here using the following steps; 1. for each direction through the soil domain, i  1, 2 , 2. estimate the semi-variogram along all lines through the domain in direction i using the entire X r  x  field, 3. average the semi-variograms obtained in step 2 to obtain the final semi-variogram estimate in direction i , 4. fit a theoretical semi-variogram, having parameter  (correlation length), to the semi-variogram estimated in step 3 by minimizing the sum of squared errors (i.e. regression).


In general, when   D the estimated correlation length is overestimated, and often considerably overestimated, especially when the actual correlation length is small. This occurs because errors between the estimated trend (of any of the three types) and actual bilinear field trend (bilinear because correlation is a measure of the degree of linear dependence between random variables) are perceived in the estimation process to be caused by a strong lingering correlation (and not by an error in the original trend estimate) – hence a longer correlation length is estimated to account for the evident residual trend. Of the three trend types considered, the best is the constant sample mean and the worst is the bilinear trend (except when ns  all). The Kriged surface is slightly worse than the constant sample mean. For example, when ns  9 and  / D  0.05 , then  r /   5.6 , 10.0, and 6.3 for the constant sample mean, bilinear trend, and Kriging surface, respectively. It should be noted that the best performer, the constant sample mean, may be so only because the simulated field is assumed stationary (i.e. constant mean). At the other end of the plot, where   D , the correlation length is underestimated (  r /   1 ). In general, this is because the removal of a trend in a strongly correlated field is also removing the evidence of the strong correlation (strong correlation is evidenced by a trend having little variation off the trend) resulting in a residual field without strong correlation – hence a small correlation length. Of the three trend types considered the best performer at the large correlation length end is again the constant sample mean. For example, when ns  9 and  / D  10 , then  r /   0.08 , 0.05, and 0.06 for the constant sample mean, bilinear trend, and Kriging surface, respectively. 3

CONCLUSIONS

There is no difference between the accuracies of the trend type selected when matching the trend to the field average,  r . As expected, the accuracy improves as the number of samples and the correlation length increase. If a target standard deviation,  r , equal to 20% of the random field standard deviation,  cell , is desired, then only one sample is required if  / D  10 , while 9 or more samples are required if  / D  1 . In general, if the correlation length is small, the most accurate approach is to use a constant sample mean, which shows the best general results for all three measures of accuracy considered in this paper. Kriging is almost identical, only losing out slightly when considering the residual estimated correlation length. At the other end of the scale, when the correlation length is large, the bilinear trend is more accurate with respect to the residual standard deviation than is the constant sample mean, as expected. In the absence of knowledge about the actual correlation length, it appears that the Kriging surface removal, although not generally the best in any one measure, is very competitive and is certainly a good overall choice. 4

Figure 3. Estimated correlation length of the residual, normalized by the point correlation length, versus normalized actual correlation length.

The correlation length estimated from the residual,  r , will agree with the actual correlation length used in the simulation,  , when the ratio  r /   1 . It can be immediately seen in Figure 3 that this only occurs in general when the entire field is sampled and the correlation length is relatively small (i.e. significantly less than D ). That is, when the entire field is sampled ( ns  all), so that the sample average is equal to the actual field average, the estimated correlation length becomes equal to the actual correlation length when the samples are relatively independent (small  ).

REFERENCES

Fenton G.A. and Vanmarcke, E.H. 1990. Simulation of Random Fields via Local Average Subdivision, ASCE Journal of Engineering Mechanics, 116(8), 1733 – 1749. Heisenberg W. 1927. Über den anschaulichen Inhalt der quantentheoretischen Kinematik und Mechanik, Zeitschrift für Physik, 43(3-4), 172 – 198. Lloret-Cabot, M., Hicks, M.A., and Van Den Eijnden, A.P. 2012. Investigation of the reduction in uncertainty due to soil variability when conditioning a random field using Kriging, Géotechnique Letters, 2, 123 – 127.

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Multi-Sleeve Axial-Torsional-Piezo Friction Penetration System for Subsurface Characterization Système de pénétromètre à friction axial-torsional-piezométrique à manchons multiples pour la reconnaissance des sols superficiels Frost J. D., Martinez A.

Georgia Institute of Technology ABSTRACT: The multi-sleeve penetration system is an in-situ testing device that is derived from the cone penetration test. It incorporates a series of friction sleeves with varying surface texture along with a series of pore pressure sensors, in addition to the standard smooth friction sleeve and pore pressure sensor located directly behind the tip in the conventional CPT device. The multiple measurements made with this device allow it to provide new insight into soil type and stratigraphic variations as well as in-situ shear strengths as a function of sleeve texture height. This paper describes a third generation version of this device that incorporates torsional load sensing capabilities in addition to the standard axial load sensing capabilities. In this manner, the effects of different vertical and horizontal stress states on measured sleeve stresses can be explored. This multi-sleeve technology offers benefits over devices which are used to measure the mechanical response of soils. RÉSUMÉ : Le système multi-manchon de pénétration est un dispositif de test in situ qui est dérivée à partir de l'essai de pénétration de cône. Il comporte une série de manchons de friction avec plus ou moins de surface le long d'une série de capteurs de pression de pore, en plus de la douille de friction lisse et standard de capteur de pression de pore situé directement derrière l'extrémité dans le dispositif de CPT classique. Les multiples mesures effectuées avec cet appareil permettent d'apporter un nouvel éclairage sur le type de sol et les variations stratigraphiques ainsi que in situ la résistance au cisaillement en fonction de la hauteur de la texture manche. Cet article décrit une version de troisième génération de ce dispositif qui intègre la charge de torsion capacités de détection, en plus de la charge axiale norme capacités de détection. De cette manière, les effets des différents états de contraintes verticales et horizontales sur les contraintes manches mesurées peuvent être explorées. Cette technologie multi-douille offre des avantages par rapport à d'autres appareils qui sont utilisés pour mesurer la réponse mécanique des sols. 1

INTRODUCTION

The general trend followed for in-situ site characterization practice has been to utilize devices that incorporate only one sensor of a given type to measure desired engineering properties. While a number of different sensor types may be incorporated into a single device, they typically measure different properties and then rely on empirical correlations to predict engineering properties. The primary reason for this single sensor approach has been historical precedent as opposed to any compelling technical limitations. While this approach has proven to yield generally acceptable results for many projects, opportunities remain to improve practice. For example, as the complexity and uniqueness of investigation projects increase, the merit of conventional single sensor insitu tools decreases. Hence, recent efforts have sought to develop new tools for subsurface characterization studies configured with multiple sensors, which have the ability of providing more reliable information as part of more detailed investigations. As noted above, invasive site characterization tools have traditionally followed the approach of using “single-sensor” configurations. An example is the cone penetration test (CPT). The CPT measures, as a minimum, the penetration resistance of a conical tip inserted into the ground, the frictional force that the soil exerts on a smooth sleeve located just above the cone tip, and the pore pressure (assuming the pores are fluid filled) recorded at a location

also typically close to the penetrating tip as the probe is inserted into the subsurface. Such an in-situ tool can provide a robust set of data in the sense that it measures the bearing and frictional resistances of the soil being tested. However, one shortcoming is that it only measures the frictional response of the soil when sheared against a surface of fixed and specified low roughness. Studies by Frost and DeJong (2005) have shown that friction measurements of soil against smooth surfaces are more indicative of soil particle sliding along the surface and not of shearing against the sleeve surface. A more robust characterization of interface strength can be achieved when the soil is sheared against a range of surfaces of different roughnesses (DeJong et al., 2001). 2

MULTI-SENSOR IN-SITU TOOLS

Among the new generations of more specialized in-situ tools that exploit the multiple sensor approach are the “multi sleeve penetrometer attachments” developed at the Georgia Institute of Technology (DeJong, 2001; DeJong and Frost, 2002; Hebeler, 2005; Hebeler and Frost, 2006; Frost et al., 2012). These attachments are designed to be used behind a regular 15cm2 CPT, or as a stand-alone device behind an instrumented tip. The first and second generation devices were described in detail by DeJong and Frost (2002) and Hebeler and Frost (2006), respectively, and are briefly summarized below. The third generation device is under development and is introduced herein.

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2.1 First Generation: Multi-sleeve Friction Attachment (MFA). The first generation of multi sensor attachments deviates from the standard CPT in that the MFA is capable of measuring four different friction sleeve stresses in addition to the standard CPT measurements (qt, fs, u2). Each sleeve position offers the possibility of being equipped with a sleeve of different roughness, with the intention of inducing different degrees of shearing in the soil. Figure 1 shows a schematic of the MFA. According to studies conducted by Frost and DeJong (2005), the standard smooth CPT friction sleeve measurement is more indicative of soil sliding against the sleeve as opposed to shearing against the soil. The reason is that the conventional CPT friction sleeves are manufactured with an intentionally smooth surface. As a consequence of the MFA’s multi-sensor configuration, the device is able to determine the end bearing capacity of the soil and the relationship between interface shear strength resistance and surface texture in a single sounding. The important relationship between interface shear strength and surface roughness was originally identified through laboratory tests by Uesugi and Kishida (1986).

3

PROPOSED SITE CHARACTERIZATION TOOL

3.1. Third Generation: Multi-sleeve Piezo-Friction-Torsion Attachment (MPFTA). The third generation of multi-sensor devices being developed at the Georgia Institute of Technology incorporates both axial and torsional shear as well as pore pressure sensing capabilities. Attachment Digital Housing Attachment Digital Boards

Attachment Digital Housing Attachment Digital Board

(1.66)

(1.40)

fa4

fs#4

Friction Sleeve

(1.14)

fa3

(0.88)

Mandrel fs#3

fa2

Friction Sleeve

(0.67)

fa1 ua0

Mandrel fs#2

Friction Sleeve

Digital Housing Digital Board

fs

Friction Sleeve

u2

Pore Pressure

qc

Tip

Piezo Sensor

Piezo Sensor

Attachment Sleeve Mandrel

Piezo Sensor

Attachment Individual Piezo Sensor

Piezo Sensor

Attachment Individual Load Cell

Piezo Sensor

Digital Board

Friction Sleeve

Dual Axis Inclinometer

Replaceable Attachment Friction Sleeve

Dual Axis Inclinometer

61 cm

Conventional 15 cm2 CPT Module


Digital Housing

Mandrel fs#1


ua1

(0.81)


ua2

(1.07)


ua3

(1.33)

Mandrel

Mandrel

ua4

(1.59)

109 cm

Multi-Friction Sleeve Penetrometer Attachment

2.2 Second Generation: Multi-sleeve Piezo-Friction Attachment (MPFA). The second generation of multiple sensor devices offers the ability to directly measure the interface response over a range of counterface profiles, while simultaneously measuring the excess pore water pressure ahead of and after each friction sleeve as the device is advanced into the subsurface. This is achieved by means of its four independent load cells attached to the textured sleeves and five independent dynamic pore pressure sensors.

The coupling of axial load and pore pressure sensors gives the MPFA the ability to provide a direct measure of pore water pressure generation due to shearing against surfaces of different roughnesses. Several advantages offered by the MPFA are the ability to consider the measured interface response data within an effective stress framework which is useful for applications such as liquefaction as well as strength degradation, flow and consolidation characteristics along the penetrometer’s shaft, more detailed data for improved stratigraphy profiling, and the ability to distinguish between drained, undrained and partially drained conditions at the various sensor locations (Hebeler, 2005). Figure 2 shows a schematic of the MPFA. Examples of the unique insights resulting from the multi-sleeve sensor technology include in-situ determination of the relationship between interface friction and sleeve surface roughness (Figure 3) and soil classification using interface behavior (Figure 4).

4.37 cm

Figure 1. Schematic of the multi-sleeve friction penetrometer along with a standard CPT module.

(a)

fs

Friction Sleeve

u2

Pore Pressure

qc

Tip Load

(b)

Figure 2. Schematic of the multi-piezo-sleeve friction penetrometer along with a standard CPT module (a) schematic - brackets indicate sensor offset from tip in meters and (b) piezo friction sleeve mandrel design detail.

Its dimensions and external characteristics, with and without pore pressure sensing capabilities, are similar to the MFA and MPFA shown in Figures 1 and 2, respectively. However, the new concept incorporated into the device consists of a dual load-torsion cell being installed in each sleeve module

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and location, with the goal of measuring both axial and torsional shear responses of the soil throughout the same sounding. In this manner, the effects of special variability (vertical and horizontal) will be eliminated and more detailed information about the soil’s anisotropy and state of stress can be provided. The proposed texture of the MPFTA’s friction sleeves is the same to the texture of the MFA and MPFA’s sleeves as shown in Figure 5. The friction sleeve texture pattern consists of machined diamond shaped features with a height that typically ranges from 0.25 to 2 mm in order to induce different degrees of shearing. The configuration of penetration angle, diamond width, diagonal spacing, texture slope and areas with no textural features ensures that shearing is induced with the soil and prevents clogging of the textural features.

situ shear vane test. The shear vane is typically used to characterize the response of soft clays. The reason is that stiffer soils can compromise the structural integrity of the shear vane, resulting in blade bending. It is considered that this is not a limiting factor for the MPFTA’s frictional elements because of their different design and thus stiffer configuration. It is important to note that the MPFTA’s intent is to the surface interface strength of the soil in the axial and radial directions, while the shear vane’s intent is to measure the soil’s undrained shear strength. Finally, as shown by Chandler (1988), different diameter sizes can impose strain-rate effects; however since the diameter of the MPFTA device is constant and only the height of the diamond texture elements changes, the results of the MPFTA will not need to be corrected for this and other potential geometry effects.

Interface Sleeve Stress (kPa)

400 Values from Individual Sleeve Tests (Soundings S17-S22) S31 - CPT - Smooth - 30H.125S3 - 30H.25S3 - 30H.5S3 S32 - CPT - Smooth - 30H.5S3 - Smooth - 30H1S3 S33 - CPT - Smooth - 30H1S3 - Smooth - 30H2S3

300

200

100

0 0

2

4

6

8

Figure 5. MFA and MFPA’s friction sleeves with increasing diamond height (from left to right)

10

Normalized Roughness, Rn* (=Rmax (diamond height)/D50 (sand))

Figure 3. Relationship between surface roughness and interface friction determined using multi-sleeve technology.

Figure 4. Soil Classification Chart based on multi-sleeve data.

3.2 Sleeve Locking Mechanism. For the MFA and MPFA devices, the axial force is derived from measurements using a series of bonded strain gauges configured as the fourbranches of a wheatstone bridge. Application of the soil shear force on the textured sleeves brings them into contact with a “shoulder” and the resulting change in length of the bonded strain gauges changes the output of the Wheatstone bridge. In order to measure the torque applied when the sleeve is rotated, the sleeve is temporarily fixed to the core of the mandrel by an electromagnet which prevents rotation of the sleeve and instead induces changes in resistance of a set of orthogonally bonded strain gauges also configured as the branches of a Wheatstone bridge. Given the magnitude of the forces on even the most heavily textured sleeves, relatively low currents are required to “lock” the electromagnets and thus sleeves during torsional testing. A sketch and photograph showing the axial and torsional load application modes for the new device are shown in Figure 6. Final designs of the actual combined axial-torsional cell are being completed. Once measurements at a given sounding elevation are completed, the electromagnets are turned off and the penetration of the device and recording of axial loads is continued. In many instances, the device will be advanced so that a sleeve is advanced to the same elevation that the adjacent preceding sleeve was located at in a previous torsional test so that successive torsional test measurements are made at the same elevations with sleeves of increasing texture height. This eliminates the need to account for lateral and vertical variability since successive tests are performed on the same material.

3.1 Comparison to Existing In-Situ Testing Systems. The MPFTA device has relatively little in common with the in-

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4

CONCLUSIONS

Significant advances have been made in the last four decades in the design, use and interpretation of the results from penetrometer devices used for subsurface characterization. Similarly, over the past two decades, significant new insight has been developed into the role of surface roughness on the behavior of geotechnical interfaces. An emerging family of innovative devices has been developed in the last decade that leverages the advances in performance of penetrometer devices with the new understanding of interface behavior to produce multi-sleeve devices that allow for direct in-situ determination of the relationship between surface roughness and interface shear as well as the development of shear induced pore pressures when surfaces of various roughness are sheared against soils. A unique aspect of this family of devices is the use of multiple friction sleeves with surface of different roughness in the same sounding so that the effects of material variability can be isolated and/or eliminated. To date, all these devices rely on response of soils during axial penetration.

there are clear advantages to developing multiple sensor systems for future investigation studies. The recent development of various configurations of multisensor systems including the previously presented generation 1 MFA and generation 2 MPFA as well as the proposed generation 3 MPFTA device introduced herein represent a significant departure from traditional practice. Amongst the benefits of the latter device are: i) Up to sixteen independent measures of interface shear ranging from smooth surface sliding to textured surface soil shearing can be realized in a single sounding in contrast to the three measures possible with conventional cone penetration systems. ii) The effects of different vertical and horizontal stress states on measured sleeve stresses can be accounted for by means of the axial and torsional interface stress response. iii) The procedure for performing combined axial-torsionalpiezo penetration testing of the subsurface using the proposed MPFTA system involves a series of steps which allows them to be readily controlled from a remote location and to be performed using robotic systems. 5

ACKNOWLEDGEMENTS

The initial device development reported herein was funded in part by a grant from the US National Science Foundation to the Georgia Tech Research Corporation under Contract # CMS 9978630. 6

Figure 6. Schematic and photograph of multi-sleeve piezo- frictiontorque penetrometer showing load application modes.

This paper describes the development of a new device that embodies the attributes of the existing multi-sleeve devices but incorporates the ability to also conduct torsional friction penetrometer tests in the same sounding. In contrast to existing vane shear type devices which involve the application of a torsional force to a rigid central shaft and measure the resistance to rotation of a set of blades in a soil, the new device enables measurement of torsional resistance with the same textured sleeves used in the axial stage of the test. This is possible through the use of an innovative electro-mechanical system that allows independent measurements of axial and torsional resistance of the sleeves of the penetrometer device. The availability of complimentary axial and torsional shear forces along with the associated pore pressures generated by friction sleeves of different surface roughness represents a potential “disruptive technology” in the in-situ characterization of soil properties ranging from soil type to soil strength and deformation properties to assessment of the in-situ state of stress and associated parameters such as the in-situ stress ratio. Significant opportunities exist for dramatic advances in subsurface investigation. Single sensor historical precedent has guided the design and configuration of in-situ devices,

REFERENCES

Chandler, R.J. (1998). “The in-situ measurements of the undrained shear strength using the field vane”, Vane Shear Strength Testing in Soils. Field and Laboratory Studies. A.F. Richards (ed.), ASTM STP 1014, ASTM, Philadelphia, pp. 13-44. DeJong, J.T. (2001). “Investigation of Particulate-Continuum Interface Mechanics and Their Assessment Through a MultiFriction Sleeve Penetrometer Attachment”, PhD Dissertation, Georgia Institute of Technology, Atlanta, May, 360 pp. DeJong, J.T. and Frost, J.D. (2002). “A Multi-Friction Sleeve Attachment for the Cone Penetrometer,” ASTM Geotechnical Testing Journal, 25, No. 2, pp. 111-127. DeJong, J.T., Frost, J.D., and Cargill, P.E. (2001). “Effect of Surface Texturing on CPT Friction Sleeve Measurements.” Journal of Geotechnical and Geoenvironmental Engineering, 127, No. 2, pp. 158-168. Frost, J.D., and DeJong, J.T. (2005) “In Situ Assessment of the Role of Surface Roughness on Interface Response,” Journal of Geotechnical and Geoenvironmental Engineering, 131, No. 4, pp. 498-511. Frost, J.D., Hebeler, G.L., and Martinez, A., (2012), “Cyclic Multipiezo-friction Sleeve Penetrometer Testing for Liquefaction Assessment”, Proceedings of 4th International Conference (ISC’4) on Geotechnical and Geophysical Site Characterization, Pernambuco, Brazil, Vol. 1, pp. 629-636. Hebeler, G.L. (2005). “Multi Scale Investigations of Interface Behavior.” PhD Dissertation. Georgia Institute of Technology, Atlanta, August, 772 pp. Hebeler, G.L., and Frost, J.D., (2006), “A Multi Piezo-Friction Attachment for Penetration Testing”, Proceedings of ASCE Geo-Institute Congress: Geotechnical Engineering in the Information Technology Age, Atlanta, CD ROM. Uesugi, M. and Kishida, H., (1986) “Frictional Resistance at Yield Between Dry Sand and Mild Steel.” Soils and Foundations, Vol. 26, No. 4, pp. 139-149.

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Hydraulic Properties of Glacial Deposits Based on Large Scale Site Investigation Les propriétés hydrauliques des dépôts glaciaires basées sur une enquête de chantier à grande échelle Galaa A., Manzari M.

Coffey Geotechnics, Toronto, Ontario, Canada

Hamilton B.

CH2M Hill, Toronto, Ontario, Canada

ABSTRACT: Glacial deposits by nature comprise variable soil types in relatively short distances. Hydraulic conductivity (K) is the most important parameter in design of construction dewatering for underground structures. However, determination of proper design values for K is not an easy task. Due to the inherent variable nature of the glacial deposits, even conventional pumping tests may not provide reliable design parameter due to its smaller zone of influence compared to that of the actual dewatering for a structure. This paper describes the methodology created for establishing more representative design values for hydraulic conductivity of glacial deposits during a large scale subsurface investigation for planned tunnels. The subsurface investigation involved 400 boreholes, including 88 slug tests and 16 pumping tests. A relation was established between K obtained from the field tests (Kfield) and K calculated by applying Kozeny-Carman formula (KKC). Subsequently, the calibrated K-C formula was applied to 1,200 grain size analyses conducted on various soil types. The calculated and measured K were used to form statistical analysis of the parameter and provide more reliable design values for dewatering. RÉSUMÉ : Les dépôts glaciaires comprennent des sols variables à travers des distances relativement courtes. La conductivité hydraulique (K) est le paramètre le plus important qui est nécessaire durant la construction des structures souterraines. Cependant, la détermination des valeurs de calcul appropriées pour K n'est pas une tâche facile. à cause de la nature variable des dépôts glaciaires, même les essais de pompage peut-être ne fourniront pas des résultats fiables pour une bonne conception pour une bonne conception parce que les structures déshydratés ont une plus grande zone d'influence. Ce document décrit la méthodologie créée pour établir les paramètres de conception plus représentatives au cours d'une enquête de chantier à grande échelle pour les tunnels de métro prévues. L'étude a porté sur 16 essais de pompage avec des puits d'observation associés, et 88 essais de conductivité hydraulique. Une relation a été établie entre K obtenue à partir des essais sur le terrain (Kfield) et K calculé en appliquant la formule de Kozeny-Carman (KKC). Par la suite, la formule de K-C calibrée a été appliquée à des analyses granulométriques effectuée 1200 échantillons. Les valeurs de K calculées et mesurées ont été utilisées pour former une analyse statistique, et pour fournir des valeurs plus fiable. KEYWORDS: Kozeny-Carman formula, hydraulic conductivity, Glacial Tills, dewatering. 1

INTRODUCTION

The Greater Toronto and Hamilton Area (GTHA), located in southern Ontario, is Canada’s largest and fastest growing urban region. The Government of Ontario Province through its transportation authority known as Metrolinx, has embarked in a massive transportation plan called “The Big Move”, which is a 25-year, $50 billion plan that will transform regional transportation across the GTHA. The Eglinton Scarborough Crosstown (ESC) Light Rail Project is part of that Big Move program. The ESC is a 19-kilometre light rail transit line (LRT) that will run along Eglinton Avenue, connecting west to east of the city. Eleven kilometers of the alignment will be tunneled underground, crossing well established urban areas which are densely populated and congested. The tunnel construction is divided in two contract packages: West Twin Tunnels Construction and East Twin Tunnels Construction, with Yonge Street the dividing limit. Dewatering operations will be required for a total of twenty four structures along the tunnel alignment: sixteen cross passages, four launch and exit shafts, and six emergency exit buildings. In order to meet a very tight schedule while properly managing subsurface risk and support the design of the tunnel, an aggressive multi-phase geotechnical investigation program was undertaken. The geotechnical investigation for the west and east tunnel contracts was conducted during a two-stage program between 2010 to mid-2012; which followed by a hydrogeological study for each section. In summary, about four hundred (400) shallow and deep sampled boreholes were advanced including three hundred (300) monitoring wells along

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the subject alignment to obtain information regarding the subsurface stratigraphy and groundwater conditions. Furthermore, eighty eight (88) slug tests and sixteen (16) pumping tests (150 mm O.D.) were completed as part of the site specific hydrogeological study. At the time of preparation of this paper, only the results of eight (8) pumping tests for the west tunnels are available and used in analyses. Due to project’s very tight schedule and ongoing progress of design, the proposed locations of some structures were revised after completion of the pumping tests. Furthermore, it was not practical to conduct the pumping tests for all of the structures. Innovative techniques were developed and used to establish more representative design value of hydraulic conductivity while not having pumping test at exact location of each structure and also consider the inherent variable nature of the glacial deposits. This paper describes the methodology developed and summarizes the range of hydraulic conductivity for various types of glacial deposits obtained from this large scale subsurface investigation which is generally more refined than older published range for the same deposits. 2

GEOLOGY SETTING

A detailed regional description of the Quaternary geology of the project area can be found in the Ontario Geological Survey Map (Sharpe, 1980). The soil deposits in the project area are result of glacial depositional systems that took place during various glacial periods. From the published geological data, the GTHA experienced three glacial and two interglacial periods. This


fluctuating glacial advance and retreat produced a complex distribution of over-consolidated glacial till layers, separated by interstadial and interglacial stratified deposits of glaciolacustrine plastic silt/clays and non-plastic silt/sands. The subsurface overburden encountered during the site investigation were initially classified into 17 different soil types (Types 1 through 17). The soil classification system followed the modified version of Unified Soil Classification System. Identification of soil origin as “till” was based on their heterogeneous structure, the relatively broad grain size distributions and the documented local geology. Many of the different soil types demonstrate relatively comparable engineering characteristics and may possibly have similar geological origin. Consequently, the various soil types were consolidated into six engineering classes (Classes A through F). The six soil classes are as follows:      

Class A: Fill and Topsoil Class B: Interstadial Sand to Gravel Class C: Interstadial Silt to Sand Class D : Non-Plastic Till Class E : Plastic Glacio-lacustrine Class F : Plastic Till

Class B was divided into two subclasses based on the percentage of silt and clay particles (<75 μm). Sandy soils with less than 20% silt and clay particles were grouped under Class B2,3,4 and the rest (> 20% silty and clay) under Class B5,6. 3

ESTABLISHING HYDRAULIC CONDUCTIVITY

Glacial deposits by nature comprise of variable soils types in relatively short distances. Due to the inherent variable nature of the glacial deposits at project area, conventional filed pumping tests may not provide fully reliable results for a proper dewatering calculation as the zone of influence of a pump test may only extend a few tens of meters. On the other hand, the actual dewatering volume of a structure is affected by the characteristics of surrounding soil within a few hundreds of meters. Furthermore, the pumping tests were not necessarily at the exact location of some structures. It became necessary to complement the hydraulic conductivity values obtained through field testing in order to expand the test results to a larger domain or be able to focus on any specific area. It was decided to use the available semiempirical methods/formulae in literature to complement hydraulic conductivity values obtained through filed testing with predicted values based on index properties such as grain size distributions, pore size distributions and/or specific surface. The following sections will outline the procedure followed to predict hydraulic conductivities and provide design parameters. 3.1

Kozeny-Carman formula

Since Kozeny (1927) introduced his theory for a series of capillary tubes and Carman (1938 and 1956) followed this work and provided formulations that takes into the account the tortuosity of the flow path of a fluid in a porous medium. The following formula presented by Carman was then referred to as the Kozeny-Carman (K-C) formula (Carrier, 2003). Details of the formula can be found in the subject references. In summary, the hydraulic conductivity of the soil can be estimated as follows: 2  K 1.99  10 4 100 % / { f i /( Dli0.5  Dsi0.5 )} (1 / SF 2 )[e 3 /(1  e )] (1) Where, e is the void ratio; SF is a shape factor; fi is the fraction of particles between two sieves (%), denoting the larger sieve with (l) and the smaller one as (s) in, and Dave-i = (Dli×Dsi)0.5 is the average particle size, in cm, between two sieve sizes..

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The Kozeny-Carman formula takes into account specific surface area of full range of particle sizes and soil void ratio which leads to better accuracy than the famous Hazen formula (Lambe and Whitman 1969) in predicting the hydraulic conductivity for a wide range of soils. Notwithstanding the above, the application of K-C formula is constrained by almost the same limitations as Hazen (Carrier 2003). Such constrains, as discussed below, arise when dealing with soils at the extremes of any spectrum such as the grain size, particle size distribution, particle shape, and particles orientation (anisotropy). The formula does not account for the electrochemical forces between particles and particles and water which disqualify the formula from being applied to clayey soils. In addition, the formula assumes laminar flow, which may not be satisfied in gravels and gravelly sands. The formula does not produce a close estimate to the specific surface area of particles with extreme shapes such as platy or flakey particles. Therefore, the K-C formula may not be applicable in these cases or can be applied after replacing the calculated specific surface area by the measured value. Also, K-C formula does not account for soil anisotropy which is more pronounced in natural deposits than for laboratory constructed samples. Locat et al (1984) measured the specific surface area (S) for several clays and found that clays with low plasticity (8 < PI < 15) have S between 23 and 30 m2/kg and is independent of the percentage of soil finer than 2 m. Chapuis and Aubertin (2003) picked a constant number between 23 and 30 m2/kg as an estimate for S of the soil fraction finer than 2 m and calculated S for the fraction coarser than 2 m as per original K-C formula. Consequently, the results of these hybrid methods in using K-C formula were in good agreement with measured hydraulic conductivities in laboratory for clayey soils with PI<15. In this study, the approach proposed by Chapuis and Aubertin (2003) was followed for plastic glacial tills with PI less than 15. However, the effect of weathering and factures in the upper portion of the clayey till deposits must be considered in any assessment (McKay, 1993; Hendry, 1982). 3.2

Site specific correlation factor for K-C formula

This section outlines the work completed in the field to obtain in-situ hydraulic conductive (K) for the different soil classes and explains the approach followed to establish site specific correlation factor for using K-C formula. Hydraulic conductivities for each soil class were measured in the field by a combination of pumping tests and/or falling or rising head slug tests. The results of 8 pumping tests with associated observation wells and 88 slug tests conducted along the tunnel alignment, distributed among six soil classes are used in this study. The number of the field tests performed on the aquifers’ materials was greater than those performed on the other soil types. However, a significant number of the tests were performed on both plastic and non-plastic tills. One grain size distribution analysis was conducted, as minimum, on the soil samples recovered from within the screen interval of the 88 slug tests and pumping tests with associated observation wells. These grain size distributions were determined by undertaking sieve analysis, in accordance to ASTM C136-06, and the hydrometer test, in accordance to ASTM D422-63. These grain size distributions analyses were used to calculate K based on the K-C formula. After excluding the tests for samples with PI >15 and/or field test conducted in the clayey till deposits with obvious signs of weathering and fracture, K-C formula was applied to about 80 grain size analyses that were screened as suitable (not within the limitations of the formula) and correspond with K obtained from field tests. As a result, for every in-situ measured K in the field (Kfield) there is a corresponding predicted K from applying KC formula to the grain size analysis associated with the screen interval (KKC), as shown in Figure 1.


1.E+00 Correlation line

K (Field) (cm/sec)

1.E‐01 1.E‐02

K(field) = 3 K(KC) K(field) = 1/3 K(KC)

1.E‐03 1.E‐04 1.E‐05 1.E‐06 1.E‐07 1.E‐07

1.E‐06

1.E‐05

1.E‐04

1.E‐03

1.E‐02

1.E‐01

1.E+00

K (Kozeny‐Carman) (cm/sec)

Figure 1 In-situ measured field hydraulic conductivity versus calculated by Kozeny-Carmen Formula (KKC vs. Kfield)

The dashed line represents the equality line and the solid black line represents the site specific correlation line which has a slope shown in equation (2). log Kfield = 1.03×log KKC

(2)

The grey lines in Figure 1 represent the boundaries that encompass 90% of the data points. These lines have the same slope as the correlation line with ±0.5 offset in the log-log scale. This indicates that Kozeny-Carmen formula with incorporation of the site specific correlation factor of 1.03 (equation (2)) predicts a K value ranging between 1/3 to 3 times the in-situ measured field hydraulic conductivity (Kfield) for the glacial deposits in this specific site. These conclusions are comparable to the margin obtained from laboratory permeability test results shown by Chapuis (2002) and Chapuis and Aubertin (2003). 3.3

Overall hydraulic conductivity for each soil class

Hydraulic conductivity (K) values for each soil class of glacial deposits were calculated using the K-C formula as per method described in the previous sections for about 1,200 grain size analyses conducted on various soil types along the alignment. Equation (2) is then used to correct KKC assuming that 90% of the predicted values fall between 1/3 to 3 times the actual K in the field. The statistical parameters were calculated for the corrected KKC obtained for each soil class in conjunction with the K values directly obtained from field tests (slug and pumping tests). The statistical distribution of K for each soil class is plotted in histograms as shown in Figure 2a to 2e. The K values obtained from the field tests conducted in the plastic till deposits (Class F) with obvious signs of weathering and fracture has also been added to the calculated K values and other field measurement results which all together included in the statistical distribution of K for Class F (Figure 2a). Generally, the higher end of the K distribution in Figure 2a is associated with the field measured hydraulic conductivity in the fractured plastic till. This is in conformance with the finding of other studies in similar soil condition (e.g., D’Astous 1989, Ruland 1991). Although, some of the slug tests conducted on this fractured zone were as low as the results typically associated with soil matrix values; which could by the results of the smeared zone tend to form around augered boreholes.

Figure 2a to 2e Statistical distribution of hydraulic conductivity for various soil classes of glacial deposits obtained from the investigation.

The K values for Class B2,3,4 (interstadial sand with less than 20% fines) fit a bimodal distribution (Figure 2e). Further review of the resutls indicated that the higher peak (10-2 cm/s) is associated to sand with leass than about 10% fine; while the rest of the class resutled to the lower peak. 3.4

Design hydraulic conductivity for structures

The zone of influence for 72 hours pumping tests ranged from 15 m to less than 100 m, depending on the location. On the

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other hand, the zone of influence for actual dewatering volume of the structures would be a few hundreds of meters and therefore, the dewatering volume would be affected by the characteristics of surrounding soil within this larger zone. In order to assess the reliability of the pumping test results for dewatering calculation, the uniformity of the soil within the dewatering zone was verified using the correlation described in the previous sections. For each structure location, a zone of influence of 350 m radius is assumed. Corrected KKC in conjunction with K values directly obtained from field tests (slug and pumping tests) within the assumed zones around each structure were pulled out of the overall data available. Subsequently, the statistical distributions of K-values for every soil class encountered within the dewatering zone were prepared for each structure. Examples of the cumulative distributions are shown in Figure 3a and 3b for Structure No.1 and No.2. Based on the localized distribution of the K-values for each structure, the pumping tests results for some structures fall within 70 percentile or higher; on the other hand, the results for other structures could be as low as 20 to 50 percentile. A detailed review of the results and interpretive subsurface profile showed that generally when the zone of the influence of the pumping tests was small, the K obtained from pumping test tends to be on the lower side of the cumulative distribution. This has also been augmented where random presence of pockets/seams of Class C soil within Class B deposits has dominant effect on pumping test results. The design K-value for dewatering calculation has been selected based on the result of the localized distribution of the K-values prepared for each structure. Two examples are shown in Figure 3. 4

CONCLUSION

piezometers. Physical scale of field measurements may strongly influence the resulting hydraulic conductivity values.

Figure 3 Localized distribution of hydraulic conductivity for (a) Class B5,6 in Structure No.1 and Class B2,3,4 Structure No.2.

5

The authors would like express their gratitude to Metrolinx for authorizing the preparation of this paper. 6

Glacial deposits comprise of variable soil types in relatively short distances. Conventional pumping tests may not provide fully reliable results for a proper dewatering calculation as the zone of influence of a pump test may only extend tens of meters while the actual dewatering volume of a structure is affected by the characteristics of surrounding soil within hundreds of meters. Presence of pockets/seams with higher silt content within sandy deposits has dominant effect on pumping test results. Smaller the zone of influence of the pumping tests, K obtained from the test tends to be on the lower side of the cumulative distribution for the dewatering zone of influence. The pumping test results for some structures could be as low as 20 to 50 percentile of accumulative distribution. It is imperative to assess the reliability of the pumping test results for dewatering calculation in the variable glacial deposits; particularly when the zone of the influence of the pumping tests is relatively small. The Kozeny-Carman formula takes into account specific surface area of full range of particle sizes and soil void ratio and proven to provide reliable predictions of K for wide range of soils. Based on the results of this large scale investigation, Kozeny-Carmen formula with incorporation of the site specific correlation factor, predicts K values ranging between 1/3 to 3 times the in-situ hydraulic conductivity (Kfield) for the glacial deposits. This provides a powerful tool in verifying the reliability of pumping test results in glacial deposits. However, careful consideration must be given to proper interpretation of the field test results and applicability of the formula to site conditions. It also should be noted that K of weathered zone of clayey deposits is controlled by flow through the fractures. The field K measured in this zone could be up to a few orders of magnitude greater than the clay matrix. Field measurements in this zone may also be sensitive to smearing during the installation of

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ACKNOWLEDGEMENTS

REFERENCES

Carman, P. C. 1938. The determination of the specific surface of powders. J. Soc. Chem. Ind. Trans. 57, 225. Carman, P. C. 1956. Flow of gases through porous media, Butterworths Scientific Publications. London. Carrier, W. D. 2003. Goodbye, Hazen; Hello, Kozeny-Carman. Journal of Geotechnical and Geoenvironmental Engineering, 129(11), 1054-1056 Chapuis, R.P. and Aubertin, M. 2003. On the use of the Kozeny– Carman equation to predict the hydraulic conductivity of a soil. Canadian Geotechnical Journal, 40(3), 616-628. Chapuis R.P. 2002. The 2000 R.M. Hardy Lecture: Full-scale hydraulic performance of soil–bentonite and compacted clay liners. Canadian Geotechnical Journal, 39(2), 417-439. D’Astous et al. 1989. Fracture effects in the shallow groundwater zone in weather Sarnia-area clay. Canadian Geotechnical Journal, 26, 43-56. Hendry, M. J. 1982. Hydraulic conductivity of a glacial till in Alberta. Ground Water 20(2), 162-169. Kozeny, J. 1927. Ueber kapillare Leitung des Wassers im Boden. Wien, Akad. Wiss. 136 (2a), 271. Lambe, T. W., and Whitman, R. V. (1969). Soil mechanics. Wiley, New York. Locat, J., Lefebvre, G., Ballivy, G. 1984. Mineralogy, chemistry, and physical properties interrelationships of some sensitive clays from Eastern Canada. Canadian Geotechnical Journal, 21 (3), 530-540. McKay L., Cherry J., and Gillham R 1993. Field Experiments in a Fractured Clay Till. Water Resources Research, 29(4), 1149-1162. Ruland W. et al. 1991. The depth of active groundwater flow in a clayey till plain in southwestern Ontario. Ground Water 29(3), 405-417 Sharpe 1980. Quaternary Geology Series, Quaternary Geology – Toronto and Surrounding Area, Southern Ontario. Ontario Geological Survey Map 2204.

The seismic SPT test in a tropical soil and the G0/N ratio L'essai SPT sismique pour le sol tropicaux et la relation G0/N Giacheti H.L., Pedrini R.A.A.

Universidade Estadual Paulista, Departamento de Engenharia Civil e Ambiental, Bauru – SP – Brazil Rocha B.P. B. P. Rocha

Universidade de São Paulo, Departamento de Geotecnia, São Carlos – SP – Brazil

ABSTRACT: The seismic SPT, a test which associates the up-hole technique to the SPT, is briefly described. The maximum shear modulus (G0) can be determined together with the N value with this hybrid test. Seismic (Cross-hole, Down-hole and SCPT) and SPT test data for a Brazilian tropical sandy soil are presented and discussed emphasizing the advantage of using the interrelationship between the small strain stiffness (G0) and an ultimate strength (N value) to identify different soil behavior. A seismic SPT test was carried out in this research site and the G0/N ratio is discussed as an interesting index to help characterize tropical soils, similar to what has been suggested for the Go/qc ratio determined in a single test. RÉSUMÉ : Le SPT sismique, qui associe le up-hole au SPT est brièvement décrit. Le module de cisaillement maximale (G0) peut être déterminé avec la valeur N de ce test hybride. Des données sismiques (Cross-hole, Down-hole and SCPT) et SPT pour un sol sableux tropical du Brésil sont présentées et discutées soulignant l'avantage d'utiliser la corrélation entre (G0) et une résistance à la rupture (valeur N) afin d'identifier le comportement de différents sols. Un essai SPT sismique a été réalisé dans le site expérimental et la relation G0/N est discutée comme un indice intéressant pour aider à caractériser les sols tropicaux, de la même façon que ce qui a été proposé pour le rapport G0/qc mesuré dans un essai unique.

KEYWORDS: In situ testing, SPT, seismic, up-hole, tropical soil, G0/N ratio. 1

INTRODUCTION

Site characterization can be defined as the process of identifying the geometry of relatively homogeneous zones and developing index, strength and stiffness properties for the soils within these zones. Some in situ testing can be used as an alternative to the traditional approach of drilling, sampling and laboratory tests. Combining stratigraphic logging with a specific measurement in a in situ test is a modern approach for site characterization. Some authors have shown that it is possible to incorporate the measurement of shear wave velocities using the SPT blow by the up-hole technique. This hybrid test is known as the seismic SPT (S-SPT), which combines stratigraphic logging, estimative of geotechnical parameters and determining small strain stiffness (Go) in one single test similarly to the SCPT. This paper briefly describes a system to carry out the S-SPT test and the approach to interpret the seismic data. It also discusses the applicability of the interrelationship between (Go) and N value to identify unusual soil behavior based on the tests carried out in a research sites located in the city of Bauru, inland of São Paulo State, Brazil emphasizing the advantage of using the S-SPT test for this approach. 2 2.1

BACKGROUND Tropical Soils

Tropical soils are formed predominantly by chemical alteration of the rock, and they are considered a non-textbook type geomaterial because their peculiar behaviors that cannot be explained by the principles of classical soil mechanics. The term tropical soil includes both lateritic and saprolitic soils. Saprolitic soils are necessarily residual and retain the macro fabric of the parent rock. Lateritic soils can be either residual or transported and are distinguished by the occurrence

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of the laterization process, which is an enriching of a soil with iron and aluminum and their associated oxides, bonding a highly porous structure. Saprolitic soil has structural or chemical bonding retained from the parent rock. The contribution of this cementation to the soil stiffness depends on the strain level the soil will experience. Differences between the mechanical behaviors of the mature (lateritic) and young (saprolitic) soils have been reported for both natural and compacted conditions. 2.2

Go/qc Ratio

The pore pressure measurements cannot always be considered useful to allow an adequate classification of tropical soil based on CPTU data. The small strain stiffness (Go) and cone tip resistance (qc) ratio has been suggested as an additional information for classifying different soil types, especially to identify soils with unusual compressibility. Schnaid et al (2004) suggested that the ratio Go/qc provides a measure of the ratio of the elastic stiffness to ultimate strength and may therefore be expected to increase with sand age and cementation, primarily because the effect of these on Go are stronger than on qc. They proposed a chart and boundaries by correlating Go/qc versus normalized type resistance (qc1). This chart can be used to evaluate the possible effects of stress history, degree of cementation and ageing for a given profile. Three lines divide upper and lower bounds for cemented and uncemented sands. Giacheti & De Mio (2008) presented SCPT test results from three tropical research sites in the State of São Paulo, Brazil and plotted all the data in the Schnaid et al (2004) chart as shown in Figure 1. The authors pointed out that the SCPT test allows calculating Go/qc ratio simplifying interpretation and reducing site variability. The SCPT data interpretation indicated that the bonded structure of tropical soils gives Go/qc ratios that are systematically higher than those measured in cohesionless soils.


The results are in agreement with the propositions of Schnaid et al. (2004). They also observed that lateritic soils tends achieve a higher Go/qc ratio than the saprolitic soils.

Classification System proposed by Nogami & Villibor (1981) for tropical soils was used to define and classify the soils with regards to its lateritic behavior.

Figure 3. Bauru city, where the studied site is located.

SPT and seismic testing data

3.2 Figure 1. Relationship between Go and qc (Giacheti & De Mio, 2008).

2.3

Go/N Ratio

Schnaid et al. (2004) suggested that the N values from SPT test can also be combined with Go, using the Go/N ratio, to help assessing the presence of bonding structure. This approach is presented in Figure 2. Viana da Fonseca & Coutinho (2008) included data from experimental sites from Portugal in this figure. These authors pointed out that the bonded structure have a marked effect on the behavior of residual soils, with a Go/N values considerably higher than those observed in cohesionless materials. Lines are also shown in Figure 3 to define the upper and lower bounds for cemented and uncemented sands. Similarly to what has been presented by Giacheti & De Mio (2008) for tropical soils based on SCPT data, the interrelationship between small strain stiffness (Go) and N value could be used to identify different soil behavior using the seismic SPT similarly to the SCPT.

The typical soil profile for the studied site was defined based on the SPT tests and it is presented in Figure 4.a, together with N values correct by 60% efficiency (N60) for all SPT tests (Figure 4.b). The shear wave velocities (Vs) were determined with cross-hole, down-hole and SCPT tests (Figure 4.c). Total mass densities were obtained from undisturbed soil samples collected in a sample pit excavated at the site. They were used to calculate Go values based on Elastic Theory and the data are presented in Figure 4.d. An average Go/N60 ratio for every one meter depth was calculated, so the Goavr/N60avr values versus depth are presented in Figure 4.e. The criteria to calculate this ratio was averaging Go and N60 from all the tests and after that calculating the average ratio with the closest depth from Go and N60. Site variability can be assessed based on N60 and Vs values and these data indicate that the site is quite variable. Giacheti at al (2003) discussed variability for this site based on several CPT tests. They also concluded that the site is variable and test data can be affected by suction and cementation. The authors pointed out that the SCPT1 shows the presence of a region with low qc and high Rf between around 10 and 16 m depth. These data are quite different from those recorded with the SCPT2 test, so Vs values were not considered to calculate Go for this portion of the soil profile for the SCPT1 test. This variation is probably related to the morphogenetic and pedogenetic processes and probably reflects different degrees of cementation in the profile. N60

SPT profile 0 1

0

(a)

10

20

Go (MPa)

Vs (m/s) 30

40 0

200

400

(b)

600 0

Goavr / N60avr

100 200 300 400 0

(d)

(c)

10 20 30 40 50 60

(e)

2 3 4 5 6 7

1

8

Depth (m)

Figure 2. Relationship between Go and N60 (Schnaid et al 2004, completed by Viana da Fonseca & Coutinho, 2008).

9 10 12 13

3 3.1

STUDIED SITE

14

LA' NA'

15 16

The site

The Unesp experimental research site is located inland of the State of São Paulo, Brazil, in the city of Bauru (Figure 3). Several site characterization campaigns including SPT, DMT, PMT, CPT, SCPT, cross-hole and down-hole tests were previously carried out at the site. A sample pit was excavated to retrieve disturbed and undisturbed soil blocks to be tested in the laboratory to characterize the soils and to determine geotechnical properties. The subsoil is a sandy soil where the top 13 m has lateritic soil behavior (LA’) overlaying a soil of non-lateritic behavior (NG’) derived from weathering of Sandstone rock. The MCT

SM - SC

11

17 18 19 20

2 3

21 1-Red clayey fine sand 2-Red clayey silty fine sand 3-Red clayey fine sand

CH1 SCPT1

CH1 SCPT1

SCPT2

SCPT2

DH

DH

Figure 4. SPT and seismic testing data and Go/N60 for the studied site.

3.3

The Go/N ratio

It can be observed in Figure 4.e that the average Go/N60 tends do decrease with depth, with an average value equal to 35

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between 1 and 6 m depth, 23 between 6 to 13 m depth and 10 below 13 m depth. These results indicate that Go/N60 ratio is higher in the lateritic soil layer (1 to 13 m depth) and tends to decrease as the residual soil is less developed. The average Go and N60 values for the study site were plotted in the Go/N60 versus (N1)60 chart (Figure 5). Almost all the data points are in the upper bound for cemented sands which indicates that the bonded structure of tropical sandy soils produces Go/N60 that are systematically higher than those measured in cohesionless soils. It is also interesting to note in Figures 4.e that the lateritic soils (G0/N60=35 to the upper portion and 23 to the lower portion) present a higher cementation than the saprolitic soils (average G0/N60=10). These results are similar to what had been presented by Giacheti & De Mio (2008) based on Go/qc from SCPT test (Figure 1) and indicate the use of the ratio between the small strain stiffness (Go) and an ultimate strength (N or qc) to identify unusual soil behavior and degree of evolution of residual soils.

Ratio (Go/pa)/N60

1000

Upper bound (cemented geomaterials)

4.2

Vs from the S-SPT test

Determining Vs from the S-SPT test data is not straightforward. Bang & Kim (2007) described two methods: DTR (delay time between serial receivers) and DTS (delay time between serial sources). Pedrini (2012) suggested using the DTS method. In this method, the time interval of the S waves arrival for each sample depth in which the test was carried out is determined identifying the exact moment of the first arrival time plotting the wave receptions generated at different depths. Figure 7 shows a typical wave recordings profile as well as the point of the first S wave arrival. Another important aspect is the geometry. Bang & Kim (2007) recommend that Snell’s Law (the refraction and reflection during the propagation of waves in stratified layers of different densities) should be taken into account when determining the refracted wave path.

Lateritic Soils Saprolitic Soils

100

Unaged uncemented sands Lower bound (cemented geomaterials) 10 1

10

100

Normalized (N1)60

Figure 5. Relationship between Go and N60 for the studied site.

4 4.1

THE SEISMIC SPT TEST Figure 7. Profile of seismic wave and the identification of the common arrival point of the S waves (Pedrini et al, 2012).

Principle

It is possible to incorporate the shear wave velocity (Vs) measurements during the SPT test applying the up-hole technique. This approach has been used in the past and it is recently presented in detail by Bang & Kim (2007). This hybrid test allows measuring the SPT N value together with Vs (so Go) at the same time and in the same borehole. For each sampler depth (usually at every meter) a seismic wave is generated and it can be recorded on the ground surface. A schematic representation of the S-SPT test is show in Figure 6. Manual SPT Equipment

4.3

Trigger & Anvil

DAQ System

Case with geophones

1

H1 H2

2

H3 Hi

3 i

The refracted ray pathway calculated based on Snell’s Law depends on various wave velocities and it can be determined by considering two conditions: the Snell’s law and a geometrical criteria. The following assumptions must be done: 1) each sample layer is equal to the depth where the SPT test was carried; 2) each layer is homogeneous and the propagated wave velocity is assumed constant in each layer as show in Figure 6. An iterative method must be used to solve the equation system and determine the length (L) that the wave propagates in each soil layer. Details can be found in Bang & Kim (2007).

L1

L2

L3

Li

Figure 6. Schematic representation of an S-SPT test and a seismic refracted path (adapted from Bang & Kim, 2007 by Pedrini et al, 2012).

The test equipment is the same currently used for the SPT test. An arrangement of transducers (usually geophones) placed in appropriate boxes on the ground surface, a triggering system and the seismic source, which is the SPT sampler itself, are added for the seismic SPT test.

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The S-SPT equipment

The system for carrying out S-SPT tests and the method of analysis were implemented and described by Pedrini (2012). The main characteristics of this system are presented by Pedrini et al (2012) and will be summarized herein. Bang & Kim (2007) used the drop of the SPT weight as the source to generate waves while Pedrini (2012) used a 2 kg sledgehammer. The triggering device was digital, with one terminal (positive or negative) fitted into the anvil head and the other attached to the sledgehammer. Two geophones were installed inside of six boxes placed on the ground, one vertical and other horizontal oriented in a radial pattern. A National Instruments, model NI-USB-6353, data acquisition system was used. It has 16 bits resolution, 32 single ended channels and 16 differential channels, a digital and analogue trigger and a receiving rate of 1.25 ms/s. Software in the Labview and Matlab platforms were developed by Pedrini (2012) to trigger, capture the waves, signal processing, represent the traces, analyzing the recorded data and calculating the velocities.


The S-SPT test procedure

4.4

1000

N60

0 1

0

10

20

Vs (m/s) 30

(a)

40

(b)

0

200

Go (MPa)

400

600 0

(c)

100 200 300 400

Go / N60

(e) (d)

3 4 5 7

1

Depth (m)

8 9 10

SM - SC

11 12 13 14

LA' NA'

15 16 17 18 19 20

2 3

100

Unaged uncemented sands

1

10

100

Normalized (N1)60

Figure 9. Relationship between Go and N60 for the seismic SPT test.

5

6

21 1-Red clayey fine sand 2-Red clayey silty fine sand 3-Red clayey fine sand

Saprolitic Soils

CONCLUSIONS

It was observed that the average ratio Go/N from several SPT and seismic tests carried out in the studied site was higher in the lateritic soil than in the saprolitic soil, particularly in the top layer. The seismic SPT test was used to derive Go/N values in the same site. Similar results were achieved with this single test, which allows determining both parameters simultaneous, reducing the effects of site variability. Relating an elastic stiffness (Go) to an ultimate strength (N value) is an interesting approach to help identify tropicals soils since the low strain modulus from seismic tests reflects the weakly cemented structure of lateritic soils while the SPT sampler penetration brakes down all cementation. The preliminary results from the seismic SPT test indicate that this hybrid test opens up new possibilities for geotechnical site characterization of tropical soils, based on the relationship Go/N, which is similar to the Go/qc ratio in the SCPT test.

0 10 20 30 40 50 60

(c)(d)

2

6

Lateritic Soils

10

The N60 values measured during the S-SPT test carried out at the studied site are presented in Figures 9.b. This hybrid test allowed determined Vs simultaneously to N every 1 m interval (Figure 8.c) for calculating Go (Figure 8.d). The Go/N60 values versus depth are also presented in Figure 8.e for the studied site with no averaging. SPT profile

Upper bound (cemented geomaterials)

Lower bound (cemented geomaterials)

The S-SPT testing data

4.5

Vs - 6,0m Vs - 8,0m Vs - 10,0m Vs - 12,0m Vs avr

ACKNOWLEDGEMENTS

The authors gratefully acknowledge FAPESP (State of São Paulo Research Foundation) and CNPq (National Council for Scientific and Technological Development).

Figure 8. S-SPT testing data and Go/N60 for the studied site.

4.6

Ratio (Go/pa)/N60

An S-SPT test was carried out using this system in the studied site. Seismic data were recorded from waves generated every one meter depth up to 21 m, right after the N SPT measurement using the equipment described in the previous sub-item. A six box arrangement was placed on the ground surface after removing the top soil to enable better coupling. The distance between each box (which contains a pair of geophone) was 1.5 m and they were all placed between 4.5 m to 12.0 m away from the test borehole.

The Go/N ratio

The Go/N60 profile (Figure 8.e) obtained from the S-SPT test data are similar to what was found when averaging all SPT and seismic test data (Figure 4.e) for the top lateritic layer (1 to 6 m depth) with a lower average Go/N60 equal to 27, a bit lower than what was previously found, 35. In the lower part of the lateritic layer (6 to 13 m depth) it was found an average Go/N60 equal to 14, also lower than what was previously found (23) and the same average value for the saprolitic layer. These data were also plotted in the Go/N60 versus (N1)60 chart as shown in Figure 9. All the data points are in the upper bound for cemented sands. In this case the difference between lateritic and saprolitic soils is not so clear, just the upper portion of the lateritic layer reflects a higher degree of cementation. Soil variability in this particular site probably related to the morphogenetic and pedogenetic processes, already pointed out by Giacheti et al (2003) and Giacheti & De Mio (2008) could explain the observed differences.

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7

REFERENCES

Bang, E. S. & Kim, D.S. 2007. Evaluation of shear wave velocity proﬁle using SPT based up-hole method, Soil Dynamics and Earthquake Engineering 27, p. 741-758. Giacheti, H. L. & De Mio, G. 2008. Seismic cone penetration tests on tropical soils and the ratio Go/qc. 3rd Geotechnical and Geophysical Site Characterization Conference, ISC´3, Taiwain, v.1. p. 12891295. Giacheti, H. L.; Peixoto, A. S. P. & Marques, M. E. M. 2003. Cone Penetration Testing on Brazilian Tropical Soils. XII Panamerican Conference on Soil Mechanics and Geotechnical Engineering, Cambridge/USA, v.1. p. 397-402. Nogami, J. S. & Villibor, D. F. 1981. Uma nova classificação de solos para finalidades rodoviárias, Simpósio Brasileiro de Solos Tropicais em Engenharia, Brasil, V. 1, p. 30-41. Pedrini, R. A. A. 2012. Desenvolvimento de sistema para realização de sísmica up-hole em conjunto com sondagem SPT. M.Sc. thesis, FEB/Unesp. Bauru/SP, Brazil. Pedrini, R. A. A. & Giacheti, H. L. 2012. The seismic SPT to determine the maximum shear modulus, 4th Geotechnical and Geophysical Site Characterization Conference, ISC´4, Brazil, CD-Rom. Schnaid, F.; Lehane, B. & Fahey, M. 2004. In situ test characterization of unusual geomaterials. 2nd Geotechnical and Geophysical Site Characterization Conference, ISC´2, Portugal, v. 1. p. 49-74. Viana da Fonseca. A. & Coutinho, R. Q. 2008. Characterization of residual soils, 3rd Geotech. and Geoph. Site Characterization Conf., ISC´3, Taiwain, v. 1. p. 195-249.

Compressibility Parameters of Cohesive Soils From Piezocone Paramètres de compressibilité de sols cohésifs au piézocone Hamza M.

Faculty of Engineering, Suez Canal University & Chairman of Hamza Associates, Egypt

Shahien M.

Faculty of Engineering, Tanta University, Egypt

ABSTRACT: Drained compressibilty parameters for cohesive soils can be determined by carrying out one dimensional consolidation tests on “undisturbed” samples. The compressibility parameters include the compression and recompression indices, overconsolidation ratio and coefficient of consolidation. Some of these parameters or in other forms have been already correlated in the literature to results of piezocone. The aim of this paper is to provide additional data on drained compressibility parameters, focusing on constrained modulus and overconsolidation ratio, for cohesive soils from geotechnical investigations in seven major sites of river Nile Delta deposits in Egypt where piezocone CPTU data are also available. The results of consolidation tests are used to evaluate and modify the available correlations(s) with CPTU data. It is believed that the data and analysis in this paper shall be a valuable contribution to the literature by providing a better ground for improving the current state of the art of estimating the compressibility parameters from the CPTU data. RÉSUMÉ : Les paramètres de compressibilité drainée pour les sols cohérents peuvent être déterminés en exécutant un test de consolidation unidimensionelle sur les échantillons « intacts ». Ces paramètres incluent les indices de compression et de recompression, le taux de surconsolidation et le coefficient de consolidation. Certains de ces paramètres ont déjà été corrélés dans la bibliographie aux résultats du piézocone. L'objectif de cet article est de fournir des données supplémentaires sur les paramètres de compressibilité drainée en se concentrant sur le module contraint et sur le taux de surconsolidation pour des sols cohérents étudiés dans sept sites majeurs des dépôts du Delta de Nil en Egypte, où des données de CPTU sont aussi disponibles. Les résultats d’essais de consolidation sont utilisés pour évaluer et modifier les corrélations disponibles avec les données de CPTU. On estime que les données et l'analyse présentées ici seront une contribution valable à la bibliographie en fournissant de meilleurs fondements pour améliorer l’état de l’art actuel concernant l'estimation des paramètres de compressibilité à partir de données de CPTU. KEYWORDS: constrained modulus, overconsolidation ratio, sample quality designation, piezocone, clay 1

The aim of this paper is to provide additional data on both constrained modulus and overconsolidation ratio as determined from oedometer consolidation tests on “undisturbed” samples of cohesive soils and CPTU data from seven sites from the Nile Ddelta deposits. The authors believe that the addition of the data presented in this paper to the literature provides a better ground for improving the current state of the art of estimating drained compressibility parameters from the CPTU data. With such belief, the data are used to evaluate and modify the available correlations.

INTRODUCTION

Drained compressibilty parameters for cohesive soils are useful in; a) carrying out long term settlement analysis, b) providing key parameters for analysis and design of ground improvement, and c) profiling undrained shear strength parameters with the aid of other insitu field investigation equipments such as field vane and piezocone. Drained compressibility parameters for cohesive soils can be determined from End of Primary (EOP) void ratio versus effective stress relationship that results from carrying out incremental load one dimensional consolidation tests on “undisturbed” samples. The drained compressibility parameters include the compression and recompression indices, overconsolidation ratio and coefficient of consolidation. These parameters can be influenced with variable degrees by quality of samples used in the tests. (Jamiolkowski et al., 1985 and Terzaghi et al., 1996). Empirical correlations to estimate these parameters or equivalent in other forms, from insitu tests such as piezocone are available in the literature (e.g. Jamiolkowski et al, 1985, Lunne et al., 1997 and Mayne, 2009). Availability of such correlations provides a great aid for geotechnical engineers to estimate such parameters in continuous profiles for a site in relatively short period of time and perform fewer consolidations tests for confirmation. However, estimating drained parameters from undrained piezocone test results could be complicated and sometimes may have various degrees of uncertainties (Lunne etl. 1997). Therefore, there is a need for continuous feed of data from local experiences to confirm, validate, and even modify the existing correlations.

2

INVESTIGATED SITES

Comprehensive geotechnical investigation campaigns were carried out in seven sites of major projects along the north coast and within the Delta of the Nile River of Egypt. The seven sites provide full coverage of the Nile Delta deposits starting from Idku at west of the Nile Delta, to Metobus within the Nile Delta, to Damietta, to El-Gamil and Port Said further east of the Delta. Three of these sites were reported in Hight et al. (2000), Hamza et al. (2002), (2003) and (2005). The seven sites were used by Hamza and Shahien (2009) to investigate the correlations of estimating the efective stress friction angle from piezocone data. The stratifications of the sites are shown in Fig. (1). The statification of the sites consists of silty sand top layer over very soft to medium stiff clay layer over sand over stiff to hard clay. The thickness of the soft clay layer tends to thicken as moving from west to east of the Delta (Hamza et al., 2005).

539


Silty Sand Stiff Clay

0 10

Depth, m

20 30 40

Damietta 2

Damietta 3

Silty Sand Soft to Firm Clay

Silty Sand

50 60

Silty Clayey Sand Very Soft to Medium Stiff Clay Silty Clayey Sand Stiff to Hard Clay Silty Clayey Sand Hard Clay

Overconsolidation Ratio, OCR

Stiff Clay Damietta 4 Very Soft to Medium Stiff Clay

Silty Sand

3.0 Idku Metobus Dammietta 3 Dammietta 4 Port Said 2 El-Gamil Dammietta 2

2.5 2.0 1.5 1.0

SQD Scale B A

0.5 0.0

0

1

2

D

C 3

4

5

E

6

7

8

9

10

Volumetric Strain at 'vo, vo, %

Stiff to Hard Clay

Fig. 2 Overconsolidation ratio (OCR) versus vo as a measure of SQD

3.3. Compression Indices and Moduli

Silty Clayey Sand Hard Clay

Figure 1. Stratigraphy of the soil formations in the seven sites.

3 COMPRESSIBILITY PARAMETERS FROM OEDOMETER TESTS 3.1. General The results of total 125 consolidation tests were used in this study. The tests were carried out on clay “undisturbed” samples that were collected by means of stainless steel thin wall Shelby tubes with cutting edge sharpened to approximately 5 o. Incremental loading procedure was utilized with a load increment ratio of 2. End Of Primary (EOP) consolidation was determined for each load increment using the Taylor method. EOP void ratio versus logarithm of effective vertical pressure (e-log ’v) curves were plotted for each test. 3.2. Overconsolidation Ratio The overconsolidation ratio, OCR, is defined as the ratio between the preconsolidation or yield pressure, ’p, to in situ effective overburden pressure, ’vo. The ’p is the pressure that distinguishes between low compressibility in the recompression range and the high compressibility in the compression range. There are several mechanisms for a deposit to demonstrate a ’p (Jamiolkowski et al., 1985 and Mayne et al., 2009). Those mechanisms include; decrease in vertical effective stress, freeze-thaw cycles, repeated wetting-drying, tidal cycles, earthquake loading, desiccation, aging, cementation or geotechnical bonding. The decrease in effective stress could be caused by; mechanical removal of overburden, overburden erosion, rise in sea level, increased groundwater elevations, glaciation, and mass wasting. The conventional and most common Casagrande method is used to determine ’p from the EOP e-log ’v curves from the Oedometer tests carried out. Sample quality was evaluated on the basis of the magnitude of the volumetric strains, vo, during reconsolidation to ’vo in

540

The compression, Cc, and re-compression, Cr, indices were calculated for each test as the slopes of the e-log ’v curve in the normally consolidated and the re-compression ranges, respectively. The recompression index, Cr, was calculated as the average slope of the unloading-reloading cycle of e-log ’v curve between vertical effective stress value of twice of the preconsolidation pressure, ’p, and effective overburden pressure, ’vo or the average slope of the unloading curve from consolidation pressure of 3200 kPa. Compression index values in this study are plotted in Figure (3) versus natural water content, the Terzaghi et al. (1996) plot for filling and reference. The water content is a major variable as it reflects how much water held in the deposit to be squeezed out upon the increase in effective stress. As expected, the data show a band that compares relatively well with data from all over the world as collected originally by Terzaghi et al. (1996). The overall average of ratio of re-compression to compression indices Cr/Cc is calculated to be about 0.1. -6

2.85

Cc=4x10 w 1

Idku Metobus Dammietta 3 Dammietta 4 Port Said 2 El-Gamil Dammietta 2

w

60

Sand with Silt Occasionaly interbeded by Hard Clay

=0 .1

50

Firm to Stiff Clay

c

Sand with Silt

Sand with Silt

C

40

05 w

Limemud

Soft to Firm Clay

=0 .0

30

Silt Soft to Firm Clay

c

Soft to Med. Stiff to Very Stiff Clay, Interbeded with Very Stiff Peat

C

Alternating Soft to Med. Stiff Clay, Organic Silt & Silty Sand

03 w

Depth, m

20

Silty Sand Sand - Occas. Silt and Clay

=0 .0

Silty Sand

10

oedometer tests as suggested by Andresen and Kolstad (1979). The Sample Quality Designation (SQD) scale using vo suggested by Andresen and Kolstad (1979) and modified by Terzaghi et al. (1996) is used in this paper. Figure (2) shows the OCR values in this study versus vo. Shown also on the plot, is the above mentioned SQD scale. The scale suggests that the majority of samples have quality B to C. Such sample qualities correspond to verbal scale of very good to good samples. The OCR values for the clay are in the range of 1 to 2. It should be noted that OCR values might be influenced by sample disturbance. As sample disturbance increases (i.e. vo increases), the OCR value decreases due to the de-structuring of the samples during sampling. One possible major source for sample disturbance in Nile Delta deposits is the natural gas exsolution in the pore water (Hight et al., 2000). The OCR values, for the very few tests, that are less than 1 were corrected to 1 for use in evaluations and correlations developed in this study.

Port Said 2 Soft Clay

c

Silty Sand

Silty Sand Silt/ Silty Sand/ Clay

C

ELGamil

Compression Index, Cc

Metobus

Idku

0

0.1 10

100

Natural Water Content, %

Figure.3 Data of this study on the compression index versus natural water content Terzaghi et al (1996) relationship

Constrained modulus is another form of compressibility parameter instead of the recompression or compression indices. The following expression is used to estimate the tangent constrained modulus: (1) M= ’v/ = 2.3(1+e)’v/Cc The general definition of constrained modulus in Equ. (1) is used in the literature (e.g. Kulhawy and Mayne 1990). There are several definitions for the constrained modulus depending on which ’v and which index, Cc or Cr, used in Equ. (1). It is expected that the modulus in the compression range is different


than that in the re-compression range. Even in the compression range, the constrained modulus is dependent on ’v level (Janbu, 1963). Figure (4) introduces the several definitions of the constrained modulus using consolidation test data from the Idku site as an example. The Janbu (1963) approach can be used to define three constrained moduli as defined in Figure (4) and Equs. (2) to (4); Mi in the recompression range, Mnp or Mn@’p at ’p and Mn in the compression range that is dependent on level of ’v: Mi= 2.3(1+e)’p/Cr (2) (3) Mnp = Mn@’p = 2.3(1+e)’p/Cc (4) Mn= 2.3(1+e)’v/Cc There are investigators (e.g. Sanglerat, 1972, and Abdelrahman et al., 2005) that are using Mo at ’vo as in Equ (5)(Fig. 4): (5) Mo= 2.3(1+e)’vo/Cc The geotechnical engineer should be cautious as what modulus is reported or estimated and how it is used in settlement analysis, because in a lot of literature the reference is given to M without specifying which modulus is meant such as in Equ. (1). Mo modulus can be used to estimate both Mi and Mn using Equs. (6) and (7) to be used for settlement analysis in the recompression and compression ranges, respectively. (6) Mi = MoOCR(Cc/Cr) (7) Mn = Mo(’v/ pa) where ’v is the average pressure between ’p and the final pressure due to surface load causing the settlement. 50000 1.5 1.4 1.3 1.2 1.1 1.0 0.9 0.8 0.7

30000

50000

10

100

1000

10000

Effective Vertical Stress, kPa

20000

'p

30000

20000

10000

0

10000

0

1

40000

0

1000

1500

2000

2500

3000

Mn

Mo 0

500


Mn-'p 100

200

300

400

500


’p = k (qt-vo)

Reference

k

Comment

Lefebvre & Poulin (1979) Mayne & Holtz (1988) Larson & Mulabdic (1991) Mayne (1991)

0.25- 0.4 0.4 0.29 0.33

Norway & UK sites World Data Scandinavian Soils

0.28 Eastern Canada Clays 0.305 205 Clay sites 0.2 – 0.5 0.65(Ip)-0.23 0.25 – 0.32 su/’p=constant interpretation 0.2 – 0.5 Port Said Site, Egypt 0.14 Louisiana Soils – 7 Sites 0.3 Beaufort Sea Clays Ko=1.5 0.24 Beaufort Sea Clays Ko=2.0 Robertson (2012) * SHANSEP & CSSM * k = [ [(qt-vo)/’vo]0.2 / (0.25(10.5+7log Fr)) ]1.25 where Fr = fs/(qt-vo)

5


k = 'p/(qt-vo)

0.8

Piezocone Penetration Tests with pore water pressure measurements (CPTU) were performed at the sites. A l0 cm2 Piezocone was used to carry out the testing. Records were made at 2 cm intervals. At each tested depth, cone resistance (qc), pore water pressures behind cone (u2) and side friction (fs) were measured. Typical CPTU records at some of the sites under study are shown in Hight et al. (2000), Hamza et al. (2003) and Hamza et al. (2005). The corrected tip resistance, qt, can be calculated as qt=qc+(1-)u2, where  is a cone factor. The net cone resistance, qn, can be calculated as qn= qtvo, where vo is the total overburden pressure.

Cavity Expansion & Critical State Soil Mechanics Analysis

Leroueil et al. (1995) Chen & Mayne (1996) Lunne et al. (1997) Mayne (2001) Mesri (2004) Abdelrahman et al. (2005) Pant (2007) Becker (2010)

1.0

PEIZOCONE PENETRATION TESTS

(8)

Table 1. Summary of the parameter k from the literature..

Figure 4 Definition of tangent constrained modulus concept

4

or OCR = ’p/'vo = k(qt-vo )/'vo

It should be noted that empirical constant k in both expressions in Equ. 8 is the same. Table (1) shows a summary of k values reported in the literature. According to the table, k is in the range of 0.14 to 0.5. Mayne (2001) showed that k is slightly dependent on plasticity index, while Becker (2010) showed that k is slightly dependent on coefficient of horizontal pressure at rest. Robertson (2012) suggested an expression that is dependent on (qt-vo)/'vo and sleeve friction ratio, Fr. The empirical constant is calculated for the data in this study and is plotted versus Fr in Figure (5). The expression suggested by Robertson (2012) was also plotted on the same plot. Figure (5) shows that the Robertson (2012) predicts well the range of k. However, it seems that k is slightly increasing with Fr. The calculated k values are in the range of 0.1 to 0.6 (0.18 to 0.4, if scatter is ignored) with an average of 0.32, which is consistent with the existing correlations in the literature.

0.6 0.4 0.2 0.0

Robertson (2012) Average k = 0.32

Range From Literature

Void Ratio

Mi

40000

Constrained Modulus, kPa

Constrained Modulus, kPa

Idku Site

and local heterogeneity. The most common and widely used correlation is (e.g. Lunne et al. 1997):

1

(qt-vo)/'vo 20 10 5 1

2

3

4

5

6

7 8 9 10

Friction Ratio, Fr = [fs/(qt-vo)] 100, %

Figure (5) Empirical constant k for the sites in this study

Ladd and De Groot (2003) proposed the following SHANSEP type of expression to estimate OCR: OCR = kOCR[(qt-vo )/'vo]1.25

(9)

Ladd and De Groot reported a value of 0.192 for kOCR based Boston Blue clay experience. Robertson (2009) suggested general kOCR value of 0.25. Robertson (2012) suggested the expression in Equ. (10) to estimate kOCR based on Fr:

PEIZOCONE PENETRATION TESTS

5.1. Stress History or Overconsolidation Ratio Review of the available correlations between ’p or OCR and Piezocone results was carried out by Lunne et al. (1997), Mayne (2001), Ladd and DeGroot (2003), Powell and Lunne (2005), Pant (2007), Mayne (2009), Becker (2010) and Robertson (2012). The cone parameters used in the correlations include qc, qt, qt-vo, qt-u2, u. Some of these parameters were used with or without normalization by ’vo. According to Campanella and Robertson (1988), there is no unique relationship between OCR or ’p and measured penetration induced pore water pressures and if exists, it is poor because the pore pressures measured is influenced by the location of the u measurement (i.e. u1, u2 or u3), clay sensitivity, over consolidation mechanism, soil type

541

kOCR = (2.625+1.75 log Fr)1.25

(10)

The data of Delta clay sites was used to back calculate kOCR and was plotted versus Fr in Fig. (6). The Robertson (2012) expression was also plotted on Fig. (6). Figure (6) shows that Equ. (10) predict well the range of kOCR. However, it seems that kOCR is slightly increasing with Fr. The average kOCR of the data in this study was about 0.23 that is consistent with data in literature.


kOCR = OCR/[(qt-vo)/'vo]

1.25

1.0


0.8 0.6

were recorded at the same depths of the samples. Such pairing allowed for comprehensive review of the existing empirical correlations to predict compressibility parameters from in-situ piezocone results. 4) The OCR of the Nile Delta clays can be best predicted using Equs. (8) and (9) using average k of 0.32 and average kOCR of 0.23. Figs (5) & (6) suggest that k and kOCR have the general tendency to slightly increase with friction ratio, Fr. 5) The Mo can be best predicted using Equ. (11) with average value of o of 3.5. Settlement analysis can then be carried out using Mi and Mn that can be calculated using Equs (6) and (7).

Average k OCR = 0.23

Robertson (2009)

Robertson (2012) Equ. (10)

0.4 0.2 0.0

Ladd & DeGroot (2003) 2

1

4

3

5

6

7

8 9 10

Fr = [fs/(qt-vo)] 100, %

Figure (6) Empirical constant kOCR for the sites in the study

5.2. Constrained Modulus Review of the available correlations between M and cone results for cohesive soil was carried out by Lunne et al. (1997), Mayne (2001), Pant (2007), and Robertson (2009). Attempts to correlate M of cohesive soils to cone results have started since mid sixties of the last century (Sanglerat, 1972). The following expression shows the general form of the empirical correlation: MSubscript = Subscript[qParameter]

(11)

The subscript in Equ (11) could be nothing, i, np, n, or o as in Equs (1 to 5). The empirical constant  as well as the cone parameter, qParameter, used in Equ (11) as reported in literature is summarized in Table (2). According to the table, o is in the range of 1 to 14. Sanglerat (1972) showed that o is inversely dependent on qc. Robertson (2009) suggested that o is directly related to (qt-vo)/’vo with an upper limit of 14. The empirical constant o is calculated for the data in this study and is plotted versus (qt-vo)/pa in Figure (7), where pa is a reference pressure of 100 kPa.. Ignoring some scatter, the calculated o values are in the range of 1 to 8 with an average of 3.5, which is consistent with the existing correlations in the literature. Sources of scatter in Figure (7) include but not limited to; sample disturbance with its influence on the measured compressibility and natural variation between the location of borehole from which the samples were extracted and that of the CPTU testing. Table (2) Summary of components of empirical Equ. (11) in literature Reference

qParameter Subscript  Range 2.3-7.7 qc o 1-8 * qc o 2.2-3.3 qc o 3.1 qt np 3.27 np qt-vo Kulhawy & Mayne (1990) 8.25 qt-vo  Senneset et al. (1989) 5-15 i qt-vo 8 np qt-vo Abdelrahman et al. (2005) 1.25 o qt-vo Mayne (2009) 5 qt-vo  Robertson (2009) ** o qt-vo * Dependent on type of soil and on qc values ** For Clays (Ic > 2.2) o= (qt-vo)/’vo o ≤ 14

Comment

Bachelier and Parez (1965) Sanglerat (1972) Jones & Rust (1995) Pants (2007)

Flanders Clay France & Spain Clays South African Clays Louisiana Clay Louisiana Clay Glava Clay Glava Clay Port Said Clay Vanilla Clays

30 Idku Metobus Dammietta 3 Dammietta 4 Port Said 2 El-Gamil Dammietta 2

Average o = 3.5

20 15

Range from literature

o = Mo/(qt-vo)

25

10 5 0

0

2

4

6

8

10

12

14

(qt-vo)/pa

Figure (7) Empirical constant o for the sites in the study

6

SUMMARY AND CONCLUSIONS

1)

The results of geotechnical investigations in seven sites in the Nile Delta clays were used in this paper. 2) The compressibility parameters; OCR, Cc and Cr, and Mo, were calculated from EOP e-log’v curves of total 125 consolidation tests carried out on “undisturbed” samples. The SQD of the majority of the samples was B to C. 3) The compressibility parameters of each test were paired with results from neighboring or adjacent piezocone test that

542

7

REFERENCES

Abdelrahman M., Ezzeldine O.and Salem M. 2005. The Use of Piezocone in Characterization of Cohesive Soil West of Port Said – Egypt, Proc. of 5th Int. Geot. Eng. Conf.,– Cairo University – Egypt, pp. 201-219. Bachelier M. and Parez L.1965. Cont ribution a letude de la compress ibilite’ des sols a l’aide du penetrometer a cone, Proc. 6th Int. Conf. Soil Mech. Found. Eng., Montreal, 2, 3-10.

Becker, D. E. 2010. Testing in Geotechnical Design, Geot. Eng. Jour. of the SEAGS & AGSSEA, Vol. 41, No. 1, pp. 1-12. Campanella, R.G. and Robertson P. K. 1988. Current status of piezocone test, Proc. of Int. Symp. on Penetration Testing, Orlando, USA, Vol. 1, pp. 1-24. Chen B. and Mayne P.W. 1996. Statistical relationships between piezocone measurements & stress history of clays, Can. Geot. Jour. 33(3), pp. 488–498. Jamiolkowski M., Ladd C.C., Germaine J.T., and Lancelotta R. 1985. New Development in Field and Laboratory Testing of Soils, Proc. of the 11th Int. Conf. Soil Mech. and Found. Eng., San Francisco, 1, pp. 57-153. Hamza M., Shahien M. and Ibrahim M. 2003. Ground characterization of Soft Deposits in Western Nile Delta, Proc. 13th Reg. African Conf. Soil Mech. Geot. Eng., Morocco. Hamza M., Shahien M. and Ibrahim M. 2005. Characterization and undrained shear strength of Nile delta soft deposits using piezocone, Proc. 16th Int. Conf. on Soil Mech. and Geot. Eng., Osaka, Japan Hamza M. and Shahien M. 2009. Effective stress shear strength parameters from piezocone, Proc.17th Int. Conf.Soil Mech. and Geot. Eng., Alexandria, Egypt. Hight D.W. Hamza M.M. and ElSayed A.S. 2000. Engineering characterization of the Nile Delta clays, Proc. of IS Yokohama 2000. Janbu N. 1963. Soil compressibility as determined by oedometer and triaxial tests, Proc. European Conf. Soil Mech. and Found. Eng. Wiesbaden, 1, 19–25. Jones G.A. and Rust E. 1995. Piezocone settlement prediction parameters for embankments on alluvium, Proc. Int. Symp. Cone Penetration Testing, Linköping, Sweden, 2, 501–8. Ladd, C. C. and DeGroot D. J. 2003. Recommended Practice for Soft Ground Site Characterization, Proc. 12th Panamerican Conf. Soil Mech. and Geot. Eng., Cambridge, USA Larson, R., and Mulabdic, M. 1991. Piezocone tests in clays. Swedish Geotechnical Institute report no. 42, Linkoping, 240p. Lefebvre, G. and Poulin C. 1979. A new method of sampling in sensitive clay , Canadian Geot. Journal, Vol. 16, pp. 226–233. Leroueil S., Demers D., La Rochelle P., Martel G. and Virely D. 1995. Practical use of the piezocone in Eastern Canada clays , Proc. Int. Symp. on Cone Penetration Testing, Linköping, Sweden, 2, 515–522. Lunne T., Robertson P.K., and Powell J.J.M. 1997. Cone Penetration Testing in Geotechnical Engineering Practice. p. 312. Mayne, P.W. 1991. Determination of OCR in clays by piezocone tests using cavity expansion and critical state concepts. Soils and Foundations 31 (1): 65-76. Mayne P. W. 2001. Stress-strain-strength-flow parameters from enhanced in-situ tests, Proc. Int. Conf. on In-Situ Measurement of Soil Properties & Case Histories, Bali, Indonesia, pp. 27-48. Mayne P. W., Coop M. R., Springman S. M., Huang A. and Zornberg J. G. 2009. Geomaterial behavior and testing, State of the Art Lecture, Proc. 17th Int. Conf. on Soil Mech. and Geot. Eng. Alexandria, Egypt, Vol. 4, pp. 1-96. Mayne P.W., Holtz R.D. 1988. Profiling stress history from piezocone soundings, Soils and Foundations, Vol. 28(1), pp. 16–28. Mesri G. 2001. Undrained shear strength of soft clays from push cone penetration test , Geotechnique 51, No. 2, pp. 167–168. Pant R. R. 2007. Evaluation of Consolidation Parameters of Cohesive Soils Using PCPT Method. MSc Thesis, Louisiana State University. USA Powell, J. J. M. and Lunne T. 2005. Use Of Cptu Data In Clays/Fine Grained Soils, Studia Geotechnica et Mechanica, Vol. XXVII, No. 3–4, pp. 29-66. Robertson, P. K. 2009. Interpretation of cone penetration tests – a unified approach, Canadian Geotechnical Journal, Vol. 46, pp. 1337-1355. Robertson P.K. 2012. Interpretation of in-situ tests – some insights, Proc. 4th Int. Conf. Geot. & Geoph. Site Characterization, ISC’4, Brazil, 1, pp 1-22. Sanglerat G. 1972. The penetrometer and soil exploration, Elsevier, 464 pp. Senneset K., Sandven R. and Janbu N. 1989. The evaluation of soil parameters from piezocone tests, Transportation Research Record, No. 1235, 24–37. Terzaghi K., Peck R.B. and Mesri G. 1996. Soil Mechanics in Engineering Practice, 3rd Ed. John Wiley and Sons, Inc., p. 549.

Comportement de la structure de sol amélioré par inclusions rigides, supportant une éolienne Behaviour of soil foundation improved by rigid columns, supporting a wind turbine Haza-Rozier E., Vinceslas G.

Cete Normandie-Centre/DERDI/CER

Le Kouby A.

Université Paris Est/IFSTTAR

Crochemore O. Theolia France

RÉSUMÉ: Dans le cadre du projet national ASIRi (Amélioration des Sols par Inclusions Rigides, 2006-2011), le CER (Centre d’Expérimentation et de Recherche du Cete Normandie-Centre) a instrumenté la structure de fondation d’une éolienne. Le principe de fondation est tel que l’éolienne est fixée sur une semelle rigide, coulée sur une couche granulaire de répartition de charges, déposé sur le sol en place, amélioré par 84 inclusions rigides. Des déplacements verticaux, des pressions totales transmises en tête d’inclusion et sur le sol sont mesurés. Les capteurs sont installés principalement en périphérie de l’éolienne, dans les zones qui doivent supporter les variations de contraintes les plus élevées. De plus, des extensomètres à cordes vibrantes sont installés dans deux inclusions pour accéder aux descentes de charges. Le comportement de la structure est suivi depuis la phase de terrassement, jusqu’au montage de la machine, puis durant le fonctionnement de l’éolienne, jusqu ‘à aujourd’hui. La plateforme de travail induit un confinement important des têtes d’inclusions. Les pressions se concentrent en périphérie et les déplacements restent faibles. L’effet de la vitesse du vent sur la distribution des contraintes est tout à fait significatif. ABSTRACT: Within the French National Project ASIRi (Soil reinforcement with rigid inclusions, 2006-2011), CER (Experimentations and Researches Centre, Rouen) instrumented foundations of a wind turbine. The foundation principle is such that wind turbine is fixed on a rigid slab, lying on a granular layer, allowing strength distribution on in-situ subgrade improved by 84 rigid columns. Vertical displacements and total stress sensors at the head of columns and on soil are measured. Sensors are placed on wind turbine edge essentially, in areas supporting highest stresses variations. Moreover, vibrating wire extensometers are positioned in two columns in order to measure load distribution. Behaviour of the structure has been monitored since excavation stage, till machine construction, and then during service working of the turbine wind until now. Working platform induced an important confinement of columns’ heads. Pressures are concentred on edge, displacement are still small. The effect of wind speed on load distribution has been shown to be significant. KEYWORDS: Instrumentation, rigid columns, foundation, wind turbine MOTS-CLÉS: Instrumentation, inclusions rigides, fondation, éolienne 1

INTRODUCTION

Lorsqu'une éolienne est construite sur un terrain de caractéristiques mécaniques médiocres, elle est traditionnellement fondée sur un réseau de pieux fixés sur sa semelle de fondation. Cependant, un nouveau type de fondation fait face à cette technique plus traditionnelle : la semelle de l'éolienne repose sur une épaisseur de sol, constituant une couche de répartition de charges, qui surmonte un réseau d'inclusions rigides (IR). La qualité mécanique du sol en place et sa capacité portante sont améliorées par la présence des IR. L’éolienne est alors construite avec une fondation superficielle, sur un terrain de bonne portance (Figure 1).

d’amélioration de sol par IR fonctionne (Briançon 2002), le CER a instrumenté un tel système de fondation. Construite par Theolia, dans un parc Boralex, dans la région de Neuchâtel en Bray (76), l’éolienne 3.6 a été suivie depuis sa construction en 2009 (Haza-Rozier 2011), sa mise en service en août 2010, jusqu’à ce jour (Haza-Rozier & al. 2012). Cet article décrit l’instrumentation mise en place, l’essentiel des résultats de mesures et une approche de l’effet du vent sur le comportement de la structure de fondation. 2 2.1

DESCRIPTION DE L’OUVRAGE Profil géotechnique

La coupe géotechnique locale est constituée de 1,6 m de terre végétale et de limon, puis de l’argile à silex jusqu’à 10,60 m de profondeur (avec des valeurs de module pressiométrique entre 2,2 et 18,1 MPa). Apparaît alors une frange de craie altérée sur 1 m (module pressiométrique entre 2,8 et 12,6 MPa), puis la craie de plus en plus saine, jusqu’à 20 m de profondeur (module pressiométrique entre 44 et 200 MPa).

Béton de propreté

Matelas de répartition

2.2 Figure 1. Semelle de fondation de l’éolienne.

Dans le cadre du projet national ASIRi (Amélioration des Sols par Inclusions Rigides, 2006-2011), du Réseau Génie Civil et Urbain, pour mieux comprendre comment une structure

543

Structure de fondation de l’éolienne 3.6

L’éolienne est constituée d’un mât de 78 m, fixé sur une semelle rigide de 18m de diamètre et de 2,5 m d’épaisseur en son centre (en béton, fortement ferraillée). Cette dernière est posée sur une couche granulaire de répartition de charges, de 80 cm d’épaisseur (sol 0/90 mm sur 70 cm et 0/31,5 mm sur 10 cm en


partie supérieure), dont les caractéristiques mécaniques ont été déterminées à la boite de cisaillement de grande dimension (500 x 500 mm), du CER : C = 63,7 kPa et  = 53,3°. Cette couche granulaire est mise en place sur le sol renforcé par 84 inclusions rigides, de 8 m de long et 0,36 m de diamètre (Figure 2).

descente de charges. Les capteurs sont relevés automatiquement toutes les 6 heures depuis le début de la construction. Les mesures sont faites depuis la phase de terrassement, jusqu’au montage de la machine, sur une période de six mois. Puis, le suivi est mené après la mise en service de l’éolienne, plus de deux ans après la pose de l’instrumentation. direction des vents dominants semelle « comprime » le sol

17 sur CPT1 18 sur CPT11 T13 sur CPT2 T14 entre CPT2 et B

T12 sur 15

Plateforme de grutage

15 sur CPT4

B

T11 sur CPT5 T10 sur 16 16 sur CPT9

19 sur CPT6 T15 sur 19 T16 entre CPT6 et C C

G Rampe d’accès

Figure 2. Forage et coulage des IR en fond de fouille. 3

direction des vents dominants semelle « se soulève » du sol

Figure 3, 14 CPT et 8 T) et un second en partie haute de cette couche, sous la semelle de l’éolienne (Figure 4, 8 CPT et 11 T).

T0 IR ref

(dans buse)

CPT10

CPT1

CPT11 CPT3 CPT4 T5 T4 CPT5

CPT2 T1 T2 T3

CPT12

CPT13

CPT9 4 bars

E

D

T17 entre A et E 20 sur A T18 entre A et D

Contrainte sous socle (CPT) 3 bars Tassement sous socle (CED)

Figure 4. Plan d’instrumentation en partie supérieure de la couche de répartition, sous la semelle (2ème niveau).

4

RESULTATS

Les mesures sont acquises pendant la construction de l’ouvrage et après sa mise en service. Les données propres de l’éolienne (vitesse du vent, orientation de la nacelle, puissance produite) sont acquises depuis le printemps 2012 et permettent une première observation de l’effet du vent sur le comportement de la structure de fondation. 4.1 Transfert des efforts pendant la construction

direction des vents dominants semelle « comprime » le sol

AA

CPT6

CPT7 T7

21sur CPT8 T20 sur CPT 7

L’instrumentation est composée de capteurs électriques de déplacement et de capteurs de pression totale (notés respectivement T et CPT dans la suite), positionnés en tête d’inclusion et dans le sol. Ils sont installés sur deux niveaux : un premier, sous la couche de répartition, au niveau des têtes d’IR (

T8 Rampe CPT8 d’accès CPT14

entre CPT8 et G

T19 sur CPT 14

INSTRUMENTATION

Plateforme de grutage

22

T6

direction des vents dominants semelle « se soulève » du sol

Contrainte sur sol (CPT) 4 bars Effort sur tête IR (CPT) 10 bars et 4 bars Tassement au niveau des têtes d’IR (CED) Déformation dans les IR (extensomètres à corde vibrante)

Figure 3. Plan d’instrumentation au niveau des têtes d’IR (er niveau).

Les capteurs sont installés principalement en périphérie de la semelle de l’éolienne, zone qui doit supporter les contraintes et les variations de contraintes les plus élevées. Ils sont ainsi disposés sur la ligne des vents dominants, afin de mesurer les plus forts effets du vent. De plus, des extensomètres à cordes vibrantes (notées CV) sont installés dans deux inclusions pour y déterminer la

544

Depuis la mise en place des CPT, directement sous la semelle de l’éolienne, leurs indications ne dépassent pas 65 kPa jusqu’à la mise en service. Par contre, les pressions totales mesurées sur 9 IR, avant la mise en service de l’éolienne (Figure 5), augmentent fortement lors du coulage de la semelle, sauf en son centre (CPT9), qui reste très peu sollicité. Les mesures des CPT1 et CPT2 (comme les CPT7 et CPT8) différent dès ce moment, malgré la proximité des capteurs, mais leur évolutions restent cohérentes. L'IR située sous le CPT5 est positionnée sous l’influence d’un des appuis de la virole (qui supportera la base du mât de l'éolienne) ; cela explique pourquoi elle est très fortement chargée. Les différences notables entre les valeurs mesurées par les capteurs s’initient à cette phase de la construction, pour se maintenir jusqu’à la mise en service de l’éolienne, qui va subir de fortes sollicitations avant cette date (tests d’arrêt d’urgence, survitesse). Les pressions mesurées sur les têtes d’IR s’échelonnent entre 360 kPa (36,6 kN sur l’IR) à 796 kPa (81 kN), pour une moyenne de 596 kPa. Les 34 IR situées en périphérie de la semelle supportent 16 % du poids statique de l’éolienne. La pression exercée sur le sol varie entre 100 et 200 kPa, sans être maximale en périphérie de la semelle. En fin de construction, le sol situé dans une frange de 0,5 m en périphérie de la semelle, reprend 24,4 % du poids de l'éolienne (en considérant une pression moyenne de 120 kPa exercée).


Figure 5. Pressions totales sur les IR pendant la construction.

Ainsi, cette frange périphérique (soit 22,5 % de la surface totale de la semelle), reprend 40,4 % de la charge statique totale. 4.2

Tassement

Les tassements sont calculés par rapport à un capteur de référence, assurément immobile, afin d'annuler les effets des variations hydriques et de température de la zone expérimentale. La Figure 6 présente le tassement du sol de fondation au niveau d’élévation des têtes d'IR.

Figure 7. Variation des pressions sur IR après mise en service.

Après mise en service, les pressions exercées sur le sol entre IR, bien que plus chaotiques que durant la construction, ne fluctuent quasiment pas. Un suivi de plusieurs années permettra de confirmer un tel comportement. 4.4

Déformation d’une inclusion rigide

Des extensomètres à cordes vibrantes (CV) sont installés dans deux IR, à des profondeurs différentes, pour accéder aux déformations de l’IR et ainsi au mécanisme de transfert des efforts dans la colonne. Seuls deux CV ont pu être mesurés dans la durée, malgré un fort bruit dû aux vibrations de l’ouvrage. Sur la figure 8 sont tracées les déformations mesurées en partie supérieure et à 1,36 cm de profondeur de l’IR coiffée par le CPT2. La mesure de la pression exercée sur cette IR est également présentée.

Figure 6. Tassement du sol de base, au niveau des têtes d’IR.

Au cours de la construction, le sol de fondation tasse légèrement plus que les IR, pour se stabiliser, au moment de la mise en service, à moins de 17 mm (et 10 mm pour les IR). Les variations relevées depuis n’excèdent pas  6 mm sur le sol et en surface de la couche de répartition de charge. 4.3

Transfert des efforts après mise en service

Figure 8. Déformation d’une IR et pression exercée sur sa tête (CPT2).

Après la mise en service de l'éolienne, les mesures montrent des charges aléatoires, du fait des variations de charges engendrées par le vent et la rotation des pales. La variation des pressions exercées sur les IR, depuis la mise en service de l’éolienne (Figure 7), est similaire pour toutes les IR. Les variations saisonnières sont plus en périphérie de la semelle qu’en se rapprochant du centre de la fondation. Ces variations de pression peuvent avoir une amplitude de 400kPa entre été et hiver.

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Dès le coulage de la semelle, l’IR s’est déformée, de façon plus accentuée en profondeur. Cela peut s’expliquer car la plateforme de travail (de 30 cm d’épaisseur, mise en place pour permettre l’accès à la foreuse en fond de fouille) enserre la tête de l’IR et l’empêche ainsi de se déformer. Elle reporte les efforts qu’elle reçoit à sa surface, comme ceux que lui transmet l’IR, plus en profondeur, sous sa base. Une déformation mesurée plus importante en profondeur peut faire apparaître la présence de frottement négatif le long de la colonne. Les fluctuations de pression exercée sur l’IR sollicitent également l’IR plus en profondeur.


4.5

Effets du vent

La vitesse du vent et l’orientation de la nacelle sont enregistrées en continu. En admettant que les pâles sont toujours perpendiculaires à la direction du vent, on peut accéder à la valeur de la projection de la vitesse du vent sur l’axe des vents dominants. En la superposant à celle du vent, on visualise si le vent souffle dans cet axe ou non. Lorsque la vitesse du vent dépasse 8 m/s, le pas d’acquisition des mesures des capteurs est automatiquement réduit à moins d’une minute, par le biais d’une deuxième centrale d’acquisition. Cela permet d’accéder à leur variations, exclusivement au cours d’un vent violent. Notons que seules les CPT répondent instantanément en mesure rapide, alors que les CV et les T nécessitent plusieurs secondes pour se stabiliser. 4.6

amélioré par inclusions rigides, a permis de suivre l'évolution des tassements, pressions exercées et déformation d’IR et du sol, au cours de la construction et pendant la vie de l’ouvrage. En fin de construction, les efforts se concentrent légèrement sur la périphérie de la semelle de fondation, ce qui montre un fonctionnement en fondation rigide.

Variation des pressions appliquées

En s’intéressant aux capteurs positionnés dans l’axe des vents dominants, on peut observer l’évolution des pressions appliquées, en fonction de la vitesse et du sens du vent (Figure 9). Il s’avère que, dans cet axe, lorsque la nacelle change de sens, les pressions exercées au niveau des têtes d’IR augmentent d’un côté de la semelle pour diminuer de l’autre (entouré A sur la Figure 7). Par vent fort, les variations respectives de contraintes atteignent 200 kPa (entouré B). Dans une moindre ampleur, les pressions exercées sur le sol présentent le même type de variations.

Figure 10. Pression au niveau des têtes d’IR (sol et IR), dans l’axe des vents dominant; vitesse du vent et projection sur l’axe.

Une fois en service, les pressions varient plus amplement et évoluent avec les variations de température saisonnières, qui impactent la structure de l’éolienne. La poursuite des mesures permettra d’observer éventuellement une tendance de comportement de la structure. Les tassements du sol et des IR restent faibles. La tête des IR a tassé d’1 cm, alors que le sol en place a tassé d’environ 2 cm en fin de construction. En service, l’amplitude des tassements ne dépasse pas 5 mm. Les déformations internes des IR sont difficilement accessibles par le système d’acquisition installé. Cependant, le mécanisme observé dans la partie supérieure d’une IR montre que la présence de la plateforme de travail, qui enserre la tête des IR, a un effet réel sur le comportement mécanique de la structure, en favorisant le développement de frottement négatif le long des colonnes, ce qui ramène plus en profondeur le point neutre de fonctionnement de la colonne. Nous poursuivons l’analyse des mesures de tassement et de déformation d’une IR pour accéder au comportement d’une IR au cours d’un fort coup de vent. 6

Figure 9. Pression au niveau des têtes d’IR (sol et IR), dans l’axe des vents dominants; vitesse du vent et projection sur l’axe.

4.7 vent

Évolution des tassements en fonction de la vitesse du

Sur la Figure 10 sont superposées la puissance développée par l’éolienne, des mesures de CPT et les mesures de tassement sous la semelle de l’éolienne et sur le sol au niveau des têtes d’IR, en périphérie de l’éolienne, sur l’axe des vents dominants. Lorsque l’éolienne ne produit pas d’électricité (puissance nulle, entouré sur la figure 10), les pressions sous l’ouvrage diminuent et le sol semble se relaxer, avant de retrouver sa position lorsque la rotation des pâles reprend. La présence d’eau en pied de semelle peut expliquer ce phénomène car elle n’est plus chassée par l’effet dynamique des battements dus aux rafales de vent. 5

CONCLUSION

L'instrumentation du chantier d'une éolienne du parc éolien de Boralex, au nord de Rouen, fondée sur une structure de sol

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REMERCIEMENT

Les auteurs remercient l’Agence Nationale pour la Recherche pour l’aide financière apportée au montage du projet national ASIRi, géré par l’IREX, dans le cadre du RGCU, ainsi que les sociétés Theolia et Boralex qui ont assumé une part importante du coût de l’instrumentation. Ils remercient également la société ANTEA, membre du projet national pour le montage du projet, comme la société Egis, intervenue lors du montage expérimental. 7

REFERENCES

Briançon L. 2002. Renforcement des sols par inclusions rigides, État de l’art en France et à l’étranger, Irex, Opération du réseau Génie Civil et urbain, septembre, 180 p. Haza-Rozier E. Vinceslas G. Le Kouby A. et Duprez T. 2012. Instrumentation des fondations d’une éolienne – Cas des inclusions rigides, Journées Nationales de Géotechniques et de Géologie de l'ingénieur, JNGG, 4-6 juillet 2012, Bordeaux, France, 561-568. Haza-Rozier E. 2011. Projet national ASIRi – Éolienne fondée sur inclusions rigides, rapport de recherche Cete NormandieCentre/CER, juillet, 28 p.

Seismic Response of Superstructure on Soft Soil Considering Soil-Pile-Structure Interaction Influence de l'Interaction sol- pieu- structure sur la réponse sismique de la superstructure sur sol mou Hokmabadi A.S., Fatahi B., Samali B.

School of Civil and Environmental Engineering, University of Technology Sydney (UTS), Broadway NSW 2007, Australia

ABSTRACT: This paper presents results of shaking table tests and three dimensional numerical simulations to investigate the influence of Soil-Pile-Structure Interaction (SPSI) on the seismic response of mid-rise moment resiting buildings supported by endbearing pile foundations. Three different cases have been considered, namely: (i) fixed-base structure representing the situation excluding the soil-structure interaction; (ii) structure supported by shallow foundation on soft soil; and (iii) structure supported by end-bearing pile foundation in soft soil. Comparison of the numerical predictions and the experimental data shows a good agreement confirming the reliability of the numerical model. Both experimental and numerical results indicate that soil-structure interaction induces significant increase in the lateral deflections and inter-storey drifts of the structures on both shallow and end-bearing pile foundations in comparison to the fixed base structures. This increase in the lateral deformations and in turn inter-storey drifts can change the performance level of the structure during earthquakes which may be safety threatening. RÉSUMÉ : Cet article présente les résultats des essais sur table vibrante et trois dimensions simulations numériques pour étudier l'influence de l'Interaction sol-pieu-structure (ISPS) sur la réponse sismique des bâtiments pris en charge par les fondations sur pieux. Trois cas différents ont été examinés, à savoir: (i) la structure de base fixe sans interaction sol-structure; (ii) la structure soutenue par la fondation superficielle sur sol mou; et (iii) la structure soutenue par la fondation sur pieux dans le sol mou. Les prédictions numériques et les données expérimentales montrent un bon accord. Résultats expérimentaux et numériques indiquent que l'interaction sol-structure augmente les déflexions latérales et les dérives inter étage des structures en comparaison avec les structures de base fixes. Cela peut changer le niveau de performance de la structure lors de tremblements de terre qui peuvent être un problème d'innocuité. KEYWORDS: soil-pile-structure interaction, seismic response, shaking table test, FLAC3D, end-bearing pile foundation 1

springs and dashpots are employed to represent the soil behaviour (e.g. Hokmabadi 2012); (ii) Elastic Continuum Methods, which are based on Mindlin (1936) closed form solution for the application of point loads to a semi-infinite elastic media; and (iii) Numerical Methods. The substructure methods are the simplest and most commonly used methods, however, these methods adopting the substructuring concept rely on the principle of superposition, and consequently, are limited to either the linear elastic or the viscoelastic domain (Pitilakis et al. 2008). The dynamic equation of motion of the soil and structure system can be written as:

INTRODUCTION

The problem of soil-pile-structure interaction in the seismic analysis and design of structures has become increasingly important, as it may be inevitable to build structures at locations with less favourable geotechnical conditions in seismically active regions. Influence of the underlying soil on seismic response of the structure can be ignored if the ground is stiff enough, and the structure can be analysed considering fixedbase conditions. However, the same structure behaves differently when it is constructed on the soft soil deposit. Earthquake characteristics, travel path, local soil properties, and soil-structure interaction are the factors affecting the seismic excitation experienced by structures. The result of the first three of these factors can be summarised as free-field ground motion. However, the foundation is not able to follow the deformation of the free field motion due to its stiffness, and the dynamic response of the structure itself would induce deformation of the supporting soil (Kramer 1996). Over the past decades, several researchers (e.g. Tajimi 1969, Gazetas 1991, Shiming and Gang 1998, Hokmabadi et al. 2011, Carbonari et al. 2011, Tabatabaiefar et al. 2013) have studied the seismic soil-pile-structure interaction (SSPSI) and the effect of this phenomena on the response of the structures. The developed analytical methods for studying the soil-pile-structure interaction may be categorised into three groups: (i) Substructure Methods (or Winkler methods), in which series of

[M]{ü}+[C]{ů}+[K]{u}= -[M]{m}üg+{Fv}

(1)

where, {u}, {ů}, and {ü} are the nodal displacements, velocities and accelerations with respect to the underlying soil foundation, respectively. [M], [C] and [K] are the mass, damping, and stiffness matrices of the structure, respectively. It is more appropriate to use the incremental form of Equation (1) when plasticity is included, and then the matrix [K] should be the tangential matrix and {ü} is the earthquake induced acceleration at the level of the bedrock. For example, if only the horizontal acceleration is considered, then {m}=[1,0,1,0,....1,0]T. {Fv} is the force vector corresponding to the viscous boundaries. This vector is nonzero only when there is a difference between the

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motion on the near side of the artificial boundary and the motion in the free field (Wolf 1985). The present research aims to study the effects of SSPSI on the seismic response of the superstructure by employing the fully nonlinear method in which main components of the interaction including subsoil, pile foundation, and superstructure are modelled simultaneously. For this purpose, a threedimensional explicit finite-difference program, FLAC3D (Itasca 2009), is used to numerically model and examine the influence of the soil-structure interaction on the seismic response of a 15storey moment resiting building. Two types of foundations including shallow foundations and end-bearing pile foundations have been considered. The proposed numerical soil-structure model has been verified and validated against experimental shaking table test results. 2 2.1

Control room

Model structure

Displacement transducers

Soil mix

SHAKING TABLE EXPERIMENTAL TESTS Prototype characteristics and scaling factors

In order to provide a calibration benchmarks for the numerical simulation and to make quantitative predictions of the prototype response several of shaking table tests have been conducted. Previous researchers (e.g. Meymand 1998, Chau et al. 2009) modeled the superstructure as a simplified single degree of freedom oscillator in which the behaviour of the soil-structure system may not be completely conform to reality and the higher modes would not be captured. In the current model tests, unlike the previous efforts, a multi-storey frame for the superstructure is adopted representing most of the dynamic properties of the prototype structure such as natural frequency of the first and higher modes, number of stories, and density. The experimental model tests have been carried out utilising the 3×3 m shaking table facilities located at structures laboratory of the University of Technology Sydney (UTS). The selected prototype structure is a fifteen-storey concrete moment resisting building frame with the total height of 45 m and width of 12 m consisting of three spans, representing the conventional types of mid-rise moment resisting buildings. The spacing between the frames into the page is 4 m. Natural frequency of the prototype building is 0.384 Hz and its total mass is 953 tonnes. The soil medium beneath the structure is a clayey soil with the shear wave velocity of 200 m/s and density of 1470 kg/m3. The horizontal distance of the soil lateral boundaries and bedrock depth has been selected to be 60 m and 30 m, respectively. The building is resting on a footing which is 4 m wide and 12 m long. For the pile foundations case, a 4×4 reinforced concrete pile group with equal spacing and pile diameter of 1.25 m and 30 long are considered. The piles are embedded into the bedrock representing typical end-bearing pile foundations. In order to achieve a reasonable scale model, a dynamic similarity between the model and the prototype is applied as described by Meymand (1998). Dynamic similarity governs a condition where homologous parts of the model and prototype experience homologous net forces. Although small scale models could save cost, the precision of the results could be substantially reduced. Considering the specifications of UTS shaking table, scaling factor of 1:30 is adopted for experimental shaking table tests on the scale model which provides the largest achievable scale model with rational scales, maximum payload, and overturning moment meeting the facility limitations. 2.2

The model structure has been designed employing SAP2000 (CSI 2010) software to meet the required characteristics, and finally a 500×500×10 mm steel plate as baseplate, fifteen 400×400×5 mm horizontal steel plates as the floors and four 500×40×2 mm vertical steel plates as the columns are adopted. The completed structural model is shown in Figure 1.

Shaking table tests model components

The developed soil-structure model for shaking table tests possesses four main components including the model structure, the model pile foundations, the laminar soil container, and the soil mix. Employing geometric scaling factor of 1:30, height, length, and width of the structural model are determined to be, 1.50 m, 0.40 m, and 0.40 m, respectively. In addition, the required natural frequency of the structural model is 2.11 Hz.

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Laminar Soil container

Shaking table

Figure 1. Final setup of the shaking table tests for the structure with end-bearing pile foundation

Similar to the model structure, the model pile is subjected to the competing scale model criteria. The model piles have a diameter of 40 mm with L/d ratio of 25. By selecting a commercial Polyethylene high pressure pipe with Standard Dimension Ratio (SDR) of 7.4 the model piles fall in the range of acceptable criteria with 5% deviation from the target value for EI. The ideal soil container should simulate the free field soil response by minimising boundary effects. Since the seismic behaviour of the soil container affects the interaction between the soil and structure, the performance of the soil container is of the key importance for conducting seismic soil-structure interaction model tests successfully (Pitilakis et al. 2008). A laminar soil container with final length, width, and depth of 2.10m, 1.30m, and 1.10m, respectively, are designed and constructed for this study. The employed laminar soil container consists of a rectangular laminar box made of aluminium rectangular hollow section frames separated by rubber layers. The aluminium frames provide lateral confinement of the soil, while the rubber layers allow the container to deform in a shear beam manner. A synthetic clay mixture was designed to provide soil medium for the shaking table testing considering required dynamic similarity characteristics. Several mixtures were examined and finally the desired soil mix (60% Q38 kaolinite clay, 20% Active-bond 23 Bentonite, 20% class F fly ash and lime, and water, 120% of the dry mix) produced the required scaled shear wave velocity of 36 m/s at the second day of its cure age. Accordingly, the soil density and undrained shear strength on the second day were determined to be 1450 kg/m3 and 3.14 kPa, respectively. The shaking table tests have been carried out in three stages: fixed-base condition, shallow foundations, and end bearing pile foundations. Since the properties of the designed soil mix is time depended, the second and third stages should be carried out


in the same age in order to make the results comparable, without being interrupted by variation of the soil mix dynamic properties. Two scaled near field shaking events including Kobe, 1995, Northridge, 1994, and two scaled far field earthquakes including El Centro, 1940, and Hachinohe, 1968 are adopted. The characteristics of the mentioned benchmark earthquakes are summarised in Table 1. Displacement transducers (levels 3, 5, 7, 11, 13, and 15) and accelerometers (at levels 3, 5, 7, 9, 11, 13, and 15) were installed on the structure in order to monitor the dynamic response of the structure and to primarily measure the structural lateral displacements. The recorded accelerations can be used to check the consistency and accuracy of obtained displacements through a double integration in time domain. The final setup of the tests for the end-bearing pile foundation system on the shaking table is shown in Figure 1. Table 1. Utilised earthquake base motions Earthquake

3

Year

PGA (g)

Mw (R)

Duration (S)

Northridge

1994

0.843

6.7

30.0

Kobe El Centro Hachinohe

1995 1940 1968

0.833 0.349 0.229

6.8 6.9 7.5

56.0 56.5 36.0

4.0) for modelling the superstructure increases the execution time dramatically and leads to less accurate results. Because of the different characteristics of the soil and the superstructure/piles, sliding and separation may occur at the soil–structure interfaces. Two sets of interface elements are modelled in this study. For the shallow foundation case, the interface elements are placed between the foundation and the soil surface. However, for the pile foundation case, the interface elements were attached to the outer perimeter of the piles. It should be noted that in the pile foundation case, there is no interface or attachment between the foundation and the surface soil as some gap in the shaking table tests is considered to avoid any pile-raft behaviour. Therefore, there is not any direct stress transfer between the foundation slab and the subsoil in the pile foundation cases. The interfaces were modelled as linear spring–slider systems, while the shear strength of the interfaces was defined by Mohr–Coulomb failure criterion. The lateral and axial stiffness of the interface elements are estimated for both sets separately based on the recommended method given by Itasca (2009) to ensure that the interface stiffness has minimal influence on system compliance. Finally, fully nonlinear timehistory analysis is conducted under the influence of the scaled earthquake records and results in terms of maximum inelastic lateral deflections, determined for the three mentioned cases, are recorded.

DEVELOPMENT OF 3D NUMERICAL MODEL

Three-dimensional explicit finite-difference based program called FLAC3D (Itasca 2009) has been employed to develop the numerical model for the shaking table tests and to simulate the response under the seismic loading. Three cases including fixed-base conditions, the structure supported by shallow foundations, and the structure supported by end-bearing pile foundations have been modelled separately and the results are compared. The dimensions of the numerical models were chosen similar to the experimental tests. The reason for choosing the soil deposit thickness of 30 m for the both experimental and numerical models is that most amplification occurred within the first 30 m of the soil profile, which is in agreement with most modern seismic codes calculating local site effects based on the properties of the top 30 m of the soil profile (Rayhani and El Naggar 2008). Experience gained from the parametric study helped to finalise the adopted mesh size and the maximum unbalanced force at the grid points to optimize the accuracy and the computation speed simultaneously. The numerical grid and model components in FLAC3D are shown in Figure 2. Adjusting the boundary conditions, in the static analysis in which the system is under the gravity loads only, the bottom face of the mesh is fixed in all directions, while the side boundaries are fixed in the horizontal directions. During the dynamic time-history analysis, the earthquake acceleration is applied horizontally at the entire base, while free-field boundary conditions are assigned to the side boundaries. Solid elements are used to model the soil deposits, and Mohr-Coulomb failure criterion is adopted. In addition, Hysteretic damping of the soil is implemented using the built-in tangent modulus function as developed by Hardin and Drnevich (1972). The pile elements and superstructure are modelled with solid elements considering elastic-perfectly plastic behaviour with yielding criteria for the elements to control the possibly of inelastic behaviour in both superstructure and piles. As a calibration, a FLAC3D analysis was first conducted on a cantilever pile while the pile was fixed at one end into ground without the surrounding soil and the different lateral loads were applied on the free end of the cantilever pile. The recorded deflection from the FLAC3D model shows less than 2% difference from analytical predictions, confirming the accuracy of the model. It should be noted that using the structural elements such as beam and shell elements in FLAC3D (version

549

Connection of piles to the base plate 15‐storey model structure

Free field boundaries

Interfaces between piles and soil

Figure 2. Numerical grid and model components in FLAC3D for the structure with end-bearing pile foundation

4

RESULTS AND DISCUSSION

The average values of the 3D numerical predictions versus experimental shaking table results for the maximum lateral displacements of the fixed-base, shallow foundations, and endbearing pile foundations were determined and compared in Figure 3. Evaluation of the predicted and observed values of the maximum lateral displacements indicates that the trend and the values of the 3D numerical predictions are in a good agreement and consistent with the experimental shaking table test results. Therefore, the 3D numerical model can replicate the behaviour of the soil-pile-structure system with acceptable accuracy and is


rational and appropriate for further studies of the soil-pilestructure interaction effects. Accordingly, the maximum lateral deflection of the structure supported by end-bearing pile foundations is increased by 17% based on the experimental values and 19% based on the 3D numerical predictions in comparison to the fixed base structure. Moreover, the maximum lateral deflection of the structure supported by shallow foundation is increased by 55% based on the experimental values and 59% based on the 3D numerical predictions. Thus, pile foundations reduce the lateral drifts in comparison to the shallow foundation case. This is due to the presence of stiff pile elements in the soft soil which increase the stiffness of the ground and influences the dynamic properties of the whole system such as the natural frequency and damping. However, in comparison with the fix-based case, soil-pilestructure interaction tends to increase the lateral deformation of the structure. 15 14 13 12 11

Storey Number

10 9 8 7 6

Fixed base Numerical Results Fixed base Exp. Results Shallow foundation Numerical Results Shallow foundation Exp. Results End_bearing piles Numerical Results End_bearing piles Exp. Results

5 4 3 2 1 0

0

10

20

30

Maximum Lateral Deflection (mm)

Figure 3. Average values of maximum lateral displacements: Shaking table experimental values versus 3D numerical predictions

The corresponding inter-storey drifts of the average values of 3D numerical model are plotted in Figure 4. Inter-storey drifts are the most commonly used damage parameters, and based on FEMA (BSSC 1997) maximum inter-storey drift of 1.5% is the defined border between life safe and near collapse levels. According to Figure 4, seismic soil-structure interaction tends to increase the inter-storey drifts of the superstructure from life safe zone toward near collapse or even total collapse. 15 14 13 12

Storey Number

11 10 9 8 7 6 5 4

Fixed base

3

Shallow foundation

2

End_bearing pile foundation

1 0

0

0.5

1

1.5

2

2.5

Inter-storey Drift (%)

Figure 4. Average experimental inter-storey drifts for: (a) fixed-base structure; (b) Structure supported by shallow foundation; (c) structure supported by end-bearing pile foundation

The natural period of the system increases due to the soilstructure interaction. Therefore, such increases in the natural period considerably alter the response of the building frames under seismic excitation. This is due to the fact that the natural period lies in the long period region of the response spectrum curve. Hence, the displacement response tends to increase. 5

CONCLUSIONS

In this paper, a three-dimensional finite difference numerical model on a soil-pile-structure system has been conducted together with the experimental shaking table tests. By

550

comparing predicted and observed results, it has been concluded that the numerical modelling method is rational and is suitable for the simulation of the soil-pile-structure interaction under strong ground motions. In addition, based on the shaking table results and 3D numerical investigations it is observed that the lateral deflections of the structures siting on the end-bearing pile foundations amplified in comparison to the fixed base model (approximately 18% in this study). This amplification for the structure siting on the shallow foundations is more severe (approximately 57% in this study). Consequently, considering soil-structure interaction in both cases with and without pile foundations is vital, and conventional design procedures excluding soil-structure interaction are not adequate to guarantee the structural safety for the moment resisting buildings resting on soft soils. 6

REFERENCES

BSSC. 1997. NEHRP Guidelines for the Seismic Rehabilitation of Buildings, 1997 Edition, Part 1: Provisions and Part 2: Commentary. In: Federal Emergency Management Agency. Carbonari, S., Dezi, F., and Leoni, G. 2011. Linear soil-structure interaction of coupled wall-frame structures on pile foundations. Soil Dynamics and Earthquake Engineering 31 (9): 1296-1309. Chau, K.T., Shen, C.Y., and Guo, X. 2009. Nonlinear seismic soil-pilestructure interactions: Shaking table tests and FEM analyses. Soil Dynamics and Earthquake Engineering 29 (2): 300-310. SAP2000 v14 Analysis Reference Manual. CSI (Computers and Structures Inc.), Berkley, California. Gazetas, G. 1991. Formulas and Charts for Impedances of Surface and Embedded Foundations. Journal of Geotechnical Engineering 117 (9): 1363-1381. Hardin, B.O., and Drnevich, V.P. 1972. Shear modulus and damping in soils: desing equations and curves. Journal of the Soil Mechanics and Foundations Division 98 (7): 667-692. Hokmabadi, A.S., Fakher, A., and Fatahi, B. 2011. Seismic strain wedge model for analysis of single piles under lateral seismic loading. Australian Geomechanics 46 (1): 31-41. Hokmabadi, A.S., Fakher, A., and Fatahi, B. 2012. Full scale lateral behaviour of monopiles in granular marine soils. Marine Structures 29(1): 198-210. Tabatabaiefar, S., Fatahi, B., and Samali, B. Seismic Behaviour of Building Frames Considering Dynamic Soil-Structure Interaction. International Journal of Geomechanics (doi: 10.1061/(ASCE) GM.1943-5622.0000231). FLAC3D version 4.00 Fast Lagrangian Analysis of Continua in three dimentions, User's Manual. Itasca Consulting Group, Inc, Minneapolis, Minnesota, USA. Kramer, S.L. 1996. Geotechnical earthquake engineering. Prentice Hall. Meymand, P.J. 1998. Shaking table scale model tests of nonlinear soilpile-superstructure in soft clay. PhD PhD thesis in Civil Engineering University of California, Berkley. Mindlin, R.D. 1936. Force at a Point in the Interior of a Semi-Infinite Solid. Physics 7 (5): 195-202. Pitilakis, D., Dietz, M., Wood, D.M., Clouteau, D., and Modaressi, A. 2008. Numerical simulation of dynamic soil-structure interaction in shaking table testing. Soil Dynamics and Earthquake Engineering 28 (6): 453-467. Rayhani, M., and El Naggar, M. 2008. Numerical Modeling of Seismic Response of Rigid Foundation on Soft Soil. International Journal of Geomechanics 8 (6): 336-346. Shiming, W., and Gang, G. 1998. Dynamic soil-structure interaction for high-rise buildings. In Developments in Geotechnical Engineering, eds. Chuhan Zhang and P. Wolf John: Elsevier. 203-216. Tajimi, H. 1969. Dynamic Analysis of a Structure Embedded in an Elastic Stratum. In Proc. 4th World Conf. Earthquake Eng. Santiago, USA. 53-69. Wolf, J.P. 1985. Dynamic soil-structure interaction. Prentice-Hall, Englewood Cliffs, New Jersey.

Applicability of the RNK-method for geotechnical 3D-modelling in soft rocks Applicabilité de la RNK-méthode pour la modélisation géotechnique en 3D en roches tendres Ivšić T.

University of Zagreb, Faculty of Civil Engineering, Kačićeva 26, 10000 Zagreb, Croatia

Ortolan Ž.

J. J. Strossmayer University of Osijek, Faculty of Civil Engineering, Drinska 16a, 31000 Osijek, Croatia

Kavur B.

Institut IGH d.d., Janka Rakuše 1, 10000 Zagreb, Croatia

ABSTRACT: The RNK-method or the Reference Level of Correlation method represents a procedure for spatial engineeringgeological and/or geotechnical modeling, that was tested on many landslides in Croatia. The method gives the opportunity of differentiation of minimum shear strength zones, zones of different hydraulic conductivities, and zones of various soil densities. The application and verification of the RNK-method in soft rock formations found on the landslide area in Gorica Svetojanska (Croatia) is presented. The presentations providing the full set of relevant information needed to develop representative geotechnical profiles are also shown. The established geotechnical sliding model is verified by measurements of lateral movements in the landslide area and by results of corresponding stability analyses RÉSUMÉ : La RNK-méthode (méthode du niveau de corrélation de référence) représente une procédure de modélisation spatiale en génie géologique et/ou géotechnique, qui a été testée sur plusieurs glissements de terrain en Croatie. La méthode permet la différenciation des zones de la résistance de cisaillement minimale, des zones des conductivités hydrauliques différentes, et des zones de densité du sol diverse. L'application et la vérification de la méthode RNK aux formations rocheuses tendres, trouvés sur un site de glissement de terrain á Gorica Svetojanska (Croatie) sont présentées. On présente aussi un ensemble complet d’informations pertinentes pour développer les profils géotechniques représentatifs. Le modèle géotechnique de glissement établi est vérifié par les mesures de mouvements latéraux dans la zone de glissement, et par les résultats d’analyse de stabilité correspondante. KEYWORDS: RNK-method, plasticity index, shear strength, slope stability, spatial geotechnical model. 1 1.1

1.2

INTRODUCTION The site description

The village Gorica Svetojanska is located in hills area in northwestern part of Croatia. In last several years the intensive cracking of the walls of local church has been observed. Also, the soil movements at the slope with graveyard down the church have been noticed, as well as damages of the small mortuary structure. The church of St. Anastasia (St. Ana, “Jana” in local dialect) is situated at the plateau of narrow ridge dominating the nearby valley (Figure 1). The church at this position is mentioned in historical parish records from second half of 18th century. It was several times reconstructed and strengthened after damages caused by stronger earthquakes in late 19th century.

Local conditions

The site is in seismically active region and in Figure 2 the frequency of earthquakes (with I > 4º) in last 200 years is shown, supporting the parish records. The seismic intensities at the church location have been estimated by common attenuation function compiling the catalogue records of earthquakes with epicentres in radius R = 75 km from the site (GZ, 2005).

Figure 2. The frequency of moderate and strong earthquakes at the site

In geological profile, generally, the ridges and hills in the vicinity have the less permeable soft rocks and clayey soils in upper part, and older, permeable aquifers in lower part of profile. The aquifers are recharging at higher elevations, producing artesian or sub artesian groundwater pressures at the village site. Also, in the vicinity, the mineral water is commercially extracted and bottled.

Figure 1. The St. Ana church with graveyard.

551


The complex investigations at the location have been made, including borings and sampling, laboratory tests, water level measurements and monitoring of church wall movements. The thorough engineering geology investigations were also performed, and, in order to obtain reliable geotechnical model of landslide, the so-called RNK method was used. 2

RNK METHOD - FUNDAMENTAL NOTIONS AND BASIC DEFINITIONS

O

PEAK - P OR RESIDUAL FRICTION ANGLE - R ( )

The RNK method (RNK-the acronym in Croatian language) or the Reference Level of Correlation Method (Ortolan 1996) is a fully developed method for engineering-geological and geotechnical modelling. It is primarily intended for the landslide recognition and the analysis of the slope stability of soils and soft rock formations. However, the “sedimentation fingertip” obtained by geotechnical correlation column can be also used for reliable association of other test results in clayey sloppy profiles (Ivsic et al., 2005) The RNK (Reference Level of Correlation) is defined as an unequivocally recognizable and visually identifiable (or graphically defined!) bedding plane or any other reference plane within a structural feature, in relation to which an altitude of all studied profiles can be unambiguously defined, with individual point analysis of any material property. Such plane is a part of a single vertical lithostratigraphical i.e. engineering geological and/or geotechnical sequence (engineering-geological and/or geotechnical correlation column). The importance of correlation for the slip-surface and/or slip-zone determination is emphasized by Ortolan (1990). The plasticity index has proven to be a reliable strength indicator for cohesive materials (Ortolan 1996, Ortolan & Mihalinec 1998, Ortolan et al., 2009). The highest values of plasticity index, but also the liquid limit, correspond to the lowest expected values of friction angle. This fact allows a new approach to exact geotechnical modelling. Therefore, testing of Atterberg plasticity limits on point samples can be recommended for the identification of zones with lowest shear strengths. The sample length should not exceed 10cm (sometimes it should be aslittle as several centimetres, and even several millimetres). The sampling interval of 0.5–1.0m is usually considered sufficient. The correlation between the plasticity index and angle of internal friction is given in Figure 3, as developed by various authors, systemized by Ortolan & Mihalinec (1998) and enriched by new carefully obtained data. 40

35

Residual friction angle (Ortolan & Mihalinec, 1998) Peak friction angle (Ortolan & Mihalinec, 1998) Residual friction angle (1998-2006) Landslide Hospital Merkur in Zagreb ( 2005/2006) Landslide Jarpetar near Buje - Istra (2002) Landslide Česmički west in Zagreb (2002) Landslide Zalesina: Triassic clays and shales (Ortolan, 1996) Residual friction angle: Landslide Gorica Svetojanska

46,0

118

Allophane: JAVA

Halloasyte: JAVA 119

4

30

27 5

89 7 90

88

60 112

25

8

47

3 49

9 6

91 62 66

107 104

87

110 103 101

26

55 95

25 19

99

44

108

(29-32): Clay from Carboniferous Shales and Mudstones Cucaracha Shale: (15)

35 36 29 15

84 82 85

65

114

1 53 54

115

31

5

18

86 10 113

102

32

10

P

64 94

109

(25-27): Materials Containing Hydrous Mica

15

30 33 42 45

69 22

41 24 37

28

83

Soft Clays (47-49)

77

48

116 34 59

23

98

40

20 58

76 78

93 56 70

80

38 52

39

111 72 67 43 79 71

73 46

16 17

11

75

50 14

63

12

13

21

74

R

(20-24): Materials containing montmorillonite

0 0

10

20

30

40

50

60

70

80

90

DESCRIPTION OF THE LANDSLIDE AND GEOTECHNICAL PROPERTIES OF MATERIALS

The topographic presentation of the neighboring terrain in Gorica Svetojanska with the contour of the landslide is given in detailed engineering geology map of the area (Figure 4). Results of laboratory and in situ investigations, presented in form of geotechnical correlation column are presented in Figure 5. Plasticity chart with encircled critical geotechnical zone-2 is presented in Figure 6. Formations found on the landslide (calcitic clays and clayey marls) date back to the Pontian.

Halloasyte

51

97

106

3

117=120

96 68

105

(100-107):Triassic clays and shales 100

20

Very sensitive clays: 4-8 (OTAWA-KANADA: 8)

92 2

61

under study, geotechnical correlation column, and engineeringgeological map with slip-plane contour lines and with clearly delineated slip areas and hydro-isohypses or hydro-isopiestic lines at the slip-plane level (Ortolan 1996, 2000). The geotechnical correlation column is a consistent engineering-geological and/or geotechnical soil model (design cross section) in which adequate parameters (defined in laboratory or in situ either by point method or continuously) can reasonably be allocated to every defined layer (and portions of such layers) along the entire height of the vertical sequence of formations covered by the study. From such geotechnical correlation column we may in principle differentiate zones of minimum residual shear resistance, with their thicknesses and continuities, but also layers with different moisture content, hydraulic conductivity, natural compaction, compressibility, etc. The engineering-geological and/or geotechnical correlation column of an analyzed area is the "key" to the interpretation of overall engineering-geological and/or geotechnical relationships in a required number of profiles selected at will for 2D and spatial analysis, which is especially significant in 3D analysis of stability. The consistent use of the RNK-method in the period from 1995 to the present day has resulted in the elaboration of threedimensional geotechnical models for some fifty landslides. In all of these cases the following parameters were successfully defined: sliding body geometry, pore pressures and shear strength parameters for materials along zones of minimum shear resistance. In combination with existing topographical documents, this enabled accurate stability analyses and definition of optimum improvement procedures. The Podsused landslide may be described as one of the most complex urban landslide projects in the world (Ortolan 1996, 2000). It is precisely on this project that the RNK-method has been developed in full detail, and the reliability of the model was confirmed with photogrammetric measurements (Ortolan et al. 1995) as well as with three-dimensional stability analyses (Mihalinec & Stanić, 1991). Most of the studied landslides have been stabilized, in all cases with great success, and the supervisory work conducted during remedial works provided positive feedback information about the correctness of adopted engineering-geological and geotechnical landslide models, (e.g. at the Granice landslide; Jurak et al., 2004), and about reliability of the engineeringgeological and geotechnical correlation column (design cross section). On some projects the reliability of the model was checked and confirmed by appropriate inclinometer, piezometer and benchmark measurements.

100 110 120 130 PLASTICITY INDEX - PI (%)

Figure 3. Correlation between index of plasticity and angle of internal friction – both peak and residual.

The following supporting documents are most often used in the study of landslides: general geological map of the wider area

552

4 4.1

ANALYSES Wall movements

The investigation program included the measurements of relative rotation of church walls using several horizontal and vertical tilt meters, and, also the change of crack widths during monitoring period (originally found cracks were 15-20mm wide). The particular results are shown in Figure 7.


Figure 4. Detailed engineering geology map of the investigated area 11

9

WL (ALL BOREHOLES)

CH

SAMPLE OR SPT POSITION WITH RESPECT TO REFFERENCE LEVEL OF CORRELATION

(RNK ± m)

8 SPT (BOREHOLES NEAR THE CHURCH)

7 6

SPT (BOREHOLES AT THE GRAVEYARD)

5

3 2

1

1

RNK

0

60

50

40

30

CL

2

-1

GEOTECHNICAL ZONE - 2

CI

PI: SAMPLES FOR RING SHEAR TEST

4

PLASTICITY INDEX - PI (%)

PI (ALL BOREHOLES)

10

MH OH

20

-2 -3

3

-4

MI OI

-5

SAMPLES FROM BOREHOLES NEAR THE CHURCH

-6

4

-7

0

-9 -10 -11 -12

5

-13 -14

GEOTECHNICAL ZONE

-15 -16 0

10

20

30

40

50

60

70

80

90

100

LIQUID LIMIT - LL & PLASTICITY INDEX - PI (%); SPT (NUMBER OF BLOWS - TUBE)

Figure 5. Geotechnical correlation column of the landslide. Figure 6. Plasticity chart of materials from the landslide. The encircled zone contains samples from preferred slip-zone

553

10

SAMPLE FOR RING SHEAR

ML OL

SC

-8

10

SAMPLES FROM BOREHOLES AT THE GRAVEYARD

SF

20

30

40

50

60

70

80

0 90 100 LIQUID LIMIT - LL (%)

The unexpected “swaying” of eastern part of church was recorded (i.e. the movements of whole church block had alterative directions). This has been confirmed by independent records in horizontal and vertical tilt meters on neighbouring east and south wall, also accompanied with relative closing or opening of cracks. The ground water levels were not measured in the same frequency, but the collected data indicate possible correlation of seasonal variations of water levels with the directions of wall movements.


The whole situation at the site can be described as unstable (labile) or at the limits of equilibrium. However, even the recent observations have shown that the net effect of “swaying” are slow irreversible displacements in direction of sliding, with cumulative displacements of 4 - 8 mm in last several years. Also, some cracks have opened at the western part of church which was “quiet” during the intensive monitoring period.

conditions - by obtaining the safety factors near Fs = 1. Also shown is the expected trend of lowering of safety factor with rise of ground water level. These rough estimates are demonstrated for static conditions, implying that even the low or moderate seismic activity can significantly reduce slope stability. 5

CONCLUSION

The interaction of engineering geology and geotechnics in the process of designing geotechnical structures is very important. In the study of landslides or stability levels of natural and artificially shaped slopes, unequivocal results can be obtained by the correlation of formations. This can be done by introducing the reference level of correlation (RNK-method) and by looking for the zone of minimum shear strength in the engineering-geological and geotechnical correlation column. The creation of reliable geotechnical model is a center of this process, and it is crucial for the quality of the entire project. The correlation of the friction angle with the liquid limit or plasticity index is suggested for correct assessment of shear strength. 6

Figure 7. Monitoring of church wall displacements. 4.2

Stability analyses

The presented charts using the RNK method describe the landslide underground conditions and enable the construction of geotechnical models for engineering analyses in various crosssections.

Figure 8. Model and results of stability analyses

The cross-section A-A (shown in Figure 4) which includes the church ridge and downhill slope was used for common stability analyses (Spencer limit equilibrium method). The layers corresponding to the geotechnical zones in Figure 4, with several slip surfaces in the layers of lowest strength are shown in Figure 8. The strength parameters taken in analyses were: cohesion c’=0 kPa (for all layers), and friction angle ’ = 28, 24, 30º (for zones 1, 2-4, 5, respectively). The minimal friction angle ’ = 24º corresponds to the results of ring shear test and correlation chart. The ground water levels were varied few meters from referent level to estimate the influence of possible variations. This type of numerical modelling might be understood as too crude or too approximate for such a complex geologic situation at the site. However, the results (factors of safety) reveal that the established geotechnical model (with sequence of layers, friction angles, water levels) and slip surfaces respecting the established weakest zones, demonstrate the unstable

554

REFERENCES

GZ. 2005. The catalogue of earthquakes in Croatia, Department of geophysics, Faculty of Science, University of Zagreb Ivšić, T., Ocvirk, E., Pavlin, Ž. 2005. Geotechnical Aspects of Small Retention Dam Vir in Croatia. Proc. Ninth International Symposium on Water Management and Hydraulic Engineering. Nachtnebel H.P. (ed.). Ottenstein : BOKU-University of Natural Resources and Applied Life Sciences, 2005. 221-228 Jurak V., Ortolan Ž., Slovenec D. & Mihalinec Z. 2004. Verification of Engineering-Geological / Geotechnical Correlation Column and Reference Level of Correlation (RNK) Method by Observations in the Slip-Plane Zone. Geologia Croatica 57(2): 191-203 Mihalinec Z. & Stanić B. 1991. Three-dimensional slide analysis procedure (in Croatian). Građevinar, 42(9): 441-447 Ortolan Ž. 1990. Le rôle de la methode de correlation dans la determination des zones de parametres minimaux de resistance au cisaillement. Proc. of the Sixth Int. Congress IAEG, 6-10 Aug. 1990, Amsterdam. Balkema: 1675-1679 Ortolan Ž. 1996. The creation of a spatial engineering-geological model of deep multi-layered landslide on an example of the Podsused landslide in Zagreb (in Croatian). PhD Thesis. University of Zagreb Ortolan Ž. 2000. A Novel Aproach to the Modeling of Deep Complex Landslides with Several Sliding Planes. In E. Bromhead, N. Dixon, M.I. Ibsen (eds) Landslides in Research, Theory and Practice. 3: 1153-1158, Thomas Telford Ortolan Ž. & Mihalinec, Z. 1998. Plasticity index - Indicator of shear strength and a major axis of geotechnical modeling. In B. Marić et al. (eds) Geotechnical hazards, Proc. of the XI-th DanubeEuropean conference on soil mechanics and geotechnical engineering, Poreč, Croatia, 25-29 May 1998. Balkema: 743-750 Ortolan Ž., Mihalinec, Z., Stanić, B. & Pleško, J. 1995. Application of Repeated Photogrammetric Measurements at Shaping Geotechnical Models of Multi-layer Landslides. Proc. 6th Int. Symp. on Landslides. Balkema: 1685-1691 Ortolan Ž., Zlatović S. & Vrkljan I. 2009. Geotechnical 3D modeling in soft rocks using the RNK method. Rock engineering in difficult ground conditions – soft rocks and karst. Proc. of the reggional symp. of the ISRM, Eurock 2009, Dubrovnik, Croatia, 29-31 october 2009. CRC Press, Taylor & Francis Group: 489-494.

Une nouvelle sonde permettant de mesurer sans extrapoler la pression limite pressiométrique des sols A new probe for measuring the pressuremeter limit pressure of soils without extrapolation Jacquard C., Rispal M.

Fondasol, Avignon, France

Puech A., Geisler J., Durand F.

Fugro GeoConsulting, Nanterre, France

Cour F.

Calyf, Maisons Lafitte, France

Burlon S., Reiffsteck P.h.

IFSTTAR, Marne-la-Vallée, France RÉSUMÉ: Une limite actuelle des essais pressiométriques de type Ménard est liée à la difficulté d’atteindre des volumes d’expansion et des pressions importants sans risque systématique d’éclatement. Une nouvelle sonde a été développée qui permet d’atteindre, même sous pressions élevées, le doublement du volume du trou et donc la mesure directe de la pression limite conventionnelle du sol. On décrit les innovations technologiques qui ont conduit à accroitre les performances et la fiabilité des sondes. On présente ensuite des essais comparatifs sur différents sites montrant les apports techniques et opérationnels du nouveau concept. ABSTRACT: A present limitation of Menard type pressuremeter tests is due to the difficulty of reaching large expansion volumes and high pressures without exposing to significant risks of bursting. A new probe has been developed allowing the volume of the hole to be doubled, even under high pressures: the conventional limit pressure can then be directly measured. Technological innovations increasing the capabilities and reliability of pressuremeter probes are described. Comparative tests on different sites are presented demonstrating the technical and operational contribution of the new concept. MOTS-CLÉS : essai pressiométrique Menard, sonde, mesure, membrane KEYWORDS: Menard pressuremeter test, probe, measurement, membrane 1.

INTRODUCTION

profil limite à partir duquel elle oppose une résistance très élevée à toute dilatation complémentaire. Dans le cas de la sonde pressiométrique, ce profil limite, en forme de fuseau, correspond à un volume d'injection de 1100 cm3 dans la cellule centrale (Figure 1b). Un dispositif similaire de gaine textile de contention a été mis en application pour la membrane de la cellule centrale (résistance propre pm= 30 kPa). La cellule, munie de ce dispositif présente une section parfaitement cylindrique sur l'ensemble de sa plage d'injection, jusqu'à un volume de 1100cm3 (Figure 1a) tout en présentant une faible résistance propre.

La quasi-totalité des essais pressiométriques réalisés à partir de sondes de type Ménard sont arrêtés avant d’atteindre la pression limite du sol, définie comme la pression correspondant au doublement du volume initial du trou (normes NFP 94-110). Les tentatives pour atteindre cette pression limite avec les matériels couramment utilisés se soldent très fréquemment par l’éclatement de la sonde. Cette situation n’est évidemment pas satisfaisante et a conduit à rechercher des améliorations (Cour et al., 2005). L’article présente les caractéristiques et les performances de la sonde pressiométrique Francis Cour (en abrégé sonde FC) conçue de manière à atteindre quasi systématiquement le doublement du volume de la cavité sans éclatement et pour des niveaux de pression nettement supérieurs à ceux tolérés par les sondes standard. Des essais comparatifs menés dans différentes formations, avec des matériels standards et avec la sonde FC, illustrent les capacités de la sonde et ses performances opérationnelles.

1a

1

1

2. SONDE PRESSIOMÉTRIQUE FRANCIS COUR La sonde pressiométrique FC est, selon la norme NF P94110-1, une sonde du type G à gaine souple.La longueur de la cellule centrale est de 210mm ; celle des deux cellules de garde est de 105mm.. Les cellules ont un diamètre extérieur de 58 mm. La principale originalité de la sonde réside dans l'adjonction, autour de la gaine extérieure en élastomère, d'une gaine textile de contention qui a fait l’objet d’un brevet déposé en 2006 par Francis Cour. Cette gaine, de forme cylindrique au repos (Figure 1b), a la propriété de se dilater en opposant une très faible résistance, jusqu'à atteindre un

Figure 1. Vues de la membrane centrale gonflée à 1100 cm3 (1a), de la gaine de contention non gonflée (1b) et gonflée à 1100 cm3 (1c)

La nouvelle sonde pressiométrique est au final composée de la membrane de la cellule centrale et de la gaine décrites cidessus, recouvertes d'une sur-gaine en polyuréthane et de lamelles métalliques (Figure 2).

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Tableau 2 : Essais comparatifs dans l’argile des Flandres 2a

2b

Figure 2. Sur-gaine (2a) et lamelles métalliques (2b) constituant la sonde FC.

Les propriétés particulières de la sonde en termes de capacité de gonflement en volume et pression, et de robustesse, sont illustrées par le diagramme d'étalonnage de la sonde, à l'air libre, poussé à 6 MPa (Figure 3).

VL

[m]

[cm3]

6 7 8 9 10 11 12

668 686 782 742 656 620 720

SONDE FC

EM

Pl*

[MPa] [MPa] 11.5 12.5 15.3 16.5 15.6 14.0 19.7

0.96 1.02 1.15 1.24 1.32 1.33 1.45

VL

[cm3] 680 712 688 716 699 719 763

EM

Pl*

[MPa] [MPa] 8.1 9.4 13.1 10.3 14.3 13.0 11.2

0.99 1.01 1.28 1.29 1.62 1.67 1.68

On constate que : - entre 6 et 9 m les valeurs de pression limite sont en bon accord ; au-delà les valeurs obtenues par la sonde FC sont légèrement supérieures mais restent dans la limite des variations locales indiquées par les essais au CPT électrique réalisés à proximité immédiate (Figure 5); - les modules pressiométriques mesurés à la sonde FC sont un peu plus faibles (EM,FC / EM, SN ~ 0.75) ; - les rapports EM/Pl* sont en conséquence plus faibles avec la sonde FC (EM/Pl* ~ 8.5) qu’avec la sonde SN (EM/Pl* ~ 12.5)

Figure 3. Courbe d’étalonnage de la sonde FC, menée à 6 MPa.

3.

Profondeur

SONDE SN

ESSAIS COMPARATIFS

Des essais comparatifs ont été menés d’une part avec une sonde FC, d’autre part avec une sonde classiquement utilisée par la profession sur plusieurs sites. Le mode opératoire a été identique pour chaque sonde, et a respecté les critères de la norme NF P94-110-1. Le tableau 1 présente les caractéristiques d’étalonnage et de calibrage réalisés sur les quatre chantiers présentés par la suite. On note que la sonde FC vérifie bien les critères de la norme NF P94-110-1 d’une sonde à gaine souple.

1000

Essai

Volume corrigé [cm3]

800

Tableau 1. Caractéristiques de la sonde FC et du système CPVtubulures. SN: sonde nue ; TF: Tube fendu ; Vs, pel selon NF P 94-110-1 Merville Dunkerque Grand Paris Londres FC SN FC SN FC TF FC Vs [cm3] 522 484 595 475 553 510 483 pel[MN/m²] 0.27 0.17 0.41 0.34 0.32 0.39 0.35

VL=VS+2V1

VL=763 cm3

600

PL=1,79

400 EM=11.2 MPa

200

0

3.1 Argile des Flandres (Merville)

Etalonnage

P2=0,96

P1=0,39

0.0

0.2

0.4

0.6

0.8

1.0 1.2 1.4 Pression corrigée [MPa]

1.6

1.8

Figure 4 : Essai avec sonde FC dans l’argile des Flandres (-12m)

L’argile surconsolidée des Flandres est un sol bien connu, particulièrement homogène sur le site expérimental de Merville. On dispose de nombreux sondages pressiométriques et pénétrométriques (Puech et al., 2013). On a réalisé à titre comparatif deux forages espacés de 5m réalisés à la tarière simple Ø63mm à sec, tubés en tête et par passes de forage de 3m. Pour les deux sondes, les conditions opératoires sont identiques: i) pression différentielle (pr+ph)-pk voisine de 0.14MPa, ii) 30m de tubulures coaxiales Ø=3mm, iii) eau pure. Pour la mesure de grands volumes, la sonde FC a été couplée à un CPV piloté PREVO (Jean Lutz S.A.). Les résultats sont donnés dans le tableau 2. Les valeurs de Pl* pour la sonde nue (SN) ont été calculées par extrapolation conformément à la norme NF P94-110-1 car le doublement de la cavité (VL=VS+2V1) n’a pas été atteint (sauf essais à 7, 10 et 11m). Les valeurs en gras sont obtenues par interpolation linéaire sur les points mesurés. La valeur de Pl* est toujours mesurée directement avec la sonde FC comme illustré sur le Figure 4 pour l’essai à 12m.

556

2.0


Figure 6. Essai avec la sonde FC dans les alluvions (-3.5m).

La figure 6 montre l’essai à 3.5m de profondeur dans les alluvions avec la sonde FC. Une expansion à 700 cm3 a permis de mesurer directement la pression limite

Figure 5 : Résultats comparatifs des essais PMT et CPT à Merville

Sables de Dunkerque

Sur ce site correspondant à un chantier de production classique, la sonde FC a été utilisée dans les mêmes conditions de forage au bilame sous boue, et avec les mêmes critères d’arrêt que la sonde nue SN, soit trois points après le fluage. Une analyse statistique des valeurs mesurées a été menée : elle a concerné 11 sondages au total (5 SN et 6 FC) Les valeurs ont été regroupées selon 3 classes de sable (1a : sables moyennement compacts, 1b : sables compacts, 1c : sables très compacts). Le tableau 3 montre que pour une mise en œuvre identique les résultats avec les deux types de sonde sont équivalents pour un total de 191 essais analysés.

Figure 7. Sondages Grand Paris : pressions et volumes atteints pour chaque essai

Deux essais sur treize réalisés avec la sonde TF ont dû être arrêtés prématurément par éclatement de la membrane de sorte que les volumes injectés ont dû être limités à 440 cm3 (Figure 7). Avec la sonde FC, aucun éclatement n’est à déplorer pour des essais menés jusqu’à 800 cm3 (limite correspondant à la capacité du CPV utilisé), ou arrêtés à 8 MPa, critère d’arrêt fixé au cahier des charges (Tableau 4). Tableau 4. Valeurs pressiométriques comparées (sondages Grand Paris)-

Tableau 3. Valeurs pressiométriques comparées (site de Dunkerque) EM et Pl* en MPa.

10.2

1.5

23

11.4

1.5

0.90

0.97

1b

25

23.8

3.3

32

21.5

2.9

1.11

1.17

1c

45

36.3

4.8

28

37.5

4.7

0.97

1.02

3.3

Région parisienne

Deux sondages pressiométriques ont été réalisés à 5m de distance sur un site de la région parisienne dans le cadre du chantier “Grand Paris”. La sonde classique est de type G avec tube fendu (TF) et cellule courte de 44 mm. La coupe lithologique au droit des sondages est : - 0-7m : alluvions anciennes : limons, sables et graves - 7 -10m : calcaires de St-Ouen - 10-20m : sables de Beauchamp.

557

pl*[MPa]

38

EMTF[MPa]

1a

Vl [cm3]

pl*FC /pl*SN

plFC3*[MPa]

EMFC /EMSN

plFC*[MPa]

Ratios FC/SN

EMFC [MPa]

Sonde standard SN 83 Pl* essais EM

Vl [cm3]

Sonde FC 108 essais EM Pl*

Sonde TF

Sonde FC Profondeur

3.2

3.5

697

30

5.5

4.3

750

74

4.5

5.0

667

41

6.3

4.8

970

36

5.9

6.5

931

61

10.7

10.0

1030

36

>3.5

7.9

839

30

2.8

> 1.8

970

19

>1.7

55

5.8

11.0

629

102

8.4

8.0

730

12.5

889

105 >7.8

>5.3

710 128 >7.1

14.0

893

47

5.3

5.1

710

15.5

833

13

2.1

2.0

730 123 >4.1

17.0

-

39

6.1

5.6

610

69

>4.2

18.5

-

152 >7.8 >7.8

710

91

6.8

20.0

883

66

610

61

3.7

5.8

4.8

44

4.2


4. CONCLUSION ET PERSPECTIVES

On constate pour les deux sondes, une concordance globale entre modules pressiométriques (EM,FC/EM,TF ~ 1.1). En revanche le ratio pl*FC/pl*TF est de l’ordre de 1.3. Si on reprend l’interprétation des essais FC (colonne pl*FC3) en limitant l’extrapolation à trois points après P2 (ce qu’autorise la norme, et ce qui correspond à la façon dont sont traités la plupart des essais en France), on obtient des valeurs pl*FC3 systématiquement inférieures d’environ 10% aux valeurs pl*FC. L’extrapolation limitée à 3 points conduit ainsi à une sous-estimation des pressions limites. Corrélativement les valeurs de EM/Pl* se trouvent surestimées. 3.4

La sonde FC constitue une avancée technologique significative dans la mesure des paramètres pressiométriques des sols. La conception originale de sa membrane autorise dans la grande majorité des cas la mesure directe de la pression limite par doublement effectif du volume de la cavité. Sa remarquable résistance à l’éclatement y compris sous pression élevée permet son emploi dans des terrains hétérogènes et résistants, correspondant au domaine d’utilisation habituel du tube fendu, et autorise ainsi sur les chantiers des cadences accrues. La sonde est conforme aux exigences de la norme NFP 94-110-1 pour une sonde à gaine souple. Dans des conditions opérationnelles similaires, la sonde FC et une sonde nue standard (SN) fournissent des paramètres pressiométriques équivalents. Les exemples traités montrent que l’extrapolation de données obtenues avec des sondes standard à trop faible capacité d’expansion peut conduire à une sous-estimation des pressions limites. La généralisation de ce type de sonde passe par une amélioration des matériels existants notamment en ce qui concerne la capacité en volume et pression des contrôleurs pression – volume.

Essais haute pression

Des essais haute pression (12 MPa) avec la sonde FC ont été mis en œuvre entre 54 et 66m de profondeur, dans des sables fins très compacts de Londres (Thanétien). Les deux forages ont été réalisés au taillant en rotation 66 mm avec injection de boue et mise en place de tubage à l’avancement. Des essais haute pression avaient déjà été réalisés dans ces sables avec une sonde nue standard et du matériel adapté pour la circonstance (Massonnet, 2005). L’arrêt des essais à 12 MPa est dû aux limites du contrôleur pression-volume et non pas à la sonde qui a permis de mesurer des pressions limites élevées, sans aucun éclatement pour les 18 essais. La figure 8 montre un essai à 63m de profondeur avec 4 points au-delà de P2, ce qui autorise une extrapolation réaliste de la pression limite (Pl*= 13 MPa; EM= 112 MPa, pour un volume brut Vmax= 450 cm3 ; Vmax net= 409 cm3).

5. REMERCIEMENTS Les auteurs remercient la société Jean Lutz S.A. pour la mise à disposition gracieuse de son matériel et son assistance sur les chantiers expérimentaux de Fugro et de Fondasol. 6.

REFERENCES

AFNOR. 2000. Norme NF P94-110-1. Essai pressiométrique Ménard. Cour F., Puech A., Durand F. 2005. Un pressiomètre de nouvelle génération. 2005. Proc. ISP5-PRESSIO (1), 63-73 Massonnet R. 2005. Le pressiomètre sous haute pression. Proc. ISP5-PRESSIO (1), 81-90 Puech A. et Benzaria O. 2013. Effet du mode de mise en place sur le comportement statique de pieux dans l’argile fortement surconsolidée des Flandres. Proc. 18 ICSMGE, Paris

Figure 8. Essai avec la sonde FC dans les sables du Thanétien (-63m).

558

Long-term Deformation of the Reclaimed Pleistocene Foundation of the Offshore Twin Airport Déformations à long terme d’une fondation de remblai pléistocène récupéré sur mer pour un projet d’aéroport jumelé Jeon B.G.

Samsung C&T Corporation

Mimura M. Graduate School of Engineering, Kyoto University ABSTRACT: A series of elasto-viscoplastic finite element analyses is performed to assess the long-term deformation including the interactive behavior of the reclaimed Pleistocene foundation due to the adjacent construction of the offshore twin airport. Attention is paid to the modeling of permeability for the Pleistocene sand gravel layers considering the sedimentation environment. The concept of “mass permeability” is introduced to model the actual process of dissipation of excess pore water pressure in the field. It is regarded as the macroscopic capability of permeability for the individual Pleistocene sand gravel layers by evaluating the permeability not of each element but of the whole layer in one body. The mechanism for the propagation of excess pore water pressure due to construction of the adjacent reclamation is discussed through the numerical procedure using the concepts of “mass permeability”. The concept of “mass permeability” for the individual Pleistocene sand gravel layers is found to well function to assess the long-term deformation including the interactive behavior in the reclaimed Pleistocene foundation. RÉSUMÉ : Les déformations à long terme d’un remblai pléistocène en mer sont évaluées a partir d’une série d’analyses élastoviscoplastiques par éléments finis. Les interactions dues aux travaux d’aménagement d’aéroport jumelé sont aussi prises en compte. On vise plus particulièrement à modéliser la perméabilité du sable/gravier pléistocène en considérant la sédimention du milieu. La dissipation des surpressions interstitielles in-situ est calculée à partir d’une perméabilité massique de l’ensemble des couches sable/gravier. Les mécanismes de propagation de surpressions interstitielles induites par le remblai voisin sont déterminés par modélisation numérique faisant appel au concept de perméabilité massique. L’application de ce concept semble être commode pour évaluer les déformations à long terme des couches sable/gravier pléistocène en interaction avec d’autres ouvrages voisins. KEYWORDS: elasto-viscoplastic finite element analysis, mass permeability, standard hydraulic gradient 1

INTRODUCTION

The development of coastal areas accomplished in Japan has been outstanding. Kansai International Airport (KIX) was constructed in Osaka Bay as two man-made reclaimed islands to minimize noise and pollution in residential areas as well as to meet the increasing demand for air transportation. Such a largescale offshore reclamation in Osaka Bay is accompanied with large and rapid settlement of deep Pleistocene clay deposits (Mimura et al., 2003). Long-term settlement of the Pleistocene marine foundations due to huge reclamation load has been of great concern in this project. The seabed deposits of Osaka Bay have been formed due to the soil supply from the rivers and the alternating deposits of KIX have been formed due to sedimentation of clayey soils during transgression and of sandy to gravelly soils during regression on the sinking base of Osaka Bay. The Pleistocene clay deposited in Osaka Bay exhibits the behavior of the quasi-overconsolidated clay without definite mechanical overconsolidation history. Itoh et al. (2001) summarized on the basis of the data from elastic wave exploration and in-situ boring logs that the Pleistocene sand gravel deposits are not always distributed uniformly in thickness, consistently and that the amount of fine contents included in them is significant. The most serious problem originating from these sand gravel deposits is the “permeability” that controls the rate of consolidation of sandwiched Pleistocene clays. In the sense, the modeling for the quasi-overconsolidated Pleistocene clay and the evaluation of permeability for the Pleistocene sand gravel deposits are the significant factors to assess the long-term behavior of the reclaimed Pleistocene foundation due to the reclamation of the offshore twin airport. Mimura and Jang (2004) proposed a concept of compression in which viscoplastic behavior is assumed to occur even in the quasi-overconsolidated region less than pc for the Pleistocene

559

clays in Osaka Bay. The procedure has been found to be versatile and allows for the long-term settlement monitored in the reclaimed islands in Osaka Port to be described (Mimura and Jang, 2005a). In the present paper, the numerical procedure to assess the long-term behavior of the Pleistocene deposits at KIX in terms of elasto-viscoplasitc FEM is proposed by introducing the concept of “mass permeability” and “standard hydraulic gradient” for the Pleistocene sand gravel layers. The validity of the procedure is carefully discussed by comparing the performed results with in-situ measurements. 2 CONCEPTS OF “MASS PERMEABILITY” AND “STANDARD HYDRAULIC GRADIENT” Mimura and Jang (2005a) reported when the permeability of sand gravel layers is considered perfectly drained, onedimensional analysis only considering the characteristic of clayey soil can be adopted for the consolidation problem without considering the effect of permeability loss in the those sand gravel layers. However, the sand gravel layers sandwiched by the Pleistocene clay layers at KIX were recognized not to function as perfect drainage layers through the in-situ measurement of excess pore water pressure. Therefore, the two or three-dimensional analysis that considers the permeability of the Pleistocene sand gravel layers is required to assess the longterm behavior of the reclaimed Pleistocene foundation. The influential factors to evaluate the permeability of sand gravel layers are the thickness, the horizontal continuity and the fine contents of them. The permeability of them is different with places even if they are categorized as the identical ones. But, it is impossible to evaluate the permeability of sand gravel layers at every point. It is also very difficult to confirm how the sand gravel layers under the Pleistocene marine foundation are


2nd phase island Ma13 Ma12

Ma11U Ma11L Ma10 Ma9 Ma7

Doc5&Ma8

Ds6 Ds7

Ds8 Ds10

17,880m

Monitoring point 1

A' (Offshore side)

1440m 1440m

2nd phase island

Ds1

Ds2

Ds3

Ds4

Ma11U

Ds5

Ds6

Ma10

Ds7

Dtc Ma12

Ds9

Ma9

5,000m

A

1st phase island

20.2m

10,000m

(Onshore side)

(Offshore side)

A'

Ma13

588.7m

The differential settlement of the individual Pleistocene clay layers as well as the excess pore water pressure at various depths, both in the clay and the sand gravel layers, have been measured at a lot of points of KIX. Figure 1 shows the plan view of KIX together with the location of representative monitoring points on the 1st phase island. A series of elasoviscoplastic finite element analyses is carried out along the representative section shown by A-A’ at monitoring point 1 in Fig.1. Figure.2 shows the representative foundation model assumed to be horizontally even layer that have a constant thickness and continuous layer based on the boring data at the monitoring point 1. Figure.3 shows the geologically genuine foundation model having the inclined base and layers that is constructed based on the soil exploration and geological survey data (Kitada et al, 2011). The clay layers increase in thickness towards the offing and the sand gravel layers drastically change in thickness horizontally. The continuity of the individual layers is still guaranteed even for the geologically genuine foundation

(Representative section)

Ds4

Ds5

Fi gure 2. Representative foundation model of KIX for finite element analysis at representative section

3 FOUNDATION MODEL AND HYDRAULIC BOUNDARY

N

A

Ds2

Ds3

Ds9

Ma6

Ma11L

Monitoring point 1 (S1)

1st phase island

Ds1

Dtc

5,000m

1440m 1440m

(Onshore side)

10,000m

17.0m

A'

131.0m

(Offshore side)

model in the present study. Here, Ma and Ds denote marine clay

148.0m

distributed in practice. The concept of “mass permeability” is proposed to evaluate the permeability not of each element but of the whole layer in one body. It is regarded as the macroscopic capability of permeability for the individual sand gravel layers by considering the horizontal continuity, the change in thickness and the degree of fine contents of them. Mimura and Jeon (2011) evaluated the mass permeability of the Pleistocene sand gravel layers using the simple foundation model as shown in Fig.2. The distribution of sand gravel layers not only in the loading area but also in the area that can rule out the effect of the hydraulic boundary condition should be considered to assess the mechanism of the propagation/dissipation of excess pore water pressure in the coupled stress-flow analysis. In the sense, on the basis of the assumption that the hydraulic gradient derived in the representative foundation model having the horizontally even layer with constant thickness is regarded as the standard one for the individual Pleistocene sand gravel layers, the evaluated mass permeability can be the representative of the capacity of permeability for the individual Pleistocene sand gravel layers at KIX. The standard hydraulic gradient is hence applied to the geologically genuine foundation model that has been developed to consider the actual stress level not only of the monitoring point but also of the considered area for the numerical analysis. Due attention should be paid to the fact that this assumption is only considered in horizontal position for the individual Pleistocene sand gravel layers.

Ds8 Ds10 Monitoring point 1

Doc5&Ma8

Ma6 Ma7

17,880m

Figure 3. Geologically genuine foundation model of KIX for finite element analysis at representative section

and Pleistocene sand gravel layer respectively. Ma13 is the Holocene marine clay whereas others are the Pleistocene origin. For the Holocene clay deposit, Ma13, sand drains are driven in a rectangular configuration with a pitch of 2.0 to 2.5 meters to promote consolidation. The lateral boundary of the clay layers is assumed to be undrained while the one of the sand gravel layers is assumed to be fully drained. Mimura and Jang (2005b) reported that when the distance to the boundary is set to be about 10 times of the loading area, the effect of the hydraulic boundary condition can be ruled out. Based on the findings, the same condition is satisfied even for the foundation models used in the present study. The distance to the offshore and onshore boundary is set to be 10,000m and 5,000m respectively. The present two foundation models are divided into finite element mesh consisting of 8,580 nodal points and 8,378 elements.

2 nd phase island

4

1st phase island S2

LOADING CONDITION AND SOIL PARAMETERS

The prescribed final overburden due to airport fill construction amounts to about 430kPa at the 1st phase island and about 530kPa at the 2nd phase island respectively. The 2nd reclamation is started after about 13years from the 1st reclamation. In the present analysis, the permeable capability evaluated from the concept of “mass permeability” for the Pleistocene sand gravel layers is applied for the present finite element analysis. On the basis of the findings by Itoh et al. (2001), the relatively high permeable capability are assumed for Ds1,3 10 because they have been evaluated as gravelly, horizontally continuous and

S3

A (Onshore side)

S Figure 1. Plan view of Kansai International Airport and the location of monitoring points on the 1st phase island

560


having enough thickness. On the other hand, very low permeable capability is assumed for Ds6 and 7 that have been evaluated to have insufficient thickness with high degree of fine contents and poorly continuous. The other layers have been evaluated as the ordinary permeable capability. The used all soil parameters for analysis are also exactly the same with that used by Mimura and Jeon (2011). 5

RESULTS AND DISCUSSIONS

The calculated distribution of excess pore water pressure before and after the construction of the 2nd phase island is shown in Fig.4 for two foundation models respectively. As shown in Fig.4, the similar distribution tendency of excess pore water pressure can be seen for two foundation models. It should be noted that a large amount of excess pore water pressure still remains undissipated in the middle Pleistocene clay layers, Ma10, 9 and Doc5&Ma8 as well as sand gravel layers, Ds6 and 7 before the construction of the 2nd phase island because of poor permeability of sand gravel layers, Ds6 and 7. In contrast, the excess pore water pressure in the upper and lower Pleistocene layers such as Dtc, Ma12,11,7,6 and Ds1,3,9,10 is monotonically dissipated with time because of high permeability of sand gravel layer, Ds1,3 and 10. At the completion of the 2nd reclamation, a large amount of excess pore water pressure is concentrated in the upper and middle Pleistocene layers such as Ma12, 10, 9 and Doc5&Ma8 beneath Before construction of 2nd phase reclamation (after 13years from construction)

Elevation(m)

-40

Ma12

-60

Ma11L

-80

Ma7

-140 -160 -166

Ma11U

Ds6 Ds7

Doc5&Ma8

0

2000

Ds8 Ds9 6000

8000

10000

Distance(m)nd

-60

Ma11L

-80

Ma7

-140

2000

16000

17880

4000

Ds4

Middle

Ds7

Ds8 Ds9

Lower

8000

10000

Distance(m)

12000

14000

16000

P.W.P

Ds3

500 400 300 200 100

P.W.P

Ds6

17880

6000

9000

12000

Ds8 Ds10

4000

6000

8000

10000

Ma13

Dtc Ma12 Ma11L

-230

Ma11U

Ma9

-380 -430

Ma7

-480

Ds5

Ma10

-280 -330

Ds1 Ds3

Ds7 Doc5&Ma8

Ds9

12000

14000

16000

17880

st

1 phase -6 -26

Upper Middle Lower

Ds2 Ds4 Ds6 Ds8 Ds10

Ma6

0

Ds10

15000

Horizontal distance (m)

2000

4000

6000

8000

10000

Distance(m)

12000

14000

16000

17880

Figure 4 (b). Contour of excess pore water pressure for geologically genuine foundation model at before and completion of 2nd phase reclamation

P.W.P

3000

2000

-180

-580 -619

0

Geologically genuine foundation Representative foundation

Ds6

Ds7

Distance(m)nd 2 phase Completion of 2 nd phase reclamation (after 19years from construction)

Excess pore water pressure(kPa)


500 400 300 200 100 0

0

0

Center of 1st phase island

Center of 2nd phase island

Ds5

Ds9

-6 -26

Ds4

Ma6

Ma7

-480

Ds3

(Unit: kPa)

Ds2

-530

Ds10

6000

Doc5&Ma8

-380 -430

Ds1

Ma10

Ma9

-330

-80

Figure 4(a). Contour of excess pore water pressure for representative foundation model at before and completion of 2nd phase reclamation

500 400 300 200 100 0

-280

-130

Ds6

Doc5&Ma8

Ma11U

Ma11L

-230

-580 -619

Upper

Ds5

Ma6 0

-180

Ds2

Ds3

Ma11U

Ma9

-120

Ma12

2 phase 1st phase

Ds1

Dtc

Ma10

-100

14000

-130

-30

Ma13 Ma12

12000

Dtc

1 phase

-530

Ds10

Completion of 2nd phase reclamation (after 19years from construction)

-40

-160 -166

4000

Ma13

-80

Ds4

Ds5

Ma6

-18

Elevation(m)

Ds3

Ma9

-120

Ds2

st

Before construction of 2 nd phase reclamation (after 13years from construction)

-30

Ds1

Dtc

Ma10

-100

(Unit: kPa)

Elevation(m)

Ma13

Elevation(m)

-18

1st phase

the foundation of the 2nd phase island. Here, a due attention should also be paid to the fact that the increased excess pore water pressure beneath the foundation of the 2nd phase island is propagated to that of the 1st phase island. Since the permeability of the upper and lower Pleistocene sand gravel layers is higher than the one of the middle layers, a larger amount of excess pore water pressure in the upper and lower Pleistocene layers is propagated compared to the one in the middle layers of the foundation of the 1st phase island. The calculated horizontal distribution of excess pore water pressure in the representative Pleistocene sand gravel layers (Ds3, 6, 10) are shown in Fig. 5 at the time before and after the construction of the 2nd phase reclamation for both foundation models. In the present study, the identical permeable capability for the individual Pleistocene sand gravel layers in two foundation models is applied by considering the concepts of “mass permeability” and “standard hydraulic gradient”. However, in Fig.5, it should be noted that the distribution of excess pore water pressure near the 1st phase island almost shows a good match for two foundation modes by applying the concept “standard hydraulic gradient” whereas the one of the other region shows the discrepancy distribution with the stress level. The stress level beneath the foundation of the 1st phase island is almost the same for two foundation models because the representative model was developed based on the monitoring point 1 whereas the one beneath the foundation of the 2nd phase island is different each other due to change in thickness of geologically genuine foundation model. It is noteworthy that

18000

(a) Before the 2nd phase reclamation

Center of 1st phase island

Center of 2nd phase island

500 400 300 200 100 0

P.W.P

Ds3

500 400 300 200 100 0

P.W.P

Ds6

500 400 300 200 100 00

Ds10

Geologically genuine foundation Representative foundation

P.W.P

3000

6000

9000

12000

15000

Horizontal distance (m)

18000

(b) Completion the 2nd phase reclamation

Figure 5. Horizontal distribution of excess pore water pressure for the representative Pleistocene sand gravel layers (Ds3, 6, 10) in a horizontal position

561


400 350 300 250 200 150 100 50 0 400 350 300 250 200 150 100 50 0 400 350 300 250 200 150 100 50 0

points S2 or S3 shown in Fig. 2.

Ds3

Start of 2 nd Completion of 2nd phase island phase island

Ds6

Settlement(m)


although the identical permeable capability for the individual

Geologically genuine foundation Representative foundation Measured Ds10

0

10

20

30

40

Elapsed time(years)

50

60

Start of 2 nd Completion of 2nd phase island phase island

Ma12

Ma10

Ma6

0

Geologically genuine foundation Representative foundation Measured 10

20

30

40

Elapsed time(years)

50

60

Figure 7. Comparison of measured and calculated settlement with time for the representative Pleistocene clay layers

Figure 6. Comparison of measured and calculated excess pore water pressure with time for the representative Pleistocene sand gravel layers

Pleistocene sand gravel layers was applied, the calculated results of excess pore water pressure could show the difference with the stress level. The calculated excess pore water pressure – time relations for two foundation models are shown in Fig. 6 together with the measured results for the representative Pleistocene sand gravel layers at the monitoring point 1. It is noteworthy that the excess pore water pressure in the upper (Ds3) and lower (Ds10) Pleistocene sand gravel layers is increased but the one of the middle layer (Ds6) is not increased due to the construction of the 2nd phase island. The long-term settlement associated with the phenomenon of propagation of excess pore water pressure is another serious problem for KIX. When the excess pore water pressure increases or the dissipation of excess pore water pressure is hindered due to the construction of the 2nd phase island, the settlement is also retarded or slight upheaval can happen (see Fig.7). It is also found that the calculated performance at the monitoring point 1 shows a good match for two foundation models by applying the concept of “standard hydraulic gradient” and can also well describe the whole process of deformation. 6

0 20 40 60 80 100 120 140 160 180 200 0 20 40 60 80 100 120 140 0 20 40 60 80 100 120 140 160

CONCLUSIONS

The long-term deformation of the reclaimed Pleistocene foundation of the offshore twin airport was numerically evaluated through the elasto-viscoplastic finite element analyses considering the concepts of “mass permeability” and “standard hydraulic gradient” for the Pleistocene sand gravel layers. The concept of “mass permeability” was evaluated as the representative permeable capacity of sand gravel layers of KIX. The representative permeable capacity of sand gravel layers was applied to the geologically genuine foundation model by introducing the concept of “standard hydraulic gradient” for the coupled stress-flow analysis. The concept of mass permeability for the sand gravel layers was found to well function to assess the process of excess pore water pressure generation/ dissipation/propagation and long-term settlement in the reclaimed foundations of KIX. The concept of standard hydraulic gradient was also found to well reproduce the representative permeable capacity by comparing the calculated results for two foundation models. The validity and objectivity of the proposed concepts will be investigated by applying them to the additional review sections including the monitoring

562

7

REFERENCES

Itoh, Y., Takemura, K., Kawabata, D., Tanaka, Y. and Nakaseko, K. 2001. Quaternary Tectonic Warping and Strata Formation in the Southern Osaka Basin Inferred from Reflection Seismic Interpretation and Borehole Sequences, Journal of Asian Earth Science, 20, 45-58. Kitada, N., Inoue, N., Takemura, K., Fukada, K. and Emura, T. 2011. Subsurface Structure Model Around Kansai Airport According to Re- Interpretation of Borehole Data based on Result of KIX18-1 Core. International Symposium on Advances in Ground Technology and Geo-Information, IS-AGTG, 137-142. Mimura, M. and Jang, W.Y. 2004. Description of time-dependent behavior of quasi-overconsolidated Osaka Pleistocene clays using elasto-viscoplastic finite element analyses, Soils and Foundations, 44(4), 41-52. Mimura, M. and Jang, W.Y. 2005a. Verification of the Elastoviscoplastic Approach Assessing the Long-term Deformation of the Quasi-overconsolidated Pleistocene Clay Deposits, Soils and Foundations, 45(1), 37-49. Mimura, M. and Jang, W.Y. 2005b. Long-term Settlement of the Pleistocene Deposits due to Construction of KIA, Proceedings of the Symposium on Geotechnical Aspects of Kansai International Airport, , 77-85. Mimura, M. and Jeon, B.G. 2011. Numerical Assessment for the Behavior of the Pleistocene Marine Foundations Due to Construction of the 1st Phase Island of Kansai International Airport, Soils and Foundations, 51(6), 1115-1128. Mimura, M., Takeda, K., Yamamoto, K., Fujiwara, T. and Jang, W.Y. 2003. Long-term settlement of the reclaimed quasioverconsolidated Pleistocene clay deposits in Osaka Bay, Soils and Foundations 43(6), 141-153.

Assessment of Scour Potential of a Circular Pier in Silty Sand Using ISEEP Caractérisation par ISEEP du potentiel d'érosion d'une pile circulaire dans un sable silteux Kayser M., Gabr M.

Department of Civil, Construction and Environmental Engineering North Carolina State University, Box 7908, Stinson Drive, Raleigh, NC 27695-7908; PH. (919)515-7904; ABSTRACT: Work in this paper describes an approach for the assessment of soil scour potential through the use of an In Situ Erosion Evaluation Probe (ISEEP) that is advanced by water jetting. Soil erosion parameters are assessed for silty sand in terms of critical stream power, and therefore, critical shear stress, and detachment rate coefficient. Scour depth around a circular bridge pier was computed using ISEEP data and compared with an empirical approach available in literature for estimating scour depth in soil similar to the tested in the study. The application of the idea and the utility of this technique to assess scourability profile are presented and discussed. RÉSUMÉ: Le travail présenté dans cet article décrit une approche pour l’évaluation du risque d'affouillement d’un sol in situ en utilisant une sonde d’érosion équipée de jet d’eau. Les paramètres d’érosion sont évalués pour les sables limoneux en fonction d’une puissance critique et par conséquent en termes d’une contrainte de cisaillement critique et d’un coefficient exprimant le taux de détachement. La profondeur d’affouillement autour d’une pile de pont a été calculée en utilisant les données issues de la sonde. Elle a été comparée à celle issue de l’approche empirique pour un sol similaire au sol étudié. L’applicabilité de l’approche proposée et son utilité pour l’évaluation du profil d’érodabilité sont présentées et discutées. KEYWORDS: Bridges, Erosion, Foundation, In Situ, Pier, Probe, Scour, Shear, Soil 1

INTRODUCTION

The assessment of scour and erosion rates of soil profiles supporting hydraulic structures and critical bridges is vital for ensuring safe performance under normal flow conditions, as well as the integrity of their foundation systems during and after severe storms. Richardson and Davis (2001) highlighted the importance of assessment of local scour around bridge piers as it is one of the most common causes of bridge failure. Several approaches ranging from simple steel sounding rods to remote sensing have been developed to assess scour depth after it has occurred. As presented by Lu et al. (2008) the more sophisticated approaches, including acoustic doppler and ground penetrations radars, have a high cost and require frequent maintenance and repair. Even then, these approaches do not provide an estimate of scour under future storm events. Current techniques for providing such information require either the removal of soil samples for laboratory testing, in a device such as the Erosion Function Apparatus (EFA) by Briaud et al. (2001), or limiting the measurements to erodibility of the surface sediments. Gabr et al. (2012) presented a prototype device, termed ISEEP (In Situ Erosion Evaluation Probe), for assessment of scour parameters with depth. ISEEP has been constructed as simple stainless steel tubes fitted with truncated cone tip. The cone-tipped vertical probe is attached to a digitally controlled centrifugal pump that provides controllable and repeatable water velocity at the tip, with sustained flow rate against any induced back pressure. As the water jet is induced through the cone tip, it mobilizes the soil particles. The test data are analyzed using the stream power (bed shear stress multiplied by the flow velocity) concept proposed by Annandale (2006) to account for the nature of the flow conditions induced during testing. The results from the tests are reduced to provide critical shear stress (c) and a rate of scour per unit

shear stress (kd). These two values are used in conjunction with the applied shear stresses (applied) per a given flow type and as appropriate to the structure being analyzed, to compute the scour rate (E) using the excess shear model as follows (Annandale 2006): E = k d (  applied -  c )

(1)

In this study, experimental work and analyses are conducted, using ISEEP-estimated data, for evaluating erosion parameters for a soil with 15% clay and 85% sand. The soil is classified as silty sand according to the Unified Soil Classification System. Tests are performed with different jet velocities and critical stream power value (Pc) and the corresponding c, and kd are evaluated using the data reduction scheme proposed by Gabr et al. (2012). An example showing the computation of the scour depth around a bridge pier using ISEEP-estimated data is presented. The results are compared with values obtained using empirical equations reported in literature, and the estimated scour depth using both approaches is presented and discussed. 2

BRIDGE PIER SCOUR

The magnitude and geometry of local scour at bridge piers in soil profiles with percent fines content have been documented in literature (e.g. Hosny 1995, Molinas and Hosny 1999, Briaud et al. 1999, 2001, and 2004). Hosny (1995), and Molinas and Hosny (1999) proposed empirical equations to assess scour depth for saturated and unsaturated compacted soil with a percent fines that lends a degree of cohesion to the soil. They reported that the scour depth decreased as compaction density was increased for the unsaturated “cohesive” soil conditions, and scour depth

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decreased with the decrease of the initial water content for the saturated “cohesive” soil. Briaud et al. (1999, 2001 and 2004) presented a method termed SRICOS for predicting scour in “cohesive” soils, with such an approach being the most comprehensive to date in literature. Scouring experiments around cylinders using “clay”– sand mixtures were carried out by Hosny (1995), Ansari et al. (2002), and Debnath and Chaudhuri (2010a, 2010b), among others, with fines fraction in the range of 0.05–0.4, 0.1–0.6, 0.2–1.0 and 0.05–0.35, respectively. Hosny (1995), and Debnath and Chaudhuri (2010a, 2010b) concluded that maximum scour depth decreases with the increase of “fines” content whereas Ansari et al. (2002) indicated that the

maximum equilibrium scour depth in sediments with fine contents could be higher than that of non-cohesive sediments under similar experimental conditions. Perhaps one reason for the difference in conclusions is attributed to the nature of fine being used in the study. In Ansari et al.’s (2002) study, the soil is reported as having zero Plasticity Index (PI). Table 1 shows several empirical equations to estimate scour depth, with the corresponding fines fraction and Froude number range for their applicability. The equations proposed by Hosny (1995), and Debnath and Chaudhuri (2010a, 2010b) are only applicable for a rather narrow Froude number (i.e. 0.13 – 0.33) range, in comparison to Ansari et al.’s (2002) range.

Table 1. Empirical Equations for Estimating Scour Around Bridge Piers for Soils with Fine Contents Reference

Equation

Conditions

Hosny (1995)

ds/b =18.9(Fr/(1+C))2

C ≤ 0.4 and 0.18 ≤ Fr ≤ 0.33

Ansari et al. (2002)

dsmc/dsms = 1.51(C*/φ*)0.2

Debnath and Chaudhuri (2010b)

3

ds/b = 8.2Fr

0.79

-0.28

C

PI = 0 and 0.16 ≤ Fr ≤ 0.69

0.15

(IWC)

2 -0.38

(τs/ρV )

0.13 ≤ Fr ≤ 0.20, W.C. ≤ 0.4, C ≤ 0.4 and 0.78 ≤

≤ 1.65

Comment b = pier diameter, Fr = Froude number = V/(gd)0.5, V = approach flow velocity, g = gravitational acceleration, d = depth of flow and C = clay fraction. dsmc = maximum scour depth for cohesive sediments, dsms = maximum scour depth for cohesionless sediments (estimated using equation proposed by Kothyari et al. (1992)), C* = [%Pc. Cu] / [(γs-γw).da], φ* = [%Pc . tan φc + (1 – %Pc) * tan φs] / tan φs, %Pc = percentage of clay content, Cu = undrained shear strength of soil, γs = unit weight of soil, γw = unit weight of water, da = arithmetic mean diameter, φc = angle of internal friction for clay and φs = angle of internal friction for sand. C = clay fraction, IWC = initial water content, τs = bed shear strength, ρ = density of water, = V/Vcs, Vcs is critical threshold velocity for sand and V = approach flow velocity.

EXPERIMENTAL PROGRAM

Testing was conducted in a circular chamber with a diameter of 1.0 m (3.3 ft) and a depth of 1.0 m (3.3 ft). Two 1.5 m long probe sections, with the bottom section fitted with 19 mm truncated tip, were used for testing. Figure 1 shows the probe set up prior to testing. 3.1 Test Soil The test soil was composed of 15% fine grained particles and 85% sand by dry weight. Percent dispersion of the fine grained fraction was estimated by performing Double Hydrometer test. Percent dispersion is the ratio of the dry mass of particles smaller than 0.005 mm diameter, without a chemical dispersant, to the same type of data from the hydrometer test but with a chemical dispersant, expressed as a percentage. A dispersion value higher than 50% was obtained for the fine grained soil, and therefore the fine fraction is classified as dispersive. The sand and the fine soil components were mixed thoroughly with an electrical mixer, in a drum, until a uniform mix was obtained. The mixing process was repeated after the soil was transferred to the test chamber (shown in Figure 1).

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Figure 1. Photograph of the Probe Set Up in the Chamber Prior to Testing

The chamber was filled up to 1m mark with the silty sand soil and approximately 0.45 kN weight was applied on the top, to induce consolidation, for a week. Several specimens were then retrieved for physical characterization of the test soil. Initial water content of the mixture ranged of 18% 23%. The results from the particle size analysis for three types of soils are shown in Figure 2, with the test soil designated as “Silty Sand” (all soils have been designated according to Unified Soil Classification System). Table 2 shows the physical and strength properties of the test soil, with the undrained strength estimated using the Fall Cone test.


which no scour is assumed to take place is estimated using the equation proposed by Briaud et al. (1999): Poorly Graded Sand

Percent Passing

100 90 80 70 60 50 40 30 20 10 0

 max  0.094 V 2 (

Sandy Clay

where, ρ is water density, Re = VD/υ is Reynolds number, V is depth average flow velocity at the location of the pier if the bridge were not there, D is pier diameter and υ is kinematic viscosity of water. In this case, assuming a pier diameter of 1 m and a depth of flow= 2 m, the critical shear stress value is estimated equal to 1.75 Pa.

Silty Sand

10

1 0.1 0.01 Particle Size (mm)

0.001

Figure 4 shows the equivalent penetration rate per shear stress function for the averaged test data. During testing, the probe tip is in close proximity of the soil mass, and erosion occurs within the jet potential core. The applied shear stress in this case is estimated using the relationship presented by Annandale (2006) as:

Figure 2. Grain Size Distribution of Test Soil: Silty Sand Table 2. Properties of the Test Soil Dry unit weight (kN/m3) = 17.7

4

Mean Particle Diameter, D50 (mm) = 0.26

Undrained Shear Strength, Cu (kPa) =5-8

PI= NonPlastic

  C f U 2

Figure 3 shows the results from the testing using four different run times. Based on the results from previous testing in a sand pit, Gabr et al. (2012) evaluated a critical stream power (Pc) value = 24 Watts/m2 for a sand with D50 = 0.30 mm. Using a similar technique of extrapolation approach, the data in Figure 3 is extrapolated to zero penetration rate to yield an average Pc value of 16 Watts/m2 for the test soil. Similar to the observation by Gabr et al. (2012), a minimum of “45 sec” run time is needed to provide a reliable measurement of the penetration rate. To calculate critical shear velocity from the Pc, the following equation is used (Annandale, 2006): (2)

Figure 4. Computed Scour Around Circular Bridge Pier on Silty Sand Bed; Range between Arrows is Values Estimated using Empirical Equation .

Penetration rate (cm/s)

where, γ is unit weight of water, q is the discharge per unit area, and h is the hydraulic head including the jet velocity head. 3.5 3 2.5 2 1.5 1 0.5 0

Run time = 15s Run time = 30s Run time = 45s Run time = 60s 10

(4)

where:  = applied shear stress to bed in N/m2, U = average velocity of water at the tip (m/s),  = density (kg/m3), and Cf is a friction coefficient = 0.016 according to Annandale (2006).

TEST DATA INTERPRETATION

Pc  qh

1 1  ) log Re 10

100 Stream Power (Watts/m2)

Figure 3. Extrapolation of the Stream Power to Assess Critical Stream Power Value (Pc)

Using Equation 2, a critical velocity is back calculated as 0.32 m/s corresponding to a critical stream power value of 16 Watts/m2. As flow field changes around the pier, and therefore the flow-related shear stress, the shear stress below

The slope of the data for the 45 and 60 secs in Figure 3 provides a parameter equivalent to the detachment rate coefficient (kd) proposed by Mehta (1991). The kd values of 0.017 cm/sec per N/m2 and 0.015 cm/sec per N/m2 are estimated, respectively, for the test soil at run times of 45 and 60 seconds. In comparison, kd values of 0.017 cm/sec per N/m2 and 0.013 cm/sec per N/m2 were observed respectively for sand at run times of 45 and 60 seconds by Gabr et al (2012). The Kd value obtained at 60 sec for the test soil is approximately 13% higher than the value obtained for sand. This observation agrees with the conclusion made by Ansari et al. (2002), where the authors indicated that in a lower fines content (<20%) type of soil, non-plastic fine particles are carried away as the resistance due to “cohesion” becomes insignificant. 5

SCOUR AROUND BRIDGE PIERS

Local scour around bridge piers occurs due to induced shear stresses associated with flow field changes. Ettema et al.

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(2011) indicated that the estimation of time-dependent clear water scour magnitude at bridge piers remains a challenge during the limited duration of excessive flow as, for example, in the case of a storm surge. Equation 3 proposed by Briaud et al. (1999) provides an estimate of the max as a function of the flow velocity at a round pier. However, in order to assess the scour depth with time using ISEEP data, a reduction in max with the progression of scour depth is needed. While a significant number of studies have been performed for the assessment of maximum scour at piers, these approaches were not specifically concerned with evolution of shear stresses with time and flow field. Data presented by Briaud et al. (1999) indicated a nearly linear relationship between the tmax and scour depth) /d (pier diameter), where tis the shear stress with the progression of scour depth with a minimum value of critical. An iterative approach is used to estimate tsince the maximum depth of scour is not known a’priori. The scour depth around a bridge pier is estimated for a flow velocity range of 1.0 m/s to 2.0 m/s (Froude number 0.23 to 0.45) with a pier diameter of 1 m and a depth of flow = 2 m. The computations are performed based on the ISEEP data and compared with the values from the empirical equation by Ansari et al. (2002), as the conditions for Ansari et al. (2002) empirical equation are in agreement with the percent fines in the test soil. Figure 4 shows scour depth, normalized with respect to the pier diameter ratio, versus time for different flow velocities. The scour depth from the equation by Ansari et al. (2002) is within 4.6-8 D (D = pier diameter) for a flow velocity range of 1.0 m/s to 2.0 m/s, which is higher than values estimated using the ISEEP data. A reason for the deviation can be attributed to the fact that the maximum shear stress equation developed by Briaud et al. (1999) was for clay, while the soil in this study is 85% sand. Furthermore, the application of Ansari et al. (2002) approach required the definition of the scour level in sand first which can widely vary depending on the parameters assumed. 6

SUMMARY AND CONCLUSIONS

The ISEEP approach developed by Gabr et al. (2012) was used to provide parameters for evaluating scour potential with time in a 15%-85% silty sand mixed bed. Soil erosion parameters included critical shear stress and detachment rate coefficient. Higher detachment rate was obtained for the silty sand than the sand soil. Application of the ISEEP data to assess magnitude of scour with time for a circular bridge pier indicated a scour depth on the order of 1-6 m versus 4.6-8 m estimated by an empirical equation in literature. The difference in results is attributed to the difference in the approach for scour computation and the limitations of estimating the evolution of shear stresses with time and flow field. In this case, the applicability of the empirical equation was somewhat limited since the testing conditions deviated in terms of soil type and moisture conditions. 7

ACKNOWLEDGEMENTS

This work is supported by the US Department of Homeland Security under Award Number: 2008-ST-061-ND 0001. The

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views and conclusions contained in this document are those of the authors and should not be interpreted as necessarily representing the official policies, either expressed or implied, of the U.S. Department of Homeland Security. 8

REFERENCES

Annandale, G. W. and Parkhill, D. L. 1995. Stream Bank Erosion: Application of the Erodibility Index Method. Proceedings of International Water Resources Engineering Conference, 2, 1570-1574. Annandale, G.W. 2006. Scour Technology: Mechanics and Practice. McGraw Hill, New York. Ansari, S.A., Kothyari, U.C. and Ranga Raju, K.G. 2002. Influence of Cohesion on Scour around Bridge Piers. Journal of Hydraulic Research, 40 (6), 717–729. Briaud, J.-L, Ting, F. C. K., Chen, H. C., Gudavalli, R., Perugu, S. and Wei, G. 1999. SRICOS: Prediction of Scour Rate in Cohesive Soils at Bridge Piers. Journal of Geotechnical and Geoenvironmental Engineering, 125 (4), 237–246. Briaud, J.-L., Ting, F.C.K., Chen, H.C., Cao, Y., Han, S.-W. and Kawk, K. 2001. Erosion Function Apparatus for Scour Rate Predictions. Journal of Geotechnical and Geoenvironmental Engineering, 127 (2), 105-113. Briaud J.-L., Chen, H.-C., Li, Y., Nurtjahyo, P., Wang, J. 2004. The SRICOS-EFA method for complex piers in fine grained soils,” Journal of Geotechnical and Geoenvironmental Engineering, ASCE, Vol 130, No. 11, p1180-1191. Debnath, K. and Chaudhuri, S. 2010a. Laboratory Experiments on Local Scour Around Cylinder for Clay and Clay–Sand Mixed Beds. Engineering Geology, 111, 51–61. Debnath, K. and Chaudhuri, S. 2010b. Bridge Pier Scour in ClaySand Mixed Sediments at Near-Threshold for Sand. Journal of Hydraulic Engineering, 136(9), 597-609. Ettema, R., Constantinescu, G. and Melville, B. 2011. Evaluation of Bridge Scour Research: Pier Scour Processes and Predictions , Contractor’s Final Report for NCHRP Project 24-27(01), National Cooperative Highway Research Program, March 2011, 181. Gabr, M., Caruso, C., Key, A. and Kayser, M. 2012. Assessment of In Situ Scour Profile in Sand Using a Jet Probe. Journal article accepted in ASTM Geotechnical Testing Journal, In Press. Hanson, G. J. and Cook, K. R. 2004. Apparatus, Test Procedures, and Analytical Methods to Measure Soil Erodibility In Situ. Applied Engineering in Agriculture, Vol. 20, No. 4, 455-462. Hosny 1995. Experimental Study of Scour around Circular Piers in Cohesive Soils. Ph.D. Dissertation, Colorado State University. Kothyari, U.C., Garde, R.J. and Ranga Raju, K.G. 1992. Temporal variation of scour around circular bridge piers. Journal of Hydraulic Engineering, 118(8), 1091-1105. Lu, J-Y., Hong, J-H., Su, C-C., Wang, C-Y. and Lai, J-S. 2008. Field Measurements and Simulation of Bridge Scour Depth Variations during Floods. Journal of Hydraulic Engineering, 134(6), 810-821. Mehta, A. J. 1991. Review Notes on Cohesive Sediment Erosion. In: N.C. Kraus, K.J. Gingerich, and D.L. Kriebel, (eds.), Coastal sediment ’91, Proceedings of Specialty Conference on Quantitative Approaches to Coastal Sediment Processes, ASCE; pp.40– 53. Molinas, A. and Hosny, M. 1999. Experimental Study on Scour around Circular Piers in Cohesive Soil. Publication No. FHWA-RD- 99-186, Federal Highway Administration, U.S. Department of Transportation, McLean, VA. Richardson, E.V., and Davis, S.R., 2001, Evaluating Scour at Bridges (4th ed.), Federal Highway Administration Hydraulic Engineering Circular No. 18, FHWA NHI 01-001.

Practical Reviews on CO2 Sequestration in Korean Sedimentary Basins and Geophysical Responses of CO2-injected Sediments Le comportement pratiques sur la séquestration du CO2 dans les bassins sédimentaires coréens et responses géophysiques de CO2 injectées sédiments Kim A.R., Cho G.C. Department of Civil and Environmental Engineering, KAIST, Korea Kwon T.H. Department of Civil and Environmental Engineering, Washington State University, USA Chang I.H. Geotechnical Engineering Research Division, Korea Institute of Construction Technology, KICT, Korea, Co. ABSTRACT: Geological CO2 sequestration is an effective means of reducing the emission of carbon dioxide. The Korean government aims to reduce CO2 emissions by 30% comparing to the usual amounts of emissions by 2020. It is expected that geological CO2 storage technology will account for more than 10% of the reduction of CO2 emissions. The forward strategies and technologies of CO2 sequestration in Korea need to be determined depending on the geological conditions of potential sites in Korea; moreover, the geophysical characteristics of CO2 and the reservoirs depend on the geological conditions. However, previous domestic studies related to geological conditions and the geophysical behavior of Korean sedimentary basins are rare thus far, with only a few studies focusing on numerical modeling. This study aims to review the geological characteristics of CO2 storage projects around the world and in Korea while also discussing the suitability for CO2 sequestration. Moreover, a laboratory approach simulating an in-situ high effective stress condition with silty sand from the Bukpyeong basin is attempted in an effort to determine the geophysical behaviors. This study offers an improved understanding of the possibility and potential of CO2 sequestration in Korea. RÉSUMÉ : La séquestration géologique du CO2 est un moyen efficace de réduire les émissions de dioxyde de carbone. Le gouvernement coréen a pour objectif de réduire les émissions de CO2 de 30 % à l’échéance de 2020. Il est prévu que la technologie géologique de stockage de CO2 représentera plus de 10% de la réduction des émissions de CO2. Les stratégies futures et les technologies de séquestration du CO2 en Corée doivent être déterminées en fonction des conditions géologiques des sites potentiels en Corée, d'ailleurs, les caractéristiques géophysiques de CO2 et les réservoirs dépendent des conditions géologiques. Toutefois, les précédentes études nationales relatives aux conditions géologiques et géophysiques sur le comportement des bassins sédimentaires de la Corée sont rares à ce jour, avec seulement quelques études mettant l'accent sur la modélisation numérique. Cette étude vise à examiner les caractéristiques géologiques des projets de stockage de CO2 dans le monde et en Corée tout en discutant de leur pertinence pour la séquestration du CO2. En outre, une approche de laboratoire simulant un état in situ à haute contrainte effective avec du sable limoneux du bassin Bukpyeong est tentée dans le but de déterminer les comportements géophysiques. Cette étude améliore la compréhension et la possibilité ainsi que le potentiel de séquestration du CO2 en Corée. KEYWORDS: CO2 sequestration, korean marine sediment, geological condition, geophysical behavior 1

INTRODUCTION

Recently, several methods have been proposed to mitigate carbon dioxide (CO2) emissions and to decrease the atmospheric concentration of CO2, including material recycling, the usage of renewable energy, and nuclear fusion. Among these, carbon capture and storage (CCS) strategies are considered as effective methods of reducing the atmospheric concentration of CO2 in a relatively short time at a low cost compared to other technologies (Espinoza et al. 2011; Pires et al. 2011). In light of this, approximately 40 CCS projects (including pilot-tocommercial scale applications) are in the planning or operational stages around the world (Hosa et al. 2010). It has been reported that the increasing rate of CO2 emissions in South Korea is the highest among OECD member countries, making Korea the seventh largest CO2 emitting nation in the world (BP 2011). Since 1999, the atmospheric concentration of CO2 in Korea has always been higher than the global average (Figure 1). The Korean government plans to reduce its CO2 emissions by 30% compared to the current business as usual (BAU) value by 2020 (i.e., about 244 Mton/yr; Presidential Committee on Green Growth, 2011). It is prospected that geological CO2 storage technology will account for more than 10% of global CO2 emissions (approximately 25 Mton/yr). As a part of this effort, pilot (10000 tons of CO2) and demonstration (100000 tons of CO2) scale CO2 sequestration projects, capable of storing more than 1 Mton of CO2 in total, are currently planned and being conducted with commercial considerations in

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Korea (Presidential Committee on Green Growth 2011). This effort mostly relies on existing geologic data and resources that were gathered during oil explorations and drilling projects. However, the current geologic information on onshore and offshore deep subsurface areas of Korea is insufficient. Comprehensive geological exploration and database construction activities are critical for characterizing, selecting, or at least screening potential storage sites for CO2 sequestration.

Figure 1. Atmospheric concentration of CO2 of the world and in Korea. Data were gathered from Climate Change Information Center (www.climate.go.kr) and National Oceanic and Atmospheric Administration (www.noaa.gov).


Proper strategies and technologies for storing CO2 in geologic formations must be determined with deep consideration of the geological conditions of the potential sites. Geological conditions (e.g., the pressure, geothermal gradient, geology, geochemical characteristics, and mineralogy) govern the interpretation of the geophysical responses of CO2-storage reservoirs. Therefore, a fundamental understanding of the geophysical responses of CO2-containing sediments in Korean sedimentary basins is required. However, while a few studies have been conducted on numerical reservoir modeling, there have been few efforts to examine the geological suitability for CO2 storage and on geophysical characterizations of Korean sedimentary basins to date. This study provides a review and discussion of the geological conditions and suitability of the potential CO2 storage sites in Korea as well as CO2 storage sites around the world. Moreover, laboratory experiment results on the P-wave velocity and the electrical responses of CO2-injected sediments are presented, in which high in-situ effective stress conditions were simulated on natural samples cored from the eastern Bukpyeong basin, which is one of the candidate sites for geologic CO2 storage. 2 SITE CHARACTERIZATION FOR CO2 SEQUESTRATION 2.1

Site-dependent strategies

CO2 can be stored in various geological formations, such as (1) deep saline formations, (2) coal beds, (3) depleted oil and gas fields, and (4) oil and gas reservoirs during enhanced oil and gas recovery efforts (EOR and EGR). Coal beds absorb and contain gaseous CO2 in micro-pores in coal. However, the temperature and pressure effects on the CO2 trapping process in coals are not well understood (Larsen, 2003). CO2 storage in depleted oil and gas fields or the use of CO2 for EOR (or EGR) have been proven as effective GCS methods due to the geological suitability for CO2 storage, the existence of geophysical and geological data, and the ready-made infrastructure used for oil and gas production. However, several problems remain poorly identified, including well plugging, leakage induced by the overpressure of pore fluids, and the injection depth. Deep saline formation is the most promising method for safe and effective CO2 storage due to its vast capacity. The potential storage capacity of deep saline formation is expected to be at least 1000 GtCO2, which is approximately 200 to 300 times higher than the potential storage offered by oil or gas fields and coal seams (IPCC, 2005). In particular, sedimentary basins that have permeable formations (e.g., sandstone) with overlying low-permeable seals (caprocks) are effective for both CO2 injection and CO2 leakage prevention. The target depth is deeper than 1000 meters below the ground surface. 2.2

Selected sites for CO2 sequestration in America, Europe, and Asia

There are more than 800 sedimentary basins around the world (St John et al., 1984). Approximately 40 CO2 storage projects are under operation or in planning in North America, Europe, Australia, and Asia. Among them, ten onshore projects in the USA and two onshore projects in Canada are being conducted. Two onshore projects and three offshore projects in Europe are being undertaken, and two demo onshore projects in Japan are being tested (Hosa et al., 2010). Additionally, other potential sedimentary basins have been investigated and proposed for pilot-scale testing. For instance, the Alberta basin in Canada, where natural hydrocarbon resources had been found, was evaluated to be the most suitable basin in Canada owing to the existence of adjacent infrastructure (Bachu, 2003). The offshore Gippsland basin in Australia is considered as an effective target for CO2 storage due to its complex stratigraphy,

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high injectivity, low-permeable marginal reservoir, the existence of several depleted oil fields, and its long migration pathways (Gibson-Poole et al., 2008). In China, an ECBM (enhanced coalbed methane recovery) pilot test and a singlewell micro-pilot test were successfully performed at the South Qinshui basin (Wong et al., 2010). One of the most well-known CO2 storage attempts is the Sleipner project, targeting the Utsira Sand formation, which was launched in 1996. It was the first commercial-scale project to store CO2 in a saline formation. The geologic condition of this site is a brine-saturated sandstone layer (250 m thick) with an overlying thin shale cap layer. Its storage capacity is expected to be 25 MtCO2 (Hosa et al., 2010). The Nagaoka project at Nagaoka City, Japan, was the first pilot-scale attempt in Asia. In this pilot-scale test, CO2 was injected into a Haizume-formation sandstone layer. The injection efficiency differs between the two formations (e.g., the Utsira Sand formation and the Haizume formation). In detail, the Utsira Sand formation (2800 tons/day) has a storage capacity of approximately 70 times that of the Haizume formation (a maximum 40 tons/day), as the permeability of the Haizume formation (i.e., 6 mD) is much lower than that of the Utsira Sand formation (i.e., 5 D) (Hosa et al., 2010) though both formations have relatively high porosities (37% for the Utsira formation and 22.5% for the Haizume formation) and similar injection depths (about 1000‒1100 m). Therefore, it can be tentatively concluded that the permeability is a major controlling parameter for CO2 injectivity rather than the porosity or injection depth. 2.3

Geologic characteristics of sedimentary basins in Korea

There are several potential geologic formations for GCS on the Korean Peninsula. The porosity, storage capacity, and geologic characteristics of those proposed sites are listed in Table 1. However, the geological information of offshore basins is still poorly identified due to insufficient exploration. Table 1. Porosity and storage capacity of potential Korean geological storage sites Porosity Storage Capacity Geologic (%) (Mton) Characterstics Bukpyeong N.A. 8771 Saline aquifer Ulleung 10.62 3,0182 EGR Jeju 15.72 95,1012 Saline aquifer Gunsan 102 2542 Saline aquifer Heuksan N.A. N.A. Saline aquifer Pohang N.A. 382 Saline aquifer 1 The values are from Kim et al. (2011) and 2MEST (2008). Basin

Korean offshore sedimentary basins have thickness ranges from 3 km to 10 km (MEST, 2008). The vast coverage area and high porosity of Korean offshore sedimentary basins are expected to show that these have larger storage capacities than onshore basins. Moreover, Korean offshore sedimentary basins show geologically structural similarity with natural hydrocarbon reservoirs, which indicates good suitability for GCS. For an example, the Gunsan basin and the Jeju basin show high potential for GCS because the geologic structures are similar to natural hydrocarbon reservoirs in analogous Chinese basins (Hong et al., 2005). The Ulleung basin contains natural gas deposits and is located more than 1000 m below the sea level. Thus, structural trapping may be feasible (Hong et al., 2005). To evaluate the storage and economic efficiencies of sedimentary basins in Korea, a systematic and quantitative evaluation method (Bachu 2003) was employed in this study. Fifteen criteria (e.g., geological characteristics, basin resources, maturity, and infrastructure, among others) are considered with weight factors to assess the suitability (Table 2). Bachu’s (2003) method classifies the proposed sites with dimensionless values between 0 and 1. The value can be used as a decision criterion


for assessing the suitability of the proposed sedimentary basins for geological sequestration and for comparing it with basins in other countries in which pilot- and commercial-scale projects are already underway. Table 2. Scores and weight assigned to the criteria and classes for assessing sedimentary basin in terms of their suitability for CO2 sequestration in geological media (Bachu, 2003) Scores Weight Criterion j=1 j=2 j=3 j=4 j=5 Tectonic setting 1 3 7 15 15 0.07 Size 1 3 5 9 0.06 Depth 1 3 5 0.07 Geology 1 3 7 0.08 Hydrogeology 1 3 7 0.08 Geothermal 1 3 7 0.10 Hydrocarbon potential 1 3 7 13 21 0.06 Maturity 1 2 4 10 0.08 Coals and CBM 1 2 5 0.04 Salts 1 2 3 0.01 On/offshore 1 4 10 0.10 Climate 1 2 4 7 11 0.08 Accessibility 1 3 6 10 0.03 Infrastructure 1 3 7 10 0.05 CO2 Sources 1 3 7 15 0.09

The evaluation results are summarized in Figure 2. Among other sites, the Ulleung basin was evaluated to be the most suitable site for geologic CO2 storage in Korea due to the presence of nearby infrastructure as constructed for natural gas recovery. 1

While CO2 leakage from CO2 stored sites could cause serious environmental problems, geophysical survey techniques are viable methods to detect CO2 leakage and to identify CO2 movement. Therefore, understanding the geophysical responses of Korean sediments during CO2 injection and storage is important to ensure safety. The most widely used geophysical methods are seismic surveys using P-waves and electrical resistivity surveys (Nakatsuka et al., 2010). In particular, feasibility of P-wave surveys for the detection of CO2 has been examined in laboratory tests (Shi et al., 2007; Siggins et al., 2010; Xue and Lei, 2006). Also, it is well known that CO2-containing formations have less stiffness than brine-saturated formations do (Daley et al., 2008; Lazaratos and Marion, 1997; Mito and Xue, 2011). Because the physical properties of unconsolidated sediments are significantly affected by the effective stress as well as the formation characteristics, achieving an in-situ effective stress condition is critical to obtain reliable physical properties of CO2-containing sediments, though this may not be the case for cemented porous media. This section presents the geophysical responses of sediments in which an in-situ effective stress condition of the potential CO2 storage sites was achieved. 3.1

Experimental study

The sediment sampled from the Bukpyeong basin (located on the east coast of Korea; see Table 1) was used in this study. The silty sand sample was compacted into a rigid-walled vessel and was saturated with water. Vertical in-situ effective stress of 15 MPa was then applied. The final porosity resulting from the applied stress condition was estimated to be 49%. Table 3. Properties of test specimen property Soil type Specific (USCS) gravity specimen Silty sand 2.73

0.8

Score

3 GEOPHYSICAL RESPONSES OF CO2-CONTAINING SEDIMENTS

0.6 0.4 0.2 0

Figure 2. Scores for suitability for of Korean sedimentary basins.

However, the proposed sites in Korea are less feasible for geologic CO2 storage compared to the basins in Canada and Australia. Specifically, the Korean basins show low scores on the following criteria: size, hydrocarbon potential, maturity, and infrastructure. Most Korean sedimentary basins, except the Ulleung basin, are estimated to be of a small-to-medium size, whereas the Otway and Gippsland basins in Australia are categorized as large and the Alberta and Williston basins in Canada are known to be large to giant in size. Moreover, a lack of boring studies and geophysical exploration exacerbate the problems of low maturity and insufficient infrastructure. Thus, more data acquisition and exploration are required to enhance the reliability of numerical modeling and simulations for the first Korean pilot project. Additionally, a new alternative approach involving the use of deep saline formations should be considered for safe and economic CO2 sequestration in Korea with consideration of the geological characteristics of Korean basins and their limitations as regards CO2 injection.

569

Permeability (cm/s) 4.2*10-5

D50 (mm) 0.15

The test was performed on saturated silty sand under a supercritical temperature condition (35°C) under the effective stress (15 MPa). The pore water pressure was maintained at 8 MPa using a back pressure regulator. The specimen was then flooded with supercritical CO2. After a pre-determined amount of CO2 was injected into a water-saturated sample, the geophysical responses (P-wave velocity VP and electrical resistivity) were measured and the amount of injected CO2 was estimated by measuring the amount of water expelled from the vessel. 3.2

Results and analysis

The total amount of water expelled by weight was 29.86 g; thus, approximately 30% of the pore water was displaced with supercritical CO2, indicating that the CO2 saturation rate was ~30%. The density and solubility of CO2 at the target temperature and pressure are 591.85 kg/m3 (NIST) and 1.35 mol/kg of water (Duan and Sun 2003), respectively. Accordingly, the total calculated amount of injected CO2 was 17.29 g. Figure 3 shows the decrease in VP during the CO2 injection process. The decrement range is relatively low compared to preexisting studies (Shi et al. 2007; Siggins et al. 2010). This can be explained by the fact that the CO2 injectivity in sandy soil specimens is lower than that in sandstone. The electrical resistivity increased rapidly as CO2 was introduced. The convergence of the electrical resistivity indicates a fully saturated condition in which no more CO2 can be injected (see Figure 4). The overall results show that CO2 can be detected by measuring the geophysical properties; however, the injected


amounts and readings of the movement of CO2 are not highly precise.

P-wave velocity (m/s)

1730 1728 1726 1724 1722 1720

0

50

100

150

Time (min)

Figure 3. The effect on the P-wave velocity of the sediment during CO2 injection.

Electrical Resistivity (Ω)

86

84

82

80

0

100

200

300

400

500

600

700

Time (sec)

Figure 4. The electrical resistivity of the sediment during CO2 injection.

4

CONCLUSION

The presented study explores practical reviews on Korean marine sediments for CO2 sequestration in relation to geological and geotechnical considerations. The geological conditions of off-shore sedimentary basins in Korea were investigated, and the suitability of the basins for CO2 storage were evaluated. The Ulleung basin were found to be the most suitable site for GCS, although their scores were lower than the scores of some basins where CO2 storage is currently undergoing or pilot-tested in Canada and Australia. Geophysical behavior of CO2-storing sediments is available for field application to monitor CO2 movements and leakages. A laboratory scale experiment simulating the in-situ condition for measuring geophysical properties, and the results showed that CO2 can be detected by measuring geophysical properties but further study is required to exact understanding geophysical behavior of CO2-storing Korean marine sediments. 5

ACKNOWLEDGEMENTS

This research was supported by a grant from the Korea Electric Power Corporation (KEPCO) and by the Block Funding Project (GP2012-030) of Korea Institute of Geoscience and Mineral Resources (KIGAM) transferred from the Energy Efficiency & Resources Program of the Korea Institute of Energy Technology Evaluation and Planning (KETEP) grant, funded by the Ministry of Knowledge Economy, Republic of Korea. 6

REFERENCES

Bachu, S. 2003. Screening and ranking of sedimentary basins for sequestration of CO2 in geological media in response to climate change. Environmental Geology, 44(3), 277-289. BP. 2011. BP Statistical Review of World Energy June 2011. The report can be downloaded at the website http://www.bp.com/statisticalreview.

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Daley, T., Myer, L., Peterson, J., Majer, E., and Hoversten, G. 2008. Time-lapse crosswell seismic and VSP monitoring of injected CO2 in a brine aquifer. Environmental Geology, 54(8), 1657-1665. Duan, Z., and Sun, R. 2003. An improved model calculating CO2 solubility in pure water and aqueous NaCl solutions from 273 to 533K and from 0 to 2000 bar. Chem. Geol. 193, 257-271. Espinoza, D., Kim, S., and Santamarina, J. C. 2011. CO2 geological storage — Geotechnical implications. KSCE Journal of Civil Engineering, 15(4), 707-719. Gibson-Poole, C. M., Svendsen, L., Underschultz, J., Watson, M. N., Ennis-King, J., van Ruth, P. J., Nelson, E. J., Daniel, R. F., and Cinar, Y. 2008. Site characterization of a basin-scale CO2 geological storage system: Gippsland Basin, southeast Australia. Environmental Geology, 54(8), 1583-1606. Hong, S. K., Lee, H., Egawa, K., Choi, T., Lee, M. K., Yoo, K. C., Kim, J. H., Lee, Y. I., and Kim, J. M. 2009. Preliminary evaluation for carbon dioxide storage capacity of the Chungnam, Taebacksan, Mungyeong and Honam basins. Journal of the Geological Society of Korea, 45(5), 449-462 (in Korean). Hosa, A., Esentia, M., Stewart, J., and Haszeldine, S. 2010. Injection of CO2 into saline formations: benchmarking worldwide projects. Chemical Engineering Research and Design, doi:10.1016/j.cherd.2011.04.003. IPCC, 2005, Special Report on Carbon Dioxide Capture and Storage. In: Metz., B., Davidson, O., de Coninck, H. C., Loos, M., Meyer, L. A., (eds.), Prepared by Working Group III of the Intergovernmental Panel on Climate Change, Cambridge University Press, Cambridge, UK and New York, NY, USA Kim, J. M., Kim, J. H., and Park, S. W. 2011. Evaluation of CO2 storage capacity of Bukpyeong Basin using three-dimensional modelling and thermal-hydrological nemerical modelling. 1st Korea CCS Conference (Abstract), Jeju, April 13-15 (in Korean). Larsen, J.W. 2003. The effects of dissolved CO2 on coal structure and properties. International Journal of Coal Geology, 57, 63-70. Ministry of Educational Science and Technology, 2008, Characterization and evaluation of geologic formation for geological sequestration of carbon dioxide. 21st Century Frontier R&D Program (in Korean). Mito, S., and Xue, Z. 2011. Post-Injection monitoring of stored CO2 at the Nagaoka pilot site: 5 years time-lapse well logging results. Energy Procedia, 4, 3284-3289. Nakatsuka, Y., Xue, Z., Garcia, H., and Matsuoka, T. 2010. Experimental study on CO2 monitoring and quantification of stored CO2 in saline formations using resistivity measurements. International Journal of Greenhouse Gas Control, 4(2), 209-216. NIST Thermophysical Properties of Fluid System, 2007. National Institute of Standards and Technology, Available on line: (http://webbook.nist.gov/chemistry/fluid). Accessed: October 2007. Pires, J. C. M., Martins, F. G., Alvim-Ferraz, M. C. M., and Simoes, M. 2011. Recent developments on carbon capture and storage: An overview. Chemical Engineering Research and Design, In Press, Corrected Proof. Hosa, A., Esentia, M., Stewart, J., and Haszeldine, S. 2010. Injection of CO2 into saline formations: benchmarking worldwide projects. Chemical Engineering Research and Design, doi:10.1016/j.cherd.2011.04.003. Shi, J.-Q., Xue, Z., and Durucan, S. 2007. Seismic monitoring and modelling of supercritical CO2 injection into a water-saturated sandstone: Interpretation of P-wave velocity data. International Journal of Greenhouse Gas Control, 1(4), 473-480. Siggins, A. F., Lwin, M., and Wisman, P. 2010. Laboratory calibration of the seismo-acoustic response of CO2 saturated sandstones. International Journal of Greenhouse Gas Control, 4(6), 920-927. St John, B., Bally, A. W., and Klemme, H. D. 1984. Sedimentary provinces of the world hydrocarbon productive and nonproductive. American Association of Petroleum Geologists, Tulsa. Wong, S., Macdonald, D., Andrei, S., Gunter, W. D., Deng, X., Law, D., Ye, J., Feng, Z., and Ho, P., 2010, Conceptual economics of full scale enhanced coalbed methane production and CO2 storage in anthracitic coals at South Qinshui basin, Shanxi, China. International Journal of Coal Geology, 82(3-4), 280-286. Xue, Z., and Lei, X. 2006. Laboratory study of CO2 migration in watersaturated anisotropic sandstone, based on P-wave velocity imaging. Exploration Geophysics, 37(1), 10-18.

Using Multi-scale Sediment Monitoring Techniques to Evaluate Remediation Effectiveness of the Tsengwen Reservoir Watershed after Sediment Disasters Induced by Typhoon Morakot

Lin B.-S., Ho H.-C., Hsiao C.-Y., Keck J., Chen C.-Y., Chi S.-Y.

Disaster Prevention Technology Research Center, Sinotech Engineering Consultants, Taiwan

Chien Y.-D., Tsai M.-F.

Soil and Water Conservation Bureau, Council of Agriculture, Taiwan

ABSTRACT: The 2009 typhoon Morakot dumped more than 3,005 mm of rain in mountain areas of the Tsengwen reservoir watershed and caused unprecedented landslide and sediment-related disasters. Subsequently, the storage capacity of the Tsengwen reservoir was drastically reduced. In order to increase the longevity of the reservoir and also protect ecosystems and the peoples living in the upper portions of the watershed, the Taiwan Executive Yuan implemented the "Tsengwen, Nanhua, Wushantou Reservoir Remediation and Water Resources Protection Act". This study aims to use multi-scale sediment monitoring techniques including field investigations and multi-stage remote sensing data to identify sediment migration patterns associated with remediated areas of the Tsengwen reservoir watershed after typhoon Morakot and to guarantee the effectiveness of remediation efforts. A case study of the Longjiao creek in Tsengwen Reservoir watershed shows that remediation works can not only reduced sediment production due to erosion and landslides, but future sediment production will also be suppressed. The reduction of sediments carried by the Tsengwen river will also lead to an increase in the service life of the Tsengwen reservoir. RÉSUMÉ : En 2009, le typhon Morakot a déversé plus de 3 005 mm de pluie dans les régions montagneuses du bassin versant du réservoir et provoqué des glissements de terrain Tsengwen sans précédent et les catastrophes sédiments. Par la suite, la capacité de stockage du réservoir Tsengwen a été considérablement réduit. Afin d'augmenter la longévité du réservoir et aussi de protéger les écosystèmes et les populations vivant dans les parties supérieures du bassin versant, le Taiwan Yuan exécutif mis en place le "Tsengwen, Nanhua, Wushantou réservoir d'assainissement et de protection des eaux Loi sur les ressources". Cette étude vise à utiliser multi-échelle des techniques de surveillance des sédiments, y compris les enquêtes sur le terrain et sur plusieurs périodes données de télédétection pour identifier les schémas de migration des sédiments associés à des zones assainies du bassin versant du réservoir Tsengwen après le typhon Morokot et de garantir l'efficacité des efforts d'assainissement. Une étude de cas du flux Longjiao dans le bassin versant du réservoir Tsengwen montre que les travaux de réhabilitation peuvent non seulement réduit la production de sédiments à cause de l'érosion et des glissements de terrain, mais la production de sédiments avenir seront également supprimés. La réduction des sédiments charriés par le fleuve Tsengwen conduira également à une augmentation de la durée de vie du réservoir Tsengwen. KEYWORDS: Tsengwen reservoir watershed, typhoon Morakot, sediment disasters, remediation effectiveness. 1 INTRODUCTION In 2009, typhoon Morakot brought heavy rainfall up to 3,005 mm, which was recorded at Alishan rainfall guage station of the Tsengwen reservoir watershed over a five day period. Also, the consecutive 72 hour accumulated rainfall exceeded historical records in Taiwan (SWCB, 2011). This typhoon event induced massive sediment-related disasters within the watershed, which caused about 91,080,000 m3 of sediment in the reservoir and exceeded the original design level (5,610,000 m3/yr). Afterward, Taiwan government passed “Tsengwen, Nanhua, Wushantou Reservoir Remediation and Water Resources Protection Act” and planned a project for managing and remediating sediment problems. The primary goals of the proposed project are to reduce reservoir turbidity levels, extend the service life of the dam and protect security of the upstream residents. Sediment transport and deposit within the watershed is an unavoidable natural process. It is very important to do field survey and monitor periodically especially in major sediment source areas including old debris flow, large-scale landslide and massive alluvial soil or river terrace deposits. Many researches has pointed out that the sedimentation of Tsengwen Reservoir has been serious in flood season due to intense geological activity. Recently, under the effect of global climate change, the probability of extreme weather occurrence has increased. In the mountain area, it can be observed that the magnitude of

571

disasters caused by water-sediment flows, induced by high intensity and long duration rainfall events, has increased (Lo et al., 2012; Lin et al., 2012). The mode of sediment transport can be classified in different ways, according to the mechanics of sediment transport process, from suspended load to debris flow. Therefore, the sediment deposited in the Tsengwen reservoir watershed comes from long-term deposits of the trunk river and soil erosion as well as slope landslides due to 2009 typhoon Morakot. It significantly affects water supply to residents and rapidly reduced storage capacity of reservoir. For validating and proving the effectiveness of remediation efforts after typhoon Morakot event, the study integrates multi-scale sediment monitoring techniques to collect time-dependent monitoring data and spatiotemporal remote sensing information including watershed scale, high-resolution airborne LiDAR DTMs. Then, using the data obtained from the remediated environmental area, remediation effectiveness of the Tsengwen reservoir watershed with regard to suppression of soil erosion, vegetative recovery rate, variation in amount of landslide and sediment trapping efficiency are quantified. Finally, the proposed procedure of this study will assist us to track remediation effectiveness, and reduce sediment yield entering a reservoir or trap eroded sediment for effective watershed management.


2 OVERVIEW OF ENVIRONMENT Study area

As illustrated in Figure 1, the Tsengwen reservoir is located in the southwestern portion of Taiwan. It is the most important water resource conservation hydraulic structure and the largest dam in Taiwan. The dam is 400 m in width and 133 m in height, and mainly serves irrigation, municipal water and power supply functions for the southern plains and downstream area of Chiayi county, Tainan and Kaohsiung city shown in Figure 2. The watershed area of Tsengwen Reservoir is approximately 481 km2, where Tsengwen river is the trunk river originating from Alishan mountain at elevation of 2,609 m a.s.l. The watershed shape is similar to a quadrilateral with elevations gradually increasing from southwest to northeast, and ranging from 100 m a.s.l and 2,700 m a.s.l. In general, most hill slopes are steeper than 28.8° and approximately represents over 60% of the study area. As for the aspect, slopes are mostly west-facing and southwest-facing in the watershed. There are many fault line and geologic structures and the geological condition of the watershed consists mostly of sandy shale, siltstone, and isolated areas of muddy sandstone, which are prone to more severe weathering and become weak layers in the rock strata. These conditions make the slope unstable during heavy rainfall or strong earthquake shaking. Hence, during the typhoon and flood season, the combination of huge rainfalls and local weak geological conditions easily permit the occurrence of sediment landslides (Lo et al., 2012). Due to high topographic relief, annual average temperature ranges from 24℃ in the plains and 11℃ in the mountainous parts of the watershed. According to Alishan rainfall gauge station, average annual accumulated ranges from 1,950 to 4,980 ㎜. Recent extreme rainfalls have caused annual accumulated rainfall of Taiwan to increase, especially for Alishan, where, since 2005, annual rainfalls have exceeded 5,000 ㎜ (see Figure 3). This rate is double the annual average precipitation (2,500 ㎜) for Taiwan and over four times of world annual average precipitation. Rainfall distribution increases from the plains to the mountains and is mostly concentrated between May and September when the watershed receives approximately 80% of the overall annual rainfall.

Figure 1. Graphical location of Tsengwen reservoir watershed in Taiwan (local coordinate system: TWD97).

572

Figure 2. Topographic map of Tsengwen reservoir watershed (local coordinate system: TWD97). 8000

Gr

7000 6000 5000 4000 3000 2000 1000 0 1959 1960 1961 1962 1963 1964 1965 1966 1967 1968 1969 1970 1971 1972 1973 1974 1975 1976 1977 1978 1979 1980 1981 1982 1983 1984 1985 1986 1987 1988 1989 1990 1991 1992 1993 1994 1995 1996 1997 1998 1999 2000 2001 2002 2003 2004 2005 2006 2007 2008 2009

2.1

Figure 3. Time series of annual rainfalls at Alishan gauge station.

2.2

Relation between reservoir sedimentation and major typhoon events

Presently, the greatest challenge of Tsengwen reservoir is sedimentation. Whether caused by anthropogenic or natural factors, both of them directly triggers problems such as increased turbidity and reduced reservoir storage volume. Figure 4 illustrates the historical trends of sedimentation in Tsengwen reservoir concerning major typhoon events. According to the figure, since completion of reservoir construction, typhoons repeatedly hit the Tsengwen reservoir. It can be found that the peaks in the historical sedimentation curve of Tsengwen reservoir correspond to major typhoon events. Before 2008, the annual average sedimentation volume is 4,760,000 m3 and still lower than the yearly designed value of 5610,000 m3. However, 2009 typhoon Morakot brought around 91,080,000 m3 of sediment into Tsengwen Reservoir, which occupies about 12% of the reservoir capacity. After the 2009 typhoon Morakot, the annual average sedimentation rapidly increases to 7,060,000 m3, exceeding the yearly designed value by 12.6 times. It is truly believed that massive amounts of sediment washed downstream. Also, this event seriously resulted in debris flows and large large-scale landsides along river flanks and close to human inhabitation in upstream areas, and threatens longevity of reservoir and significantly affects water supply to the south area in Taiwan.

Figure 4. Historical sedimentation curve of Tsengwen reservoir with the major typhoon events


3 MULTI-SCALE SEDIMENT MONITORING AND EVALUTION METHOD TO REMEDIATION EFFECTIVENESS Multi-scale sediment monitoring techniques is used in Tsengwen reservoir watershed to study remediation effectiveness and topographical changes. This section describes the method about how to systematically study and analyze soil erosion, landslide areas, and sediment trapping in the check dams from easily measured physical quantities such as depth, area, and volume by collecting time-dependent monitoring data and multi-stage remote sensing information in a watershed scale. A case study of the Longjiao subwatershed was chosen to be validated with remediation effort. The above proposed methods are detailed separately below. 3.1

Depth-based evaluation method

Soil erosion estimates were often based on empirical equations, such as USLE, MUSLE, and RUSLE, etc. These empirical equations are limited regionally and by spatial distribution of rainfalls. Therefore, this study focused on the different vegetated slopes to design erosion pins by some research reports (Schumn, 1956). Site surveys were conducted to measure surface erosion depth to investigate the state of slope soil after erosion from rainfalls. The result was used to assess the inhibition rate of soil erosion from both remediated and nonremediated hillslopes in order to understand the efficiency of remediation. To quantify the soil erosion suppression ratio (SSR) from the measured soil erosion depth of several erosion pins (see Figure 5) embedded in remediated and non-remediated hillslopes, we used an index value to depict efficiency of soil erosion retention after completing remediation. Higher index values indicate higher soil erosion suppression. Therefore, this study uses this index (SSR) to understand the remediation effectiveness of the hillslopes. SSR is defined as follows

SSR (%) 

E DR  E DN  100% E DR

(1)

Where SSR is soil erosion suppression ratio (%)；EDR is surface eroded soil depth of remediated hillslope (mm)；EDN is surface eroded soil depth of non-remediated hillslope (mm).

Figure 5. Schematic layout and photos of erosion pins embedded in remediated/non-remediated hillslope.

3.2

Area-based evaluation method

To understand the evolution of vegetation coverage of the Tsengwen reservoir watershed resulting from remediation efforts, multi-spectral high-resolution satellite images from different periods are adopted to analyze the ratio of green cover to assess the vegetation restoration after remediation. Normalized Difference Vegetation Index (NDVI) is currently a popular method to assess vegetation coverage (Kriegler et al., 1969). The NDVI is calculated from these individual measurements as follows: NDVI 

NIR  VIS NIR  VIS

(2)

573

where VIS and NIR stand for the spectral reflectance measurements acquired in the visible (red) and near-infrared regions, respectively. The NDVI value is normalized between 1.0 to 1.0. Values of NDVI above a certain threshold correspond to vegetation coverage area and values below the threshold correspond to non-vegetation coverage areas, as shown in Figure 6.

Figure 6. Orginal satellite image and the classified image from results of NDVI.

Once the NDVI has been used to classify the images into vegetated and non-vegetated zones, the ratio of vegetation coverage (VR) involved with the total area is estimated, as shown in Eq. (3).

VR (%) 

AV  100% AC

(3)

Where Ac is a given watershed area, and Av is vegetated area within a given watershed. 3.3

Volume-based evaluation method

This study compiled satellite images to identify the landslide distribution. Number of landslides, existing landslide area, incremental landslide area, and spatial distribution in key regions were obtained through digital interpretation to understand its evolution. Further, this was complemented with multi-period terrain data, established by airborne LiDAR, to quickly obtain information on terrain changes in each subcatchment area and assess the effectiveness of the remediation projects. To evaluate the effectiveness of check dams, this study utilized airborne LiDAR (Light Detection And Ranging) technology to survey and produce high resolution DEMs of the Tsengwen reservoir watershed. The pre-event DTM is subtracted from the post-event DTM. A negative value in the grid represents failure or erosion, and positive value indicates deposits. Variation in volume of a grid can be obtained by multiplying this value by the area of the unit grid (see Figure. 7). The total volume of landside material and sediment trapped by the check dam can also be obtained from multiple LiDAR generated DTMs. Then, sediment discharge and trapping efficiency of dams can be precisely calculated. It can also be applied to monitor the accumulated volume of sediment on the confluence between tributaries and river, growth of alluvial fan, and large scale wedge like slope failures. Comparison of LiDAR DEMs from different periods can also indicate terrain migration and be used to trace sediment transport from tributaries, especially in extreme typhoon disasters. Sediment trapping ratio (STR) can be assessed by measuring the volume of deposited sediment in front of the check dams (Sophie et al., 2008). If STR after remediation is higher than before remediation, it means that check dams are effectively controlling sediment transport and have adequate remediation efficiency levels. The sediment trapping ratio can be expressed as：

STR (%) 

Vd  100% Vy

(4)

Where Vd is the trapped volume in the check dam(m3)；Vy is the sediment yield from upstream (m3).


Grid Subtraction 6

11

11

13

16

14

14 16

2nd DEM Grid Map

3

10

15

17

16

19

15 = 19

3

1

-4

-4

0

-5

1st DEM Grid Map

Variation of Volume = Result x Unit grid area

No data

-1

-3

Results

Erosion Deposition

Figure 7. Grid subtraction in post-event and pre-event digital terrain elevations

4 CASE STUDY In this paper, case study chooses Longjiao creek located at downstream area of Tsengwen Reservoir (see Figure 2) for proving the effectiveness of remediation efforts after typhoon Morakot, based on real data from multi-scale sediment monitoring techniques. Then, the proposed depth-area-volume based methods are all applied to evaluate the ratios of soil erosion suppression, vegetation coverage and sediment trapping in the following sections. 4.1

Soil erosion suppression

To effectively measure surface soil loss on remediated and nonremediated hillslopes, ten erosion pins were installed on each of the hillslopes types to monitor eroded soil depth for each rainfall events. The monitoring period is from May 14., 2011 to Oct. 04, 2011. Figure 8 is the diagram of the average accumulated eroded soil depth on remediated and nonremediated hillslopes. According to the figure, soil erosion of remediated hillslopes is obviously lower than the remediated. Compared with others, remediated hillslopes can reduce soil erosion by about 1.6 mm. This data is input into Eq. (1) and then the SSR of Longjiao subwatersehd is obtained as below：

SSR (%) 

18.7  7.25  100%  61.23% 18.7

(5)

The calculated result shows that remediation of hillslops could reduce erosion amounts by 61.23% of soil loss per unit area and time. It is evident that remediation can accelerate environmental vegetation recovery and under good practical sediment control. 20.00

18.7

12.00 7.25

4.00 0.00

Figure 8. Diagram of average accumualted eroded soil depth on remediated and non-remediated hillslopes.

4.2

4.3

Sediment trapping

Soil and Water Conservation Bureau (2011) has collected three high-precision digital elevation models from aerial orthoimages and airborne LiDAR. These measurements can be divided into pre-remediation and post-remediation. Further, sediment yield is the total volume of terrain changes such as slope failures and river erosions by grid subtraction of DTMS. Sediment trapping ratio (STR) can be assessed by measuring the amount of sediment trapped in front of the check dams, which has been listed in Table 1. Compared with the results listed in table 1, post-remediation STR of Longjiao subwatershed is significantly higher than pre-remediation by 17.18 times. In the meantime, the sediment yield after remediation is lower than before remediation. Through the above results, it was found that sediment yields were effectively controlled. Table1 List of Sediment trapping ratio of Longjiao subwatershed Duration

Sediment yield (m3)

Sediment trapping (m3)

STR (%)

pre-remediation

2008~2010

1,548,300

34,540

2.2

post-remediation

2010~2011

149,143

56,373

37.8

Stages

5 CONCLUSIONS This study systematically integrates multi-scale sediment monitoring techniques to analyze soil erosion, vegetation coverage, and sediment trapping from easily measured physical quantities such as depth, area, and volume in a watershed scale. Thorough the case study, it suggests that remediation in Tsengwen Reservoir Watershed are certainly effective and are able to reduce sediment production and soil loss entering a reservoir. 6 REFERENCES

16.00

8.00

Figure 9. Evolution of vegetation coverage in pre-remediation and postremediation for Longjiao subwatshed.

Vegetation coverage

Utilizing NDVI, multi-stage vegetation recovery of the overall Longjiao subwatershed after remediation was assessed for the five events. According to Eq. (2) and (3), the ratios of vegetation coverage were calculated. Figure 9 shows that after remediation, typhoon Fanapi and typhoon Namodol repeatedly affected Longjiao subwatershed but the vegetation coverage ratio (VR) still remained over 80%. This value was estimated by satellite images and is better than the ratio after typhoon Morakot. Again, these results show that remediation including check dams, river bed foundation, and revetment as well as excavation of deposited sediments, can effectively reduce sediment yield.

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Soil and Water Conservation Bureau (SWCB), 2010. Multi-scale environmental measurements and evaluation of conservation and management efficiency for Tsengwen reservoir watershed (in Chinese). Lo, W.-C., Lin, Bor-Shiun., Ho, H.-C., Keck Jeff k, Yin, H.-Y. and H.Y., Shan 2012. “A simple and feasible process for using multi-stage high-precision DTMs, field surveys and rainfall data to study debris-flow occurrence factors of Shenmu area, Taiwan”, Natural Hazards and Earth System Sciences, 12, 3407-3419. Lin, Bor-Shiun, Cheng-Yang Hsiao,Wai-Yi Leung and Shu-Yeong Chi 2012. “Using Airborne LiDAR Technology to Analyze Landslide Hazards in Shih-men Watershed”, European Geosciences Union, General Assembly 2012, 14, EGU2012-2884, Vienna, Austria, 22~27, April. Schumm, S. 1956. “Evolution of Drainage Systems and Slopes in Badland at Perth Amboy”, Bulletin of Geological Society of America, 67, 597-646. Kriegler, F.J., Malila, W.A., Nalepka, R.F., and Richardson, W., 1969. “Preprocessing Transformations and Their Effects on Multispectral Recognition”, Proceedings of the Sixth International Symposium on Remote Sensing of Environment, 97-131. Sophie L., Tim W. E., Peter B. H., and David J. T. 2008. “Sediment trapping by a tree belt: processes and consequences for sediment delivery, Hydrological Processes, 22(17), 3523-3534.

Practice and development of the piezocone penetration test (CPTu) in geotechnical engineering of China La pratique et le développement de l'essai de pénétration au piézocône (CPTu) en Chine Liu S., Cai G., Du Y.

Institute of Geotechnical Engineering,Southeast University, Nanjing 210096, China

Puppala A.J.

Department of Civil Engineering, The University of Texas at Arlington, Arlington, Texas

ABSTRACT: The cone penetration test (CPT) technique is widely used in field site investigation due to its fast, repeatable, and costeffective advantages. It can provide near-continuous information of soil properties and has a strong theoretical background. In this paper, the history and current development status of CPT, particular the cone penetration test with pore pressure measurement (CPTu) in China practice is systematically introduced. The relationship between international standardized CPTu and China CPT is proposed based on a great number of soils. The paper then presents the review and comparison of the soil characterization methods based on CPTu tests results in China, including stress history, deformation, consolidation and permeability characteristics. RÉSUMÉ: L’essai de pénétration au cône (CPT) est largement utilisé dans les enquêtes de terrain pour ses avantages rapides, reproductibles et rentables. Il peut fournir des informations quasi continues des propriétés du sol et il a une solide théorie. Dans cet article, l’histoire et l’état et développement du CPT, notamment l’essai de pénétration au piézocône (CPTu) qui peut mesurer la pression d’eau interstitielle est systématiquement introduit en Chine. La relation entre les résultats normalisée interationale CPTu et des CPTs en Chine est proposée selon beaucoup des données. Par la suite, le document présente l’examen et la comparaison des méthodes de caractérisation des sols par les résultats des CPTUs en Chine, y compris l’historique des contraintes, des déformations, des caractéristiques de perméabilité et de consolidation. KEYWORDS: Site investigation, CPT, CPTu, Engineering Characterization 1INTRODUCTION The cone penetration test (CPT) has been used for decades to investigate the properties of soil in situ. Essentially, the test consists of pushing a penetrometer with a standard geometry (cylindrical with a diameter of 35.7 mm and a conical point with an apex angle of 60o) into the soil at a rate of 20 mm/s, while measuring a number of parameters. The cone penetration test (CPT) is widely used in-situ testing method, especially in soft soil exploration. As a new kind of in-situ test technique, the piezocone penetration test (CPTu) has been attracting wide attention and widely used in the western developed country. It has been increasingly used because of its important advantages, such as simplicity, speed and continuous profiling. The piezocone, which provides near-continuous measurements of tip resistance, sleeve friction, and pore water pressure induced during the penetration, appears to be a powerful tool for determining the stress history of soft clay deposits. The mechanical CPT like “Dutch” cone was developed by foreign engineers in Shanghai in the early 1930’s (Liu and Wu, 2004). In 1954, the Holland mechanical CPT was first introduced into China. In 1964, the first electric single bridge CPT was independently produced in China with only one measurable parameter (e.g., total specific penetration resistance). Later, the double bridge CPT was developed to measure the tip resistance and sleeve friction independently in the 1970’s, which is currently used in Chinese standards. The Holland CPTu was introduced into China in the early 1990’s by Nanjing Hydraulic Research Institute, but its follow-up development is very slow. The multifunctional CPTu was introduced into China in 2005 by Southeast University researchers. In the following years, the related theory analysis and application practice of CPTu in China have been developed

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rapidly. In this paper, the relationship between international standardized CPTu and China CPT is proposed based on a great number of soils. The paper then presents the review and comparison of the soil characterization methods based on CPTu tests results in China, including determination of stress history, deformation, consolidation and permeability characteristics. 2 COMPARISON BETWEEN INTERNATIONAL STANDARDIZED CPTU AND CHINA CPT Due to the inner geometry of a cone penetrometer, the ambient pore pressure will act on the shoulder area behind the cone. Therefore, the measured cone resistance should be corrected by the unequal area effect for the data presentation and interpretation. In literature works, most of the correlations were developed based on CPT with cone cross area of 10 cm2 as per international standardized CPT and CPTu tests. However, in China, both 15 cm2 and 20 cm2 CPT devices are frequently used. Therefore, the internationalization of Chinese CPT is inevitable (Liu and Wu, 2004). When different sizes of CPT and CPTU are employed, the question of scale effects inevitably arises. For piezocones ranging in area from 5 to 15 cm2, the usual assumption, based on experience summarized by Lunne et al. (1997), is that scale effects are negligible in soil layers of sufficient thickness relative to the cone diameter: that is, quantities such as the cone resistance and excess pore pressure do not depend on the size of the piezocone. Powell and Lunne (2005) compared the results using the 10 cm2 and 15 cm2 piezocones in UK clays. The comparison of various cone sizes and configurations between China CPT and international standard CPTu device at 28 field testing sites is presented. To avoid the variability, all the tests were performed by the same operators. The elevations of the


ground surface at different sites were measured and the difference of elevation may be considered. To quantify the differences between CPT and CPTU measurements, the ratios of the CPTU to CPT cone resistance and sleeve friction measurements were calculated for each site. The reference line positioned at an average CPTU to CPT ratio equal to one in the plots of average ratios represents the theoretical value if soil variability was eliminated and if there was no effect of cone size. In general, the ratios of cone resistance and sleeve friction measurements fluctuate near one, and the measured values increase with depth. For the soft clay sites (Figure 1), the average fs ratio of the friction sleeve is always significantly greater than the average qc ratio. For the topsoil such as fill and silty clay, the ratios CPTU to CPT fluctuate drastically. The relationships of derived key parameters are developed between China CPT and international CPTu (Table 1). From the perspective of engineering practice, it is concluded that qt = 1.03 qc ， fs-CPTu = 1.05 fs-CPT. The empirical equation can be used as useful engineering tool to guide use of 10 cm2 international CPTu in China. 0

Average of CPTU/CPT ratios

0

1

2

3

4

2 4

Cone resistance Sleeve friction

Depth/m

6 8

Table 2. Typical property values of the soils. Site

Soil type

Lianyungang Changzhou

muck clay mucky silty clay

Nanjing

Water content/% 79.6 23.5

Liquid limit 75.6 41.8

Plasticity index 35.8 19.0

41.4

36.4

13.3

Figure 2 shows the relationship between net tip resistance (qt – σvo) and the preconsolidation pressure (σ`p) measured in the laboratory oedometer test on high-quality samples. Here, n as shown in Fig. 1 is the number of data available. In Lianyungang marine clay deposits, the correlation is excellent (r2 = 0.99) for all the data, and the preconsolidation pressure varies between 29 and 139 kPa. In Changzhou clay site, the correlation is good (r2 = 0.95) for the data, and the preconsolidation pressure varies between 812 and 1789 kPa. It can be seen from Figure 5 that the relationship between net tip resistance and the preconsolidation pressure of Nanjing clay site is also pretty good (r2 = 0.98). Consequently, we can obtain the value of Nσt factor, which is equal to 2.7 according to the correlation relationship for Lianyungang lightly overconsolidated clay. Similarly, for Changzhou lightly to moderately overconsolidated clay, the Nσt factor is 2.2. For Nanjing backswamp clay deposit, the N σt factor is 2.5. Consequently, the equation defining the correlation of Lianyungang marine clay site can be expressed as follows:

10

σ 'p 

12

qt  σ v0 2 .7

(1)

14 16

1000

/ k Pa

Figure 1. Statistical analysis of qc and fs ratios

'

p- oed

Table 1. Conversion relationships between CPTu and CPT parameters. Soil types Soft clay Clay Stiff clay Silt Silty sand

Regression equation Cone resistance Sleeve friction qt = 1.04qc fs-CPTU = 1.01 fs-CPT qt = 1.02qc fs-CPTU = 1.07 fs-CPT qt = 1.01qc fs-CPTU = 1.06 fs-CPT qt = 1.03qc fs-CPTU = 1.06 fs-CPT qt = 1.03qc fs-CPTU = 1.03 fs-CPT

Li any ungang（ Ch a n g z h o u （ Na n j i n g （

10

100

2

n = 2 4 , r =0 . 9 9 ） 2

n =2 8 , r =0 . 9 5 ） 2

n =2 0 , r = 0 . 9 8 ）

1000

（

q t - v 0 ） / k Pa

Figure 2. The relationship between preconsolidation pressure measured in oedometer test and net tip resistance

3.2 3 EVALUATION OF ENGINEERING CHARACTERISTICS BASED ON CPTU TESTS IN CHINA 3.1

100

Stress history

Since the advent of CPTU in geotechnics, nearly 20 different methods have been suggested for interpreting the preconsolidation pressure and the overconsolidation ratio of clays (Mayne 1991). In this study, the three sites with all sensitive clay deposits in Jiangsu province of eastern China are selected (Liu et al., 2007). These Quaternary clay deposits are located at Lianyungang, Changzhou and Nanjing respectively. Whenever possible, the OCR values interpreted from various in situ tests were compared with the oedometer values for Lianyungang marine clay. At the other test sites, in addition to oedometer results, some field OCR values were deduced from field performance observation of trial embankments. These field values provide a reliable basis for evaluating the validity of the various interpretation methods in Jiangsu clays. Table 2 presents a summary of the typical property values of the soil layers.

576

Deformation modulus

The one-dimensional constrained modulus, M, as measured in an oedometer test, has been expressed in terms of a coefficient, αm, and cone resistance: M   m qc (2) where αm is a correlation factor. In practice, it has been usual to correlate the modulus M to a penetration resistance. To estimate one-dimensional constrained modulus M, the correlation with net cone resistance (qt -σv0) is used in the form (Kulhawy and Mayne, 1990):

M  8.25  qn  8.25   qt   vo 

(3) Comparisons between M from CPTU with laboratory oedometric modulus for various types of soil proposed by Kulhawy and Mayne (1990) showed that the ratio M -CPTU/ M lab could equal to 2.21 for high-plasticity clays and silty soils. In Figure 3 the constrained modulus estimated with relationship (3) is plotted against that determined by laboratory oedometric tests, carried out on all the types of Jiangsu lagoonal soils. In our case, the ratio M-CPTU/M-lab is always greater than the unity and is not influenced by the type of soil or by its cone resistance value (Cai et al. 2010).


values from laboratory tests and field observations. Schnaid et al. (1996) and Danziger et al. (1997) showed that, when Teh and Houlsby’s approach was employed to interpret various CPTU results, the calculated values of ch were of the same order of magnitude as those measured independently in oedometer tests in the laboratory. Abu-Farsakh and Nazzal (2005) compared seven CPTU methods and showed that Teh and Houlsby (1991) and Teh (1987) methods can estimate ch value better than the other prediction methods. Teh and Houlsby (1991) proposed a relationship between a dimensionless time factor and ch value. The dimensionless time factor, T *, is defined as:

28

Kulhawy and Mayne(1990)

24

M-CPTU/MPa

20 16 12 8

M-CPTU=2.21M-Lab

CH CL ML

2

R =0 . 9 8

4 0 0

4

8

12

16

20

Figure 3. Measured versus predicted constrained modulus values

To examine the possibility for better correlations to estimate the constrained modulus from CPTu data, the corrected cone tip resistance (qt) and the net cone resistance (qt -σvo) were plotted against the laboratory measured constrained modulus as shown in Figures 4(a) and 4(b). A linear correlation was obtained between M and qt as follows: R2 =0.78 (4) M =3.36qt And the following linear correlation was also obtained between M and (qt -σv0) given as follows: R2 =0.92 (5) M =3.73(qt -σv0) The arithmetic mean and standard deviation of (M-CPTU/Mlab) are 0.85 and 0.23 for the first correlation (M = 3.36qt), whereas 1.02 and 0.29 for the second correlation [M=3.73(qt σvo)]. 10

M=3 . 3 6 q t

Me a s u r e d M/ MPa

8

2

R =0 . 7 8

6

0

1

2

3

q t / MPa

(a) 30

Me a s u r e d M/ MPa

w

2.3 v' 0

M=3 . 73 ( q t - v 0 ) 2

R =0 . 9 2

20

15

10

Coefficent of consolidation (10-3cm2/s)

5

0

0 0

2

4

6

8

( q t - v 0 ) / MPa

(b) Figure 4. (a) qt~Measured M ; (b) (qt -σv0) ~Measured M

5

Depth (m)

3.3

(8)

RRch

where RR = compression ratio in the overconsolidated range, and can be obtained from the consolidation tests at the corresponding stress level. A comparison of the consolidation coefficient values measured by CPTu dissipation test and laboratory oedometer test is presented in Figure 5 in which the CPTU test measures ch values, whereas the conventional oedometer test measures cv. It can be seen that the cv values measured by oedometer test are lower than the ch values measured by CPTu tests. The ch values of the lacustrine clay measured by the CPTu tests are generally 4-6 times larger than the cv values measured by the conventional oedometer test, indicating anisotropic characteristic of the soil.

2

25

T50 2 r (7) t 50 where the time factor T50 is related to the location of the filter element and cone size. For a cone with a cross-sectional area of 10 cm2 and with a shoulder filter element, T50 = 0.245 (Teh and Houlsby 1991). The t50 is the measured time for 50% dissipation. The method proposed by Teh and Houlsby (1991) was used here to interpret the coefficient of consolidation for the pore pressure dissipation curves in this study. The coefficient of permeability in the horizontal direction can be estimated from a CPTU dissipation test and by means of the correlation factor (kh/kv) proposed by Jamiolkowski et al. (1985). Baligh and Levadoux (1980) recommended that the horizontal coefficient of permeability could be estimated from the expression: ch 

kh 

4

0

ch t (6) 0.5 r 2 I r  where ch = coefficient of consolidation in horizontal direction; r = radius of cone, typically 17.85 mm; Ir = rigidity index, G/Su. Among the methods available for evaluating ch from piezocone dissipation tests, the one proposed by Teh and Houlsby (1991) is probably most widely used (Robertson et al. 1992). Teh and Houlsby’s (1991) equation is as follows: T* 

28

24

M- l a b /MPa

Consolidation and permeability properties

Many theoretical and semi-empirical methods have been proposed for deriving the coefficient of consolidation from CPTu dissipation data. Teh and Houlsby (1991) proposed a relationship between a dimensionless time factor and ch value based on numerical analysis of dissipation pore pressure with the consideration of soil rigidity index parameter. Robertson et al. (1992) reviewed some dissipation data from piezocone tests, and concluded that the predicted coefficient of consolidation by Teh and Houlsby (1991) solution compared well with reference

577

0

1

2

3

4

5

CPTU Oedometer

10

15

20

25

Figure 5. Comparison of cv and ch profiles measured from CPTu and laboratory tests


The kv values measured by oedometer test and kh values deduced from CPTu test are compared with each other in Figure 6. The values of coefficient of permeability from back-analysis and falling-head permeability tests are also presented in Figure 6. The comparison shows that the kh values measured by oedometer are lower than those obtained from CPTu test within 1-2 order of magnitude. The coefficient of permeability of Taihu lacustrine clay is in the order of 10-5-10-7cm/s. The kh value measured by falling-head permeability tests agrees well with that determined by CPTu tests. 1E-8 0

1E-7

kh(cm/s)

1E-6

1E-5

Depth (m)

5

10

15

CPTU Falling-head tests Asaoka back-analysis Oedometer

20

25

Figure 6. Comparison of kh profiles measured by different methods

4 CONCLUSIONS The comparison of various cone sizes and configurations between China CPT and international standard CPTu device at 28 field testing sites is presented. The relationships of derived key parameters are developed between China CPT and international CPTu. From the perspective of engineering practice, it is concluded that qt = 1.03 qc，fs-CPTU = 1.05 fs-CPT. The empirical equation can be used as useful engineering tool to guide use of 10 cm2 international CPTu in China. The field CPTu tests were carried out in Jiangsu sites to evaluate the stress history. Consequently, we obtained the value of Nσt factor, which is equal to 2.7 according to the correlation relationship for Lianyungang lightly overconsolidated clay. The results show that ratio of M derived from Kulhawy and Mayne 1990’s method to that determined from laboratory oedometric tests, M-CPTU/M -lab, practically equals to 2.21 for high plasticity clays. A quick estimation of the magnitude of coefficient of consolidation ch is proposed by pore pressure dissipation (type u2) tests from the CPTu database. Comparisons of the results obtained by different methods indicate that the values of horizontal coefficient of consolidation determined by CPTu are typically 4 to 6 times those of laboratory tests. The coefficient of permeability values measured by laboratory tests are less than by almost 1-2 orders of magnitude with that determined by CPTu tests in Jiangsu soft clays.

5 ACKNOWLEDGEMENTS The work in this paper was funded by the National Natural Scie nce Foundation (Grant No. 41202203) of China,“Twelfth five-y ear” National Science and Technology Support Plan (Project No . 2012BAJ01 B02) and the Key Project of Natural Science Foun dation (Grant No. BK2010060) of Jiangsu Province of China.

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6 REFERENCES Abu-Farsakh, M.Y. & Nazzal, M.D. 2005. Reliability of piezocone penetration test methods for estimating the coefficient of consolidation of cohesive soils. Journal of the Transportation Research Board. No. 1913, 62-76. Baligh, M. M. & Levedoax, J. N. 1980. Pore pressure dissipation after cone penetration. PhD Thesis, Massachusetts Institute of Technology. Cambridge, Mass, 80-111. Cai, G., Liu, S., Tong, L., 2010. Field evaluation of deformation characteristics of a lacustrine clay deposit using seismic piezocone tests. Engineering Geology, 116(3-4), 251-260. Danziger, F.A.B., Almeida, M.S.S. & Sills, G.C. 1997. The significance of the strain path analysis in the interpretation of piezocone dissipation data. Geotechnique, 47(5): 901-914. Jamiolkowski, M., Ladd, C.C., Germaine, J.T., and Lancellotta, R. 1985. New developments in the field and laboratory testing of soils: State of the art. Proceedings of the 11th International Conference on Soil Mechanics and Foundation Engineering, San Francisco, 1216 August 1985. A.A. Balkema, Rotterdam. Vol. 1, pp. 57-153. Kulhawy, F.H., and Mayne, P.W. 1990. Manual on estimating soil properties for foundation design. Report No. EL-68000, Electric Power Research Institute, EPRI, August 1990. Palo Alto. 306 pp. Liu, S. Y. and Wu, Y. K. 2004. On the state-of-art and development of CPT in China. Chinese Journal of Geotechnical Engineering, 26(4), 553-556. Liu, S. Y. Cai, G. J. Tong, L. Y. Du, G. Y. 2007. On preconsolidation pressure of clays from piezocone tests. Chinese Journal of Geotechnical Engineering, 29(4): 490-495. Lunne, T., Robertson, P.K., Powell, J.J.M., 1997. Cone penetration testing in geotechnical practice, Blackie Academic & Professional, Chapman & Hall, London. Mayne, P.W. 1991. Determination of OCR in clays by piezocone tests using cavity expansion and critical state concepts. Soils and Foundations, 31(2): 65-76. Powell, J. J. M. and Lunne, T. 2005. A comparison of different sized piezocones in UK clay. Proceedings of the 16th International Conference on Soil Mechanics and Geotechnical Engineering, Osaka, 729-734. Robertson, P.K., Sully, J.P., Woeller, D.J. et. al. 1992. Estimating coefficient of consolidation from piezocone tests. Canadian Geotechnical Journal, 29(4):539-550. Schnaid, F., Sills, G.C., Soares, J.M. & Nyirenda, Z. 1996. Predictions of the coefficient of consolidation from piezocone tests. Canadian Geotechnical Journal, 34, No. 2, 315-327. Teh, C.I. 1987. An analytical study of the cone penetration test. PhD Thesis, Dept. of Civil Engineering, Oxford university. Teh, C.I. & Houlsby, G.T. 1991. An analytical study of the cone penetration test in clay. Geotechnique, 41(1): 17-34.

The use of hydro test results for design of steel tanks on stone column improved ground - a case history L’emploi des résultats des essais hydrauliques dans l’étude des réservoirs en acier sur le sol amélioré par colonnes de pierre – histoire de cas Matešić L.

Geokon-Zagreb, Zagreb, Croatia; University of Rijeka, Faculty of Civil Engineering, Rijeka, Croatia

Mihaljević I., Grget G.

Geokon-Zagreb, Zagreb, Croatia

Kvasnička P.

University of Zagreb, Faculty of Mining geology and petroleum engineering, Zagreb, Croatia ABSTRACT: This paper describes hydro tests performed on five large storage tanks (80.000 m3 each) located at the Sisak Oil Terminal, Croatia. Because of its small stiffness and low water permeability, foundation soil for each tank was improved with 660 stone columns. In order to reduce the risk of accidents, such as fire, breach or leak, a crude oil storage tank requires stringent security measures. In the case of the Sisak tanks, the hydro tests were conducted as part of technical monitoring to determine a set of documented and interconnected activities which would provide proof of proper functioning of all elements of a tank structure. In case of critical deviations from the operation expected, such activities ensure that such deviations are removed or corrected on time by taking necessary measures approved by experts. Under a procedure for the hydro tests, the phases of tank and bund filling were defined and each phase was followed by visual inspection and measurements of settlements and deformations of the steel structure. Design directions for future foundation and hydro tests of tanks were made accordingly. RÉSUMÉ : Les essais hydrauliques conduits sur cinq réservoirs de grande taille (chacun de 80,000 m3 ) situés dans le Terminal pétrolier de Sisak en Croatie, sont décrits dans l’ouvrage. Compte tenu de petite rigidité et perméabilité à l’eau peu importante, le sol de fondation pour chaque réservoir a été amélioré avec 660 colonnes de pierre. Les mesures très rigoureuses doivent être prises pour les réservoirs à pétrole brut afin de réduire le risque d’accidents tels que feu, rupture ou fuite de pétrole. Dans le cas des réservoirs de Sisak, les essais hydrauliques ont été conduits dans le cadre de la surveillance technique dont le but était de définir une série des activités bien documentées et interconnectées visées à prouver le fonctionnement impeccable de tous les éléments structurels du réservoir. Dans le cas d’une déviation critique par rapport au fonctionnement normal, ces activités permettent l’élimination ou la correction prompte de ces déviations en prenant les mesures appropriées approuvées par les experts. Dans la procédure pour les essais hydrauliques, les phases de remplissage du réservoir et de la cuvette de rétention ont été définies, et chaque phase a été suivie par une inspection visuelle et par mesurage du tassement et des déformations de la construction en acier. Les instructions d’études sont fournies pour les essais hydrauliques et les essais des fondations futurs. KEYWORDS: steel tank, stone columns, hydro test, monitoring, settlement

1

is 73,2 m and 78,2 m respectively. The tanks have 80.000 m3 in volume and their total and overflow height is 20,6 m and 19,5 m respectively (Figure 1.). Foundation soil is horizontally stratified and, therefore, the soil under all the tanks is of almost the same properties. Because of its small stiffness and low water permeability, the foundation soil for each tank was improved with hundreds of stone columns, which is a technology applied in similar cases of soil improvement (Raju et al 2004, Ambily and Gandhi 2004). In order to prevent industrial accidents, viz. fire, breach or leak of a tank, etc. to happen, a crude oil storage tank requires special safety measures. For this reason, all tank development stages such as ground investigations, design, construction, hydro tests and exploitation, were strictly controlled according to a highly elaborated plan as laid down in API 653 and EN 14015. On the basis of in situ and laboratory tests, a numerical model was created in Plaxis, and all phases of hydrostatic tests were checked before testing.

INTRODUCTION

During the years 2010 and 2011, five new crude oil storage tanks were built at the Sisak Oil Terminal.

A-2510 3 80.000 m

A

A

A-2509 3 80.000 m

A-2508 3 80.000 m

cross section A-A N

A-2511 3 80.000 m 0

bund floating roof tank 50

100

A-2507 3 80.000 m 150

200 m

2

Figure 1. The layout plan of the tanks

DESIGN REQUIREMENTS

In the near vicinity of the new tanks, three 80.000 m3 tanks with floating roofs were installed 30 years ago. During the hydro test performed on one of them, the yielding of foundation soil

All the tanks are of the same size and have an identical steel structure with a floating roof and steel bund wall designed in accordance with API 650. The diameter of the tanks and bunds

579


occurred, which caused tank shell deformations. Such deformations affected normal operation of the floating roof and consequently made the use of the tank impossible. The case described above has not been fully documented, but this is the reason why the investor imposed strict requirements for tank behaviour. Based on the documented cases of soil yielding during tank foundation works (Bell & Iwakiri, 1980) criteria for design, construction and use of tanks have been established. Criteria for maximum total and differential settlement of tanks were determined according to Marr, et al. (1982), API-650 and API-653. Allowable differential settlements are the maximum allowable design limits for deformation of the tank after allowance has been made for construction tolerances. These comprise combinations of: (a) tilt of the tank; (b) tank floor settlement along a radial line from the perimeter to the tank centre; and (c) settlement around the perimeter of the tank. Foundation and foundation soil are subjected to the highest load during hydro tests when a tank is filled with water having a density of 1 t/m3. Later, during tank use, loads on foundations and foundation soil are lower by about 15 % because the tanks are filled with crude oil having a density of 0,85 t/m3. 3

OVERVIEW OF GROUND INVESTIGATION WORKS

At the site, 45 geotechnical boreholes were drilled of which three were 70 m deep. In addition to the boreholes, 18 CPTU tests were also carried out. From the boreholes, undisturbed soil samples were continually taken or SPTs performed. A piezometer was installed in one borehole and a level of ground water monitored over a number of years. Soil classification tests as well as strength, stiffness and water permeability tests were carried out in a laboratory. The investigations showed that the soil is horizontally stratified. 4

DESCRIPTION OF FOUNDATION

The foundation soil was improved with hundreds of stone columns. After the soil had been prepared in this way, the tank shell and bund wall were installed on rigid reinforced concrete ring while the tank bottom was placed directly on the bedding prepared. 4.1

Soil improvement

As the foundation soil is horizontally stratified, the soil under all the tanks has almost the same properties. Because of its small stiffness and low water permeability, the foundation soil for each tank was improved with about 660 stone columns. The depth of the improved soil was approximately 18 m. The spacing between stone columns varied depending on their location on the layout plan. Considering that tank structure is susceptible to planar tilt settlement and non-planar settlement, stone columns were spaced more closely on the perimeter below the foundation ring and centre to achieve stronger effect of improvement. The quality of improvement was checked by CPTU and SASW tests and geodetic surveys carried out in control fields before and after soil improvement. In addition, the data relating to the installation of stone columns were analyzed. Among other things, the volume of the gravel pressed into foundation soil was determined. For each tank, it was found to be about 3% of the volume of the foundation soil improved. As geodetic surveys showed negligible soil upheave (a few millimetres), it can be considered that all the stone pressed into the soil increased directly its density, i.e. soil compaction. 4.2

Concrete ring foundation

The shells of both the tanks and bund walls were mounted directly on a rigid reinforced-concrete foundation ring of

580

rectangular cross-section b/h=350/(260-370) cm, with a central drainage gutter having a width b=60 cm and variable height h=40–80 cm. Outside and inside ring diameters are Dout=79,00 m and Din=72,00 m respectively. 4.3

Bedding of steel tank bottom

The bottoms of the steel tanks were mounted directly on the multi-layer bedding prepared as described below. The foundation soil was levelled and a layer of gravel of grain size 0-64mm was placed. The layer had 70 cm in thickness. To prevent soil pollution in case of tank leak, a HDPE geomembrane was installed in the bedding. The geomembrane was placed between a geosynthetic clay liner and clean sand to protect it from damage. Above the geomembrane, cathodic protection was installed. Additional reinforcement of the soil below the tanks was achieved by placing sand in geocells of 20 cm in height. 5

GEOTECHNICAL DESIGN ANALYSIS

On the basis of in situ and laboratory tests, an axisymmetric numerical model was created in Plaxis 2D-V8. The material behaviour is represented by the Hardening Soil model. In a numerical analysis, soil materials of five types were used and their description and some properties are shown in Table 1. The analysis included tank installation stages, hydro tests and tank exploitation. Table 1. A description of stratified foundation soil

Layer

description Surface layer of stiff clay k=0,0002[m/day]; Eoedref=9,4[MPa]; approx (1) Eurref=30[MPa]; pref=100[kPa]; 0 –6 m=0,409 Layer of soft clay k=0,0002[m/day]; Eoedref=7,8[MPa]; approx (2) Eurref=23,4[MPa]; pref=100[kPa]; 6 –13 m=0,376 Sand with silt and clay k=0,02[m/day]; Eoedref=16[MPa]; Eurref=50[MPa]; pref=100[kPa]; m=0,5 approx (3) Soil below the tanks was improved 13 –20 with stone columns as designed k=1[m/day]; Eoedref=35[MPa]; Eurref=90[MPa]; pref=100[kPa]; m=0 Alternating layers of clay and sand with silt approx (4) k=0,0002[m/day]; Eoedref=10[MPa]; 20 –70 Eurref=40[MPa]; pref=100[kPa]; m=0,376 k - permeability; Eoedref reference edometric modulus at reference stress pref; Eoed =Eoedref(/pref)m edometric modulus; Eurref= unload/reload modulus 6 6.1

depth [m]

HYDRO TEST Introduction

In the case of the Sisak tanks, hydro tests were conducted as part of technical monitoring to determine a set of documented and interconnected activities which would provide proof of proper behaviour of all elements of a tank structure. In case of malfunction or critical deviations from the expected behaviour, such activities would ensure that these deviations are removed or corrected on time by taking necessary measures approved by experts. Under a procedure for a hydro test, the phases of tank and bund filling and emptying were defined; after each phase had


been completed, visual inspection and measurements of settlements and deformations of the steel structure were made. The next phase of the hydro test could begin only after the analysis of the results of measurements obtained from a previous one had been made. Monitoring of each tank settlement involved geodetic surveys and measurement of settlements by a horizontal inclinometer. The geodetic surveys included 24 points on the outside perimeter of the foundation and one point on each of six manholes allowing access to horizontal inclinometer pipes (HI). Settlement measurements were made in three 100-m long pipes running below the tank centre and horizontally declined by 60. The manholes were located eight meters outside of the foundation perimeter. A monitoring programme was planned and carried out in a similar manner as described in the paper by Berardi and Lancellotta (2002).

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Figure 4 Differential settlements at the perimeter of Tank A-2507

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Figure 5 shows total settlements for the three representative phases of the hydro test measured by horizontal inclinometer.

settlement [cm]

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-0.1 hydro test for tank A-2507 26.3.2010 tank=17 m; bund=17m 07.6.2010 tank=0 m; bund=0m planar diff. settl. 26.3.2010 planar diff. settl. 07.06.2010 non-planar diff. settl. 26.3.2010 non-planar diff. settl. 07.06.2010

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/ 10 /10 /10 /10 /10 /10 / 10 /10 /10 /10 / 10 /10 /10 /10 / 10 /02 6/03 3/03 0/03 7/03 3/04 0/04 7/04 4/04 1/05 8/05 5/05 2/05 9/05 5/06 2 2 1 0 0 2 1 0 0 2 2 0 1 27 1

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Figure 4 illustrates the two phases in which the largest differential settlements of tank perimeter at the highest load occurred as well as the phase following tank emptying in which permanent (plastic) deformations occurred. Design allowable differential settlements for the cases of planar tilt settlement and non-planar differential settlement was 44 cm and 0,8 cm respectively. The planar tilt settlement and non-planar differential settlement obtained by calculation were 1,3 cm and 0,34 cm respectively. The results of measurement obtained for such settlements were 1,0 cm and 0,32 cm respectively.

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Figure 3 Total settlements of the tank perimeter occurred during the hydro test on Tank A-2507

Hydro test results

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By way of illustration, the results of settlements obtained from the hydro test carried out on Tank A-2507 are given. Figure 2 shows time history of tank and bund filling and emptying together with the graphs showing averaged settlements of HI pipes at manholes HI, points on the foundation perimeter and centre. 20

hydro test for tank A-2507 date / water level 26.3.2010 tank=17 m; bund=17m 07.04.2010 tank=19 m; bund=0m 07.06.2010 tank=0 m; bund=0 m plane of rigid tilt

0

manhole GR-25

hydro test for tank A-2507 horizontal inclinometers along tank diameter date / water level 26.3.2010 tank=17 m; bund=17m 07.04.2010 tank=19 m; bund=0m 07.06.2010 tank=0 m; bund=0 m

-1

Figure 2 Time history of the hydro test performed on Tank A-2507

According to settlement criteria (Marr et al 1982), the design defined allowable total and differential settlements for different settlement patterns. Thus, during the hydro test, the allowable total settlement of the tank perimeter and tank centre were 15 cm and 31 cm respectively. In a calculation, they were estimated to be 11 cm and 19,5 cm respectively. However, the results of measurement obtained for such settlements as shown on the graph were 3,6 cm and 7,2 cm respectively. Figure 3 shows total settlements of the tank perimeter for the two representative phases of the hydro tests in which the largest settlements occurred at the highest loads. In the third phase, i.e. when the tank was emptied, permanent (plastic) deformations occurred.

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manhole GR-28

tank periphery GR-13

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Figure 5 Total settlements of Tank A-2507 (cross-section)

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hydro test tank periphery - measured A-2507 A-2508 A-2509 A-2510 A-2511

Figure 7. Total settlements of tank perimeters performed on all five tanks settlement [cm]

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Soil improvement described in this paper involved stone columns installed below each tank. The gravel material pressed into the soft soil is equivalent to a layer of about 30 cm in thickness. Since the geodetic surveys showed negligible soil upheaval (a few millimetres), it can be concluded that the soil improvement prevented equivalent settlement of 30 cm. The hydro test showed, as demonstrated in the case of Tank A-2507, that total settlements are relatively small, i.e. smaller than estimated by calculation (Figure 6). For the purpose of comparing actual settlements with those given in behaviour criteria, the settlements of the tank bottoms are shown so that displacements corresponding to a rigid body rotation are given separately from displacements resulting from non-planar differential settlement. Figure 3 illustrates that the bedding was mostly displaced as a rigid body, while non-planar differential settlement was slight. For this reason, it is sure to say that maximum values of the settlements and their shapes are within the values required by the relevant standard (Figure 4).

the soil, it is clearly understandable, yet in some cases disregarded, why such procedures must be applied. The data collected about the behaviour of the tanks during the hydro tests were well documented and could be used to improve design of tanks. 20

CONCLUSIONS

0

7

8

during hydro tests

ACKNOWLEDGEMENTS

The writers particularly wish to thank Z. Korica and M. Bago for their support and assistance. hydro test for tank A-2507 tank center - measured tank periphery - measured tank center - Plaxis - calculated tank periphery - Plaxis - calculated

9

Figure 6. A comparison of measured and calculated results of settlements obtained from the hydro test on Tank A-2507.

It was found that the settlements, after the tank had been emptied, were smaller although they had the same shape. This proves that the deformations after tank emptying are mostly elastic (Figure 4). As tank loads by crude oil are less than those by water, it is expected that subsequent displacements at operating load will be less than those recorded in hydro tests, and that no further non-planar differential settlement of the tank bedding will occur. The same goes for the other four tanks (Figure 7). The diagram of the settlements of all five tanks shows that such settlements are about the same when the tanks are subjected to the same load. As this is normally expected in the case of horizontally stratified soil, this is proof of proper and correct measurement of displacement. As seen in Figure 6, the settlements obtained by calculation were significantly greater than those measured. An explanation for different values of settlements should be thoroughly investigated in further numerical analysis which will take into consideration the fact that columns and soil act together as recommended in Ambily and Gandhi (2004). In the case of the Sisak tanks, the hydro tests proved correct functioning of the floating roofs, watertightness of the shells and bottoms, and rigidity of the foundation structure for all tanks. The strictly applied procedures regarding soil investigation, design, hydro test and exploitation ensured safety in execution and further use. Considering the safety risks and loss of investment in case of non-allowable differential settlements of

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REFERENCES

Ambily, A.P.& Gandhi, S.R. 2004. Analysis of hydro test results for steel tank on stone column improved ground. Proceedings of the Indian Geotechnical Conference held at NIT Warangal, 420-423. API-650: American Petroleum Institute (API) standard 650: ”Welded tanks for oil storage”. API-653: American Petroleum Institute (API) standard 653: “Tank Inspection, Repair, Alteration, and Reconstruction” Bell, R.A. & Iwakiri, J. 1980. Settlement comparison used in tankfailure study. Journal of the Geotechnical Engineering Division 106, GT2, 153-169. Berardi, R. & Lancellotta, R. 2002. Yielding from field behavior and its influence on oil tank settlements. Journal of geotechnical and geoenvironmental engineering 5, 404-415. EN 14015:2004, Specification for the design and manufacture of site built, vertical, cylindrical, flat-bottomed, above ground, welded, steel tanks for the storage of liquids at ambient temperature and above Marr W.A., Ramos J.A. & Lambe T.W. 1982. Criteria for settlement of tanks. Journal of the Geotechnical Engineering Division 108, GT8, 1017-1039. Raju, V.R., Hari Krishna, R. and Wegner, R. 2004. Ground improvement using Vibro Replacement in Asia 1994 to 2004 – a 10 year review, Proceedings of 5th Int. Conf. on Ground Improvement Techniques, Kuala Lumpur, Malaysia.

Interrelationship between deformation moduli from CPTU and SDMT tests for overconsolidated soils La corrélation entre le module de déformation de CPTU et de tests SDMT pour les sols surconsolidés Młynarek Z., Gogolik S.

University of Life Sciences, Poznan, Poland

Sanglerat G.

Ingénieur civil des Ponts et Chaussées, Lyon, France

ABSTRACT: At the area of Poland glaciations caused overconsolidation in deep layers of the subsoil. It is imperative to take into account this fact to calculate the differential settlements of structures subjected to great moments, such as wind turbines. Paper presents the results obtained from the deformation characteristics estimated from CPTU and SDMT tests in clays Vistula and Riss glaciations with interbedded layers of fluvioglacial sands. RÉSUMÉ : En Pologne les glaciers ont provoqué la surconsolidation des couches profondes. Il est impératif de tenir compte de ce fait pour calculer les tassements différentiels des structures soumises à des moments importants telles que les eoliennes. On présente les résultats obtenus à partir des caractéristiques de compressibilité évaluées à partir des essais classiques CPTU et SDMT dans des argiles de Vistula et Riss avec intercalations de couches de sables fluvioglacieres. KEYWORDS: deformation modulus of overconsolidated soils, CPTU, SDMT 1

INTRODUCTION

Determination of representative values of constrained moduli and deformation moduli of soils found in the subsoil is a topical research problem. It is generally known that deformation and strength parameters may be determined using laboratory and in situ tests. In the laboratory method the key element in the evaluation of quality in case of e.g. an oedometric test is connected with the quality of samples collected for analyses (Młynarek 2003, Tanaka 2007). This problem is particularly evident in overconsolidated deposits. This fact indicated that soil deformation parameters need to be determined in situ using DMT, CPTU or SDMT method. Static penetration plays a particularly important role in forecasting values of deformation modulus of soils, as with the use of this method we may obtain a continuous picture of changes in moduli in the subsoil in a 1D or 3-D system (Młynarek et al. 2007). The other testing techniques determine values of moduli pointwise. In CPTU the constrained deformation modulus is determined from correlation relationships. For this reason calibration or assessment of quality of the identification of this modulus using SDMT is of considerable practical importance (Marchetti 1999). This paper discusses this problem together with an assessment of interrelationships between modulus G0 from CPTU and SDMT tests. 2

INVESTIGATIONS OBJECTS

Subsoil structure in Poland is highly complicated in terms of their stratigraphy and lithology. The contact zone of the building structure with the subsoil is comprised primarily of deposits from the two last glaciations, as well as different forms of glacilacustrine deposits.

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Figure. 1 Location of investigated plots in the region of Poland

The subsoil of the investigated area (Fig. 1) comprises glacial tills of the Riss and Vistula glaciations, Quaternary and Pleistocene soils and also Holocene fluvial deposits. The effect of diversification in terms of the genesis and lithology of soils in the discussed locations is shown in CPTU classification systems (Lunne et al. 1997) (Fig.2). In turn, Figure 3 presents examples of geotechnical profiles and results of CPTU and DMT. CPTU tests were performed using a Hyson 200 kN static probe by ap van den Berg, while dilatometer tests were performed with an original seismic dilatometer by Marchetti.


The value of OCR for soils in CPTU testing was determined using the nomogram proposed by Wierzbicki (2010), in which OCR values are established on the basis of cone resistance Qt and the plasticity index of soil Ip is considered. In the case of non-cohesive soils OCR values were also assessed applying a diagram proposed by Wierzbicki (2010). This diagram uses both tests, i.e. CPTU and DMT, as it is constructed on the basis of the formula proposed by Mayne (2000). OCR = 5.04 K01,54

(2)

Figure 2. CPTU Soil classification chart (Lunne et al. 1997).

Figure 4. Relaationship between coefficient OCRDMT/OCRCPTU and coefficient OCRCPTU (group II and III).

Figure 3. Geotechnical profile with CPTU and SDMT characteristics. Figure 5. Relationship between OCRDMT and OCRCPTU after calibration (group II and III)

3 CONCEPT FOR THE IDENTIFICATION OF THE SOIL PRECONSOLIDATION EFFECT IN SUBSOIL Identification of the relationship between the genesis of subsoil and a measure determining the overconsolidation rate, eg. overconsolidation ratio OCR, is a complex problem. Of the two discussed methods, CPTU and SDMT, the chance to determine reliable OCR values is greater for SDMT, since the effect of preconsolidation is strongly related with the geostatic stress ho. For this reason in order to obtain a continuous picture of changes in OCR of the subsoil in the examined locations calibration was performed for OCR values determined using cone resistance Qt, applying OCR values determined by SDMT. In the approach three groups were identified for the discussed locations: with complete drainage – sands (group I), intermediate soils (group II) and clays (group III) (Fig. 2). The groups of intermediate soils and clays were identified based on the content of the clay fraction and the plasticity index Ip. The values of OCR from dilatometer testing for soils of groups II and III were calculated from the relationship (Marchetti (1999): 1,56

OCR = 0.5 (KD) where: KD – horizontal stress index

(1)

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It results from Figs. 4 and 5 that the relationship between OCR values from both tests has a high statistical evaluation. This fact makes it possible to construct a direct dependence between cone resistance Qt and OCR from SDMT (Fig. 6). Values of OCR determined from this dependence were used to supplement data for statistical analysis and next in the profiles at different levels v0, where SDMT testing was not performed.

Figure 6. Relationship between cone resistance qt and OCRDMT coefficient.


4 THE RELATIONSHIP BETWEEN DEFORMATION MODULI AND SHEAR MODULI FROM CPTU AND SDMT 4.1

Constrained deformation modulus – M

Concepts for the determination of modulus M from CPTU and SDMT differ fundamentally. This results from the different techniques of parameter measurements, which are introduced to the relationship between the measured parameter and modulus M. Another factor is connected with the different location of CPTU and SDMT on the characteristics shear modulus 0 – shear strain (Mayne 2000). In the case of CPTU modulus M is determined from the relationship (Sanglerat 1972, Mayne 2000) MCPTU = 1 (qt - v0)

(3)

In DMT the dilatometer modulus ED is the starting point for the determination of modulus MDMT (Marchetti 1999) MDMT = f (ED, KD, ID) (4) where KD – horizontal stress index according to Marchetti (1999) A modified original formula according to Marchetti for the determination of RM for preconsolidated glacial tills was given by Lechowicz et al. (2011). 4.2

Figure 7. Relationship between constrained modulus M0CPTU and M0DMT.

Shear modulus G0

A function which describes the relationship between modulus G0 or G from SDMT or SCPTU and variables which describe parameters of the soil medium was given by Lee and Stokes (1986), Jamiolkowski et al. (1995) G0 = f (’v0, e0, OCR, Sr, C, K, T)

(5)

where: (’v0 – geostatic effective vertical stress, e0 – initial void ratio, OCR – overconsolidation ratio, Sr – degree of saturation, C- grain characteristics, K – soil structure, T- temperature. This relationship may be used to forecast values of modulus G0 directly on the basis of cone resistance Qt. 4.3 4.3.1

Analysis of results Constrained moduli MCPTU, MDMT

To calibrate the relationship between moduli MCPTU and MDMT individual moduli were determined from the following formulas. For the CPTU test according to Mayne (2000) (eq. no. 3). This formula was verified by oedometric tests. The analysis showed that for the tested loams and clays the mean value of coefficient i was close to 8.25. Modulus MCPTU for non-cohesive soils was calculated from dependencies supplied by Lunne et al. (1997) depending on values qc and including the degree of preconsolidation in these deposits. Moduli MDMT were calculated prior to calibration from original formulas proposed by Marchetti et al. (1999).

Figure 8. Relationship between constrained modules M0CPTU and MDMT after calibration.

For preconsolidated deposits the moduli determined by SDMT are higher than those from CPTU (Fig. 7). Obtained relationships fully confirm the opinion by Marchetti et al. (1999) on this subject. Calibration of both moduli in order to describe their changes in the subsoil with changes in v0 is presented in Figs. 8. It was assumed in the calibration process that modulus MCPTU is the reference point. 4.3.2

Shear moduli G0DMT , G0CPTU

The determination of shear modulus from CPTU – G0CPTU was based on empirical dependencies  for non-cohesive soils (group 1) (after Hegazy, Mayne 1995) Vs = 12,02  qt0,319  fs-0,0466

(6)

G0 =   V s 2

(7)



for cohesive soils (groups 2 and 3) relationship determined using multi linear regression

Group 2 G0 = 41,44  qt + 0,31  Gv0 + OCR – 1,71

585

(8)


Group 3 G0 = 41,20  qt + 0,37  Gv0 + 0,88 OCR – 28,53

using Inverse Distance Weighting Method (IDW) according to Młynarek et al. (2007).

(9)

5

CONCLUDING REMARKS

On the basis of the conducted investigations it may be concluded that the simultaneous use of CPTU and SDMT provides a continuous picture of changes in rigidity of subsoil composed of soils with diverse genesis. The effectiveness of these methods is emphasized by the high statistical evaluation for the dependence between deformation and shear strength moduli from both tests. However, to determine this dependence it is necessary to apply a calibration function. The calibration function needs to be specified for the soils, which should be grouped depending on their grain size, since this variable also influences relationships between parameters measured in CPTU and SDMT. After calibration this relationship may be a useful tool in the construction of a model for rigidity of subsoil based on shear strength moduli G0 or M0 moduli.

Figure. 9. Relationship between shear modulus G0DMT and G0CPTU

6

Figure 10. Relationship between measured G0DMT and G0CPTU after the calibration.

Figure 11. The model of subsoil stiffness calculated on the G0 values from CPTU, calibrated by SDMT results.

In order to obtain a continuous picture of changes in the shear modulus G0 the dependence of G0CPTU was calibrated using measured values of G0DMT (Fig. 10). Moduli G0CPTU determined from this relationship may be used in the construction of a model of rigidity for the subsoil composed of soils of varied genesis. An example of such a model for the foundation of a wind turbine is presented in Fig. 11. The model was constructed

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REFERENCES

Hegazy Y.A. & Mayne, P.W. 1995. Statistical correlations between Vs and CPT data for different soil types. Proc. of Symposium on Cone Penetration Testing, Vol. 2:173-178.Swedish Geotechnical Society: Linköping. Jamiolkowski M., Lancellotta R., Lo Presti D.C.F. 1995. Remarks on the stiffness at small strain of six Italian clays. International Symposium on Pre-failure Deformation Characteristics of Geomaterials. Hokkaido vol.2: 817-836. Lechowicz Z., Rabarijoely S., Galas P., Kiziewicz D. 2011. Settlement evaluation of spread foundations on heavily preconsolidated cohesive soils. Annals of Warsaw University of Life Sciences – SGGW, Land Reclam.Nr. 43 (2,) pp.113-120. Lee S. H. H., Stoke K. H. 1986. Investigation of low amplitude shear wave velocity in anisotropics materials. Geotechnical Report No. GR 86-6, Civil Engineering Department, University of Texas, Auatin. Lunne T., Robertson P. K., Powell J.J.M. 1997. Cone Penetration Testing in Geotechnical Practice, Blackie Academic & Professional. Marchetti S., Monaco P., Calabrese M., Totani G. 1999. The flat dilatometer test. A report to the ISSMGE Committee TC-16. Mayne P.W. 2000. Stress-Strain-Strength-Flow Parameters from Enhanced In-Situ Tests. Proc. International Conference on In-Situ Measurement of Soil Properties and Case Histories, Bali, Indonesia: 27-48. Młynarek Z. 2003. Influence of quality of in-situ tests on evaluation of geotechnical parameters of subsoil. Proc. of 13th European Conference on Soil Mechanics and Geotechnical Engineering. Prague. vol.3. Młynarek Z., Wierzbicki J., Wołyński W. 2007. An approach to 3subsoil model based on CPTU results. Proc. of 14th European Conference on Soil Mechanics and Geotechnical Engineering, Madrid. Sanglerat G. 1972. The penetrometer and soil exploration. Elsevier, Amsterdam. Tanaka H., Nishida K. 2007. Suction and shear wave velocity measurements for assessment of sample quality. Proc. of the 3rd International Workshop on in-situ tests and sample disturbance of clays. Studia Geotechica et Mechaica No. 1. Technical University , Wrocław. Wierzbicki J. 2010. Evaluation of subsoil overconsolidation by means of in situ tests at the aspect of its origin. (in Polish). Rozprawy Naukowe nr 410. Wydawnictwo Uniwersytetu Przyrodniczego w Poznaniu. ISSN 1896-1894, 182 pp.

Le Géomécamètre, un nouvel essai in situ adapté à la mesure des caractéristiques hydro-mécaniques du sol The Geomechameter test, a new in-situ apparatus adapted to the measurement of the hydromechanical characteristics of the soil Monnet J.

UJF-Grenoble 1 CNRS UMR 5521, Laboratoire 3SR, Grenoble

RÉSUMÉ : Un nouvel appareil d’essai in situ, le Géomécamètre a été conçu et construit. Cette évolution du pressiomètre utilise les forces produites par un écoulement d'eau vertical descendant autour de la sonde de mesure pour générer un champ de gravité artificiel. Cet écoulement hydraulique permet de régler la contrainte effective verticale au niveau de la sonde de mesure. L'influence de cette contrainte est prise en considération dans l'interprétation des résultats de l’essai. À l'origine conçu pour la caractérisation mécanique des sols et notamment pour la mesure indépendante des caractéristiques de résistance (cohésion c, angle de frottement des caractéristiques de déformation (module d’élasticité E), l'appareil a été modifié pour la mesure de la perméabilité (coefficient k) et de la sensibilité à l’érosion du sol. Cette nouvelle version de l’appareil de 2004 est équipée de la saisie numérique pour la pression, pour le déplacement radial de la membrane et d’une caméra vidéo pour la mesure de turbidité. Un des avantages de cet appareil est la mesure simultanée des principales caractéristiques hydro-mécaniques du sol, notamment dans les digues et des remblais. Le sable fin d’Hostun a été choisi comme matériau de référence pour l'étude expérimentale du modèle réduit testé en laboratoire. Les limons de l’Isère ont été choisis pour expérimenter le prototype de ce nouvel appareil d’essai in situ. Les résultats des mesures au géomécamètre sont validés par la comparaison aux résultats de l’essai triaxial pour les caractéristiques mécaniques et à l’essai Lefranc pour la perméabilité. Cette expérimentation est développée au sein du Projet national Erinoh.

ABSTRACT: A new in situ testing apparatus, the Geomechameter, has been designed and built. It is an evolution of the

pressuremeter, using the forces generated by water flow around the measurement probe. The hydraulic flow allows to control the level of the vertical stress at the test level. The influence of this stress is taken into account in the interpretation of the test results. Originally designed for the soil mechanical shearing resistance (cohesion c and friction angle ), deformation resistance (Young modulus E), the apparatus was modified for the measurement of the permeability (coefficient k) and the sensibility to erosion. This new version of the 2004 apparatus is equipped with numerical gauge for pressure and radial displacement and video camera for turbidity measurement. One of its interests is the simultaneous measurement of the main hydro-mechanical characteristics of the soil inside the soil mass, for the dams and embankments. Hostun thin sand was chosen as a material to undergo the experimental study in laboratory. Isère loam was chosen to experiment the model of this in-situ apparatus. Results of the geomechameter are validated by comparison of mechanical characteristics obtained by the triaxial test and by the Lefranc injection test for the permeability. This experiment is developed with the help of the Erinoh project. MOTS CLEFS : Erosion interne, essais in situ, modélisation numérique, pressiomètre KEYWORDS: Internal Erosion, In-Situ Test, numerical modelling, Pressuremeter 1

INTRODUCTION

L'essai pressiométrique peut être considéré comme un essai de cisaillement entre les contraintes radiales et circonférentielles dans la condition de déformation plane, avec la contrainte verticale qui est la contrainte normale appliquée sur le plan de cisaillement (Baguelin et al., 1978). La limite théorique de l'essai pressiométrique est liée au fait que la contrainte verticale est donnée par le poids des terres au repos. Ce test peut être considéré comme un essai de cisaillement unique et il peut être utilisé pour déterminer soit l'angle de frottement interne (Hughes et al., 1977 ; Monnet, 1990 ; Monnet & Khlif, 1994 ; Monnet, 2012) ou la cohésion du sol ; il peut être utilisé aussi dans la conception des travaux de génie civil (Monnet & Allagnat, 2002). Lorsque la cohésion et l'angle de frottement interne sont déterminés conjointement, ils sont reliés entre eux dans l'interprétation du pressiomètre, si bien que la valeur de la cohésion dépend de l'angle de frottement interne. L’essai au Géomécamètre a plusieurs avantages. - Il contrôle le niveau de contrainte 3D autour de la sonde. Le principe de l'essai au Géomécamètre est de créer un gradient hydraulique (Figure 1) pour contrôler la contrainte verticale en la réglant à une valeur appropriée. Ce nouvel appareil règle la

587

contrainte radiale et circonférentielle par l’expansion de la sonde. Il permet le contrôle de l'état tridimensionnel des contraintes autour de la sonde, à la paroi du forage. - Il permet la mesure de la cohésion et de frottement. Une série de trois essais réalisés avec trois valeurs différentes du gradient hydraulique externe i (Eq.3) permet d’imposer trois contraintes verticales différentes. Ces trois contraintes verticales correspondent à trois courbes différentes d’expansion radiales qui sont autant de courbe de cisaillement différentes. Ceci permet de déterminer cohésion et l'angle de frottement interne. Cette possibilité de mesure locale du couple cohésion-frottement est d'un grand intérêt en Génie Civil. - Il permet la mesure du coefficient de perméabilité et du coefficient de consolidation. Le pressiomètre peut également mesurer la dissipation de la pression interstitielle autour du forage lorsque la consolidation est atteinte dans un délai d'une heure ou plus (Clarke et al., 1979). Le Géomécamètre améliore cette mesure par une détermination simultanée du module de cisaillement et de la perméabilité et n'a pas besoin d'attendre jusqu'à la consolidation finale. - Il permet la mesure du risque de l'érosion interne du sol dans les barrages et les digues. Ce risque aussi appelé suffusion résulte des exfiltrations. La suffusion semble être la principale


cause des incidents hydrauliques observés sur les barrages et les digues. Sur l’ensemble des barrages qui présentent des dommages ou des ruptures, dans 46% des cas on trouve un phénomène d'érosion interne (Foster et al. 2000). En France, 70 cas critiques ont déjà été détecté. Dans cet article, nous présentons la nouvelle version du Géomécamètre qui a été adapté pour la mesure de la suffusion du sol dans les digues. Pour valider les principes théoriques de fonctionnement du Géomécamètre, il est nécessaire d'utiliser un sol standard qui est connu, homogène et dont l'état initial est parfaitement défini. Tous ces paramètres peuvent être mesurés dans une chambre d'étalonnage. L'essai de ce nouveau dispositif est réalisé avec une démarche qualité où chaque étape est soigneusement contrôlée. L'essai en chambre d'étalonnage est la première opération de contrôle qu’il est nécessaire d’effectuer. Nous présentons une première série de test de validation de l’appareil dans des conditions réelles in-situ. 2

de gravité. Cette action peut être considérée comme l'action inverse de l'écoulement de l'eau de boulance obtenu par la relation de Terzaghi (Eq.1) :

 icrit  sat   w   w

(1)

DESCRIPTION DU GEOMECAMETRE

2.1 La sonde du géomécamètre Le prototype de la sonde Géomécamètre est constitué de six modules différents. - Le module 0 relie la tige de forage à la sonde. - Le module 1 : le packer supérieur isole la circulation hydraulique autour de la sonde de la partie supérieure du forage. - Le module 2 : il a deux fonctions différentes, l'injection de l'eau dans le sol et la mesure de la turbidité de l'eau extraite à la base de la sonde. - Le module 3 : c'est la partie centrale de la sonde. Ses fonctions consistent à imposer une pression contrôlée au forage par l’expansion d’une membrane et à mesurer le déplacement radial au niveau de la sonde. - Le module 4 : c'est la cellule de pompage qui sert à extraire l'eau du sol. - Le module 5 : le packer inférieur qui isole la circulation hydraulique autour de la sonde du bas du forage. La sonde doit répondre à plusieurs spécifications : - Il doit être possible de démonter les différentes parties pour les besoins d'entretien de l’appareil (changement de membrane,…). - L'indépendance des quatre circulations liquides doit être assurée, notamment au niveau des liaisons entre les modules (pompage, injection, pression de l'air, la pression de l'eau). - La mesure des pressions, des déformations, des débits, de la turbidité, doit être réalisée numériquement et stockée sur microordinateur. - L’expansion de la sonde, ainsi que la mesure de la déformation doit être possible jusqu’à la pression limite (doublement du rayon du forage).

Figure 1: Le principe de l’essai géomécamétrique

2.2 Evolution de la sonde du géomécamètre Cette nouvelle version du géomécamètre (Figure 2) permet de déterminer le risque de suffusion du sol, par la mesure de la turbidité de l’eau extraite, en utilisant une vidéo caméra embarquée dans le module 2. 3

ÉTUDE THÉORIQUE

3.1 Etude analytique : circulation hydraulique autour de la sonde géomécamètre Le Géomécamètre (Figure1) est un appareil qui utilise le flux hydraulique autour d'une sonde gonflable pour augmenter localement la contrainte effective verticale dans le plan moyen de la sonde. La charge hydraulique diminue le long de la trajectoire lorsque l'eau se déplace dans le sol de la tête jusqu'au pied de la sonde. Une force est appliquée aux particules de sol dans la direction de l'écoulement. Dans l’essai au Géomécamètre, ces forces sont semblables à l'action des forces

588

Figure 2: Le Géomécamètre, version 3, juin 2012

Lorsque le coefficient de perméabilité est supérieur à 10-8 m/s, l'expansion de la sonde n'a aucune influence sur la pression d'eau interstitielle (Cambou et Bahar, 1993 ; Frank et Nahra, 1986) et la pression effective peut être utilisée, mais si le coefficient de perméabilité est inférieur à 10-10 m/s, le test ne peut pas être effectué car le sol devient non drainé. Une modélisation numérique de l'essai (Senouci et Monnet 1999) montre que dans une unité de volume du sol, la force appliquée par le débit hydraulique peut être estimée au moyen de la relation (Eq.2). Dans l’essai au Géomécamètre, l'eau est injectée dans le sol de la cellule d'injection à une pression d'injection (pi). Après circulation dans le sol, l'eau est pompée par la cellule de


cellule de pompage (à gauche en bas). On voit dans le plan médian que la charge hydraulique reste égale constante à -2,6m. Le gradient i (Eq.3) est donc constant dans le plan moyen, l’augmentation de contrainte (Eq.4) est également constante dans le plan moyen.

pompage avec une pression (pp). Le gradient hydraulique (Eq.3) est obtenu par la différence entre la charge hydraulique à l'injection et au pompage. L'augmentation de la contrainte effective verticale (Eq.4) est associée à un écoulement hydraulique de gradient i dans le sol, au niveau de la sonde de longueur le. La contrainte verticale imposée au niveau de la sonde et la profondeur simulée sont ainsi trouvées (Eq.5-6):

   F i . w .V

(2)

 i

(3)

H

i

 H p  le

 v '  ( i  w ) l e 2  v '   ' .Z sim   '  i. w . le 2   vi ' Z sim  (1  i .  w  '). l e 2  Z i

(4) (5) (6)

Figure 5 : Influence du gradient hydraulique sur l’expansion de la sonde du géomécamètre

Figure 3: Comparaison entre les résultats numériques du Géomécamètre pour une profondeur simulée Zsim=2.74m et la courbe expérimentale correspondante avec i = 5.25  = 14,2 kN/m3 ' = 50 kPa)

Figure 6 : Un exemple d’essai au Géomécamètre à 2,2m de profondeur

4

ÉTUDE EXPERIMENTALE

4.1 Essai au laboratoire – Vérification de l’influence du gradient hydraulique sur l’expansion de la sonde Des essais ont été réalisés en chambre de calibration avec le sable d’Hostun au poids volumique de 16kN/m3 (Figure 5). Ils montrent que la courbe d’expansion pour un gradient de 3 est audessus de la courbe sans écoulement. Le gradient hydraulique produit une augmentation de la rigidité apparente du sol qui peut être interprétée comme une augmentation de la contrainte verticale. 4.2 Essai in situ mécaniques

Figure 4 : Variation de la charge hydraulique autour de la sonde du Géomécamètre – résultat Plaxis

-

Détermination

des

caractéristiques

Des essais in situ ont été réalisés sur les digues de l’Isère (Figure 6). Pour l’essai présenté, le débit est d’environ 40 l/min avec 5 h de temps d'essai. Ce temps est nécessaire pour atteindre un débit stabilisé saturé, afin que la contrainte verticale soit modifiée par la sonde du géomécamètre. La simulation de la contrainte verticale est 119kPa pour une contrainte verticale au repos de 55kPa, ce qui correspond à une augmentation de 64kPa. L’essai au géomécamètre permet de mesurer, le module élastique 5MPa, sur le cycle déchargement rechargement. La comparaison avec les courbes d’expansion théoriques permet la détermination de la cohésion 5kPa et de l’angle de frottement 30,5°. Des essais triaxiaux de contrôle ont été effectués sur des échantillons de limon recompactés en conditions drainées à la même densité. Les résultats sont indiqués (Tableau 1). On peut noter pour niveau de contrainte (100kPa) proche de celui

3.2 Etude numérique : Simulation de l’essai par Plaxis La modélisation numérique (Senouci, Monnet 1999) par programme d'éléments finis Plaxis montre que la variation du gradient hydraulique calculé donne une variation de la contrainte verticale (Eq.5) qui permet de définir une profondeur simulée de la sonde (Eq.6). Les différences entre la courbe numérique trouvé par Plaxis à la profondeur simulée de 2,74 m (Eq.6) et la courbe expérimentale pour le gradient hydraulique correspondant (i = 5,25) sont très faibles (Figure 3). La variation de charge hydraulique calculée par Plaxis dans la masse du sol (Figure 4) montre l'augmentation de la charge hydraulique imposée par la cellule d'injection (sur la gauche, à hauteur moyenne) et la décharge hydraulique imposée par la

589


imposée par le géomécamètre, un module de Young qui est très proche de la valeur mesurée in situ avec une différence de 10 %, une cohésion nulle et un angle de frottement 33,5° qui est légèrement plus grand de 3° que celui trouvé in situ. Cette différence peut être expliquée par le fait que l’échantillon triaxial est remanié, reconsolidé, et que cette procédure expérimentale a supprimé la cohésion.

5

Un nouvel appareil de mesure in situ a été construit pour tester le sol dans des conditions de contrainte tridimensionnelle. Le modèle réduit du géomécamètre a montré sa capacité à augmenter la contrainte verticale autour de la sonde pendant l'expansion de la sonde. Ces conditions ont permis de déterminer les caractéristiques de cisaillement des limons de l’Isère en élasticité, cohésion, frottement avec une bonne précision. L’injection de l’eau produite par le géomécamètre a permis de mesurer la perméabilité du limon. La sonde du Géomécamètre sera équipée d'une micro caméra vidéo pour la mesure de la turbidité de l'eau extraite et ainsi détecter le seuil de suffusion à l’intérieur des massifs de sols comme les digues.

4.3 Essais in situ - Détermination de la perméabilité Le Géomécamètre permet de mesurer la perméabilité par la mesure du débit injecté, avec la relation de Ménard (Eq.7) :

k 

 wQ

 1 l    ln 2 l  p  p 0   r0 2

CONCLUSION

(7)

L’essai au géomécamètre a été réalisé le long des berges de l’Isère, dans les limons sablonneux. Le coefficient de perméabilité obtenu est de 1,4 10-6 m/s. Ces résultats peuvent être comparés avec les essais Lefranc (Tableau 2), qui ont été réalisés sur le même site pour le Symbhi. A l’exception du forage P250, la perméabilité moyenne mesurée par l'essai Lefranc est 1,1 10-5 m/s et la perméabilité minimale est de 1,1.10-6 m/s. La perméabilité mesurée par l’essai au géomécamètre se trouve dans l'échelle des valeurs de l'essai Lefranc, mais proche de la valeur minimale mesurée. D'autres expériences doivent être réalisées pour confirmer et expliquer une telle différence. Tableau 1 : Résultats de l’essai triaxial sur un échantillon remanié reconsolidé des Limons de l’Isère E

’3



c’ kPa

MPa

kPa

60 100 200 300

2,7 4,5 10 18,9

Figure 7 : Mesure de la suffusion entre un point d’injection et un point de pompage



degré

6

0,47 0,42 0,3 0,28

0

33,5°

Baguelin F, Jézéquel J.F., Shield D.H., 1978, The pressuremeter and foundation engineering, Series on Rock and Soil Mechanics, Vol. 2, Trans. Tech. Publication, 335-406. Cambou B., Bahar R., 1993, L’utilisation de l’essai pressiométrique pour l’identification de paramétres intrinsèques du comportement du sol, Revue Française Géotechnique (N°63): 39-53. Clarke B.G., Carter J.P., Wroth C.P., 1979, In-situ determination of the consolidation characteristics of saturated clays, Proc. 5th Europ. Conf. SMFE, Brighton, Vol.2, 207-213. Foster M., Fell R., Spannagle M., 2000, The statistics of embankment dam failure and accidents, Canadian Geotechnical Journal, Vol. 37, pp. 1000–1024 Frank R., Nahra R., 1986, Contribution numérique et analytique à l’étude de la consolidation autour du pressiomètre, Rapport recherche LCPC (N°137) Hughes J.M.O., Wroth C.P., Windles D., 1977, Pressuremeter tests in sand, Geotechnique, Jnl 27 (N° 4): 455-477. Monnet J, 1990, Theoretical study of elasto-plastic equilibrium around pressuremeter in sands, Proc. 3rd Int. Symp. Pressuremeter, Oxford, 137-148. Monnet J., Allagnat D. 2002, Design of a large soil retaining structure with pressuremeter analysis, Geotechnical Engineering 155, Issue 1, 71-78. Monnet J., Khlif J, 1994, Etude théorique de l’équilibre élastoplastique d’un sol pulvérulent autour du pressiomètre, Revue Française Géotechnique (N°67): 71-80 Monnet J., 2012, An Elasto-Plastic analysis of the Pressuremeter Test in Granular Soil – part 1: theory , European J. of Environmental and Civil Engineering, Vol.16, N°6, June 2012, 699-71 Senouci S.M., Monnet J, 1999, Modélisation numérique du Géomécamètre, Revue Française de Géotechnique (N°88): 21-35.

Tableau 2 : Résultats de l’essai Lefranc test réalisés sur les digues de l’Isère Profils

Profondeur m

k m/s

P 252 P 252 P 250 P 248 P 248 P 248

2 4 3 2 5,8 7,5

2,8 × 10-5 6.6 × 10-6 1,1 × 10-6 > 10-3 6,6 × 10-6 1,1 × 10-5

RÉFÉRENCES

4.4 Détermination de la suffusion La micro caméra utilisée sur le Géomécamètre a été calibrée avec différentes concentrations de sol érodé. La densité en couleur rouge a été utilisée pour la turbidité de l’eau. Le système a été calibré dans un réservoir rempli d'un mélange de sable et de gravier, et la turbidité de l'eau des effluents a été mesurée. Il semble qu'une suffusion apparaisse pour un gradient hydraulique environ 3. Après une valeur d'un gradient hydraulique de 6, l'analyse de l'image montre que l'eau est claire, mais que parfois certaines particules passent devant la caméra et modifient la valeur de la densité de couleur rouge (Figure 7). Les essais de juin 2012 n’ont pas permis de réaliser correctement la mesure insitu. De nouveaux essais sont programmés pour juin 2013.

590

Analytical approach for determining soil shear strength parameters from CPT and CPTu data Approche analytique pour déterminer la résistance au cisaillement d'un sol sur la base d'essais CPT et CPTu Motaghedi H., Eslami A., Shakeran M.

Dept. of Civil Engineering, Amirkabir University of Technology, Tehran, Iran

ABSTRACT: The common approaches for soil strength parameters determination from CPT data are on the basis of bearing capacity and cavity expansion theories. A new method is proposed for C, ϕ prediction using all quantities, qc, u and fs from CPTu considering bearing capacity mechanism of failure at cone tip and direct shear failure along penetrometer sleeve. One advantage of this method is improvement the accuracy in the case of erroneous data by using all three output of CPTu. Laboratory test results, the two sets of nonlinear equations by the proposed approach and existing correlations of C and ϕ angle parameters have been compared applying on a data base compiled from four sources. It has been considered that the internal friction angle which is obtained by current methods is almost relatively higher than the measured values. Also, the comparison indicates good consistency and low scatter for the proposed method. RÉSUMÉ: Les approches communes pour les paramètres de résistance des sols, déterminés par CPT, sont basées sur la capacité de cisaillement et les théories d’expansion des cavités. Une nouvelle méthode a été proposée pour C,  et utilise toutes les quantités de prévision, qc, u et fs de CPTu, en considérant la capacité de cisaillement et le mécanisme de failure dans type paramide et failure cisaillement direct, le long du pénétromètre manchon (sleeve). Une des avantages de cette méthode est d’améliorer exactitude dans le cas des donnés fausse utilisation, tout les trois sortie de CPTu. Les résultats des essais du laboratoire, les deux combinaisons des équations non linéaires, l’approche proposée et les corrélations existantes de c et l’angle de sont comparées sur quatre bases de données. On considère que la friction interne obtenue par la méthode courante est toujours relativement plus grande que la valeur mesurée, aussi la comparaison montre la bonne consistance et le bas scatter pour la méthode proposée. KEYWORDS: Soil shear strength, Cohesion and friction parameters, CPT and CPTu data, Bearing capacity theory 1

INTRODUCTION

Geotechnical investigation by CPTu provide continuous vertical profile of cone tip resistance (qc), sleeve friction (fs) and pore water pressure (u2) in every inch of the subsoil depth (Lunne et. al, 1997). The CPTu test is used in soft to medium deposits, and not applicable in cemented sand, hard clay and gravelly strata. The penetrometer is a useful tool to identify of thin layers where the traditional sampling procedures cannot be employed. Also, using the CPTu test may distinct the liquefiable or collapsible soil layers around 50 mm thickness in depth (Tavenas and Leroueil, 1987), (Eslami and Fellenius, 2004). In alluvial soils containing gas, determining undrained shear strength by traditional sampling procedures and using UU triaxial tests may lead to conservative results. In granular soils, determining the friction angle (ϕ) as one of the major soil strength parameters by using direct shear or triaxial tests involves uncertainties due to sampling difficulties, confining pressure simulation and limitations of size effects (Mitchell and Durgunoglu, 1983). The main advantage of CPTu versus other in situ test procedures is the relatively elimination of undisturbed sampling, performance in real condition regarding stress level and geological aspects. Furthermore, by using the continuous data in one inch interval of depth, shear strength parameters (C,ϕ), can be obtained which have significant role in geotechnical designs.

penetrometer penetration mechanism, it is assumed that cone tip resistance (qc) is equivalent with ultimate load of a deep circular foundation in subsoil and leads the soil mass to be failed. Whereas, failure assumption in cavity expansion theory is based on required pressure for forming of deep hole in an elasticplastic environment which is fitted with the pressure needed for creation and cavity expansion in the same volume under identical conditions. So far, Muromachi, 1972, Schmertmann, 1978, Mitchell and durgunoglu, 1983, Robertson and Campanella, 1988, Kulhawy and Mayne, 1990 have studied on determination of shear strength parameters from CPT and CPTu data which solely have presented Su in fine grained or ϕ in granular soils. 3 ANALYTICAL MODEL FOR C AND  BY CPTu DATA By applying two basic equations on determination of the deep foundation bearing capacity, one for tip and other for penetrometer sleeve, using the effective bearing capacity instead of total stress approach and extension of the relationships, a dual equation system with two unknowns, can be achieved as below under static loading conditions.

(1)

2 SHEAR STRENGTH PARAMETERS BY CPTU DATA Two main theories have been implemented for the estimation of shear strength parameters by using CPT and CPTu results; bearing capacity (Janbu and Senneset, 1974), (Durgunoglu, 1975) and cavity expansion (Vesic, 1972) approaches. The methods which are based on bearing capacity theories; for

591

Considering deep bearing capacity factors proposed by (Junbu, 1974 base failure model) and applying the analytical Eslami and Fellenius, (1997) model based on CPTu results, the relations can be summarized as follows:


(2)

Eq. 3 is expressed according to empirical results for

at

. Also, Nq can be achieved from Eq. 4 which is shown as below:

(3) (4) Jamiolkowski and Robertson, 1988 presented a correlation for as function of and mean in situ stresses: (5)

Where and are the vertical total and effective stresses, respectively. The lateral stress increases by increasing the relative density. Usually, in calculation, it is assumed that the lateral stress value is equal to resistant horizontal stress by acceptable accuracy as follows: (6)

in Iran. The observation of three boreholes by rotary drilling indicate that the superficial soil layer consists of alluvial gray sea sand with some silt which exists to the depth of 11m. According to USCS this layer is classified as SP, SM or SP-SM. Between depth of 10m to 14m fine clay and silt layer are located in dirty green color with the thickness of 1m to 4m which is classified as CL. The bottom layer is containing fine sea alluvial sand which is observed in depth of 11m to 30m and is classified as SM. Also, the ground water level is located below 3m of ground surface. For determining soil shear strength parameters of filed soil stratification, direct shear, triaxial and uniaxial tests are accomplished on samples. Also, according to SPT records in subsurface depths around 10m, the N values are ranged from 22 to 35, which represent medium to dense relative density for upper layer. The N values in depth of 10m to 14m and 14m to 30m vary from 12 to 25 and 22 to 45, respectively, and classified as dense to high dense coarse grained deposit. The CPTu profile in Sari Narges Hotel site is shown in Fig. 1. Site No. 3, East Changi, (Choa et al. 2004); site is a recovery site which is located in eastern costal of Changi Airport in Singapore. From geotechnical investigations, it is observed that the geomaterial is a kind of soft to medium clay. Site No. 4, University of Texas which is known as A&M Site, (Briaud and Gibbens, 1994). It is one of the international site of study in geotechnical basis and is located in Texas Province, USA. Soil deposits are formed of silty sand.

(7) (8) (9)

By substitution Eqs. 2 to 9 in two basic Eq.1 can be achieved two sets of equation.10 as follow: (10)

Figure 1. CPTu profiles in Narges Hotel Complex, (Sham-e Co., 2012)

4 EXPERIMENTAL RECORDS FOR EVALUATION Geotechnical properties and information including experimental results from the data base of four sites have been compiled. These records are containing 25 series of CPT and CPTu data and shear strength parameters measured by laboratory tests which are used for evaluating developed model. The site specifications are briefly reviewed as follows: Site No. 1, Narenjestan tourism complex, (Mandro Co., 2012); site is located in southern bank of Caspian Sea in Mazandaran Province, Iran. According to borehole operations results, observation and field tests from ground level silty sand with medium dense deposits is located to the depth of 7.5 m. Following the depth of 7.5 m the firm silt layer with high plasticity exist with thickness of 2 m. From depth of 9.5 m down to end of boring poorly graded, silty sand and sand are located with of dense condition and classified as an SM, SP. Site No. 2, Narges Hotel complex, (Sham-e Co., 2012); is located in southern Caspian Sea Shore in the suburb of Sari city

592

The accumulated results of analytical procedure in 25 cases and also laboratory test results are presented in Table 1. Table 1. Shear strength parameters according to proposed method and laboratory test results for 25 measure cases C (kPa) ϕ Site

Soil

qE

fs

Lab

No.

type

(MPa)

(kPa)

test

І І І І І І І І

SM SP SM MH SP SM SM SM

30 13 11 13 5 22 40 28

22 65 50 40 55 110 150 140

4 4.5 4 50 4 4 4 4

proposed

Lab test

3.7 5 3.8 49 5 3.5 4.5 5

30 30 31 4 33 33 35 35

Proposed

32 31 32 6 31 31 36 37


І І � � � � � Ш Ш Ш Ш Ш IV IV IV IV IV

SM SM SM SM SM CL SM CL ML SM CH CL SM SM SM SM SM

30 18 6 5 7 4 4 2 2 2 2 2 1 7 8 6 9

135 60 75 30 80 70 90 14 55 27 56 78 6 30 60 38 75

6 6 3 0 6 29 58 30 28 5 57 35 0 0 0 1 9

5.2 8.4 2.5 1 6.6 30 57 29 29 6.1 56 36 1 1 1.5 1.5 8.4

34 38 32 32 31 2 2 12 14 16 5 8 33 36 32 -

36 37 31 32 32 2.5 2.7 12 15 16 6 9 34 38 33 8 10

condition in cone tip and sleeve has been realized reasonably in proposed relations.

Fig. 2. Comparison between the measured cohesion in laboratory and estimated cohesion by proposed method

5 VALIDATION OF RESULTS AND DISCUSIONS The accomplished geotechnical study in each site has been upon borehole excavations. The samples dependent on soil stratification and from different depths are taken as distributed and undistributed specimens. For determining the shear strength parameters, direct shear, uniaxial and triaxial tests are done on samples in laboratory. Meanwhile, because of high quality of sampling in triaxial test and logicality of the test results in laboratory, it can be more adequate. Four practical cases include CPT and CPTu test results associated with laboratory test results and SPT records are used for evaluating the proposed analytical relations. The measurement results by laboratory tests and also, prediction by using analytical procedure, are presented in Table 1. Evaluation of results expressed the fact that the suggested procedures not only can spontaneously predict and determine both shear strength parameters but also it contain acceptable and reasonable results. Fig. 2 is associated to evaluation and comparison between laboratory results and suggested analytical model for determining the cohesion parameter. The measured and predicted C values show good agreement which denotes the capability of analytical approach. Also, Fig. 3 shows the comparison between measured values and analytical procedure results for internal friction angle within the range of study in four sites. As for the laboratory results which are achieved from drained triaxial test and suggested analytical model, it is observed that the proposed analytical procedures based on CPT and CPTu in cases with cohesion and internal friction angle, almost has identical to laboratory results. The laboratory results are compared with different presented procedures by researchers are shown in Fig. 4a to 4f. According to graphs, the achieved friction angle values by other procedures are always greater than the suggested analytical procedure values and laboratory results. Meanwhile, it is observed that the friction angle values from Meyerhof, (1974) results are closer to bisector line indicating close agreement between the predicted and measured values. Moreover, the presented analytical procedure and laboratory results have more coincidence and are closer to actual values. While, the values obtained from current methods, are more than the experimental results and analytical method. The current procedures do not contain any recommendation for soil cohesion and it is one of the advantages for the proposed procedure. Also, it is not depending only one of the test outputs rather, the entire CPT and CPTu outputs such as qc, fs and u are used in equations, hence the error creation reaches to minimum value in inaccurate records, because of the simultaneous employment of each three output quantities, the other advantages in the presented analytical procedure contrary to traditional procedures. Furthermore, the shear strength parameters derived from actual subsurface failure mechanisms

Fig. 3. Comparison between estimated and measured values for friction angle

Fig. 4. Comparison between estimated and measured values for friction angle

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6 CONCLUSION Geotechnical study by CPT or CPTu can determine continuous vertical profile of cone tip resistance (qc), sleeve friction (fs) and pore water pressure (u) in every inch of the subsoil depth. Hence, the shear strength parameters can be well determined which have major role in geotechnical design. In this study two main theories have been implemented for the estimation of shear strength parameters by using CPT i. e., bearing capacity in cone tip and direct mode of shear failure in along penetrometer jacket. So far, different researchers have studied on determination of shear strength parameters from CPT and CPTu data which solely have presented Su in fine grained or ϕ angle in granular soils. The entire of CPTu data, qc, fs and u are used to calculate C and ϕ, via bearing capacity theory and shear stress relation at failure condition. By combining these relations and applying the proposed analytical Eslami and Fellenius, (1997) model based on CPTu results and direct shear failure along cone sleeve, the drained shear strength parameters values include cohesion and internal friction angle can be derived simultaneously. In proposed procedure the error creation reaches to minimum value through inaccurate records, because of the simultaneous use of each three output quantities. The existence methods for determining the internal friction angle are rely on only one of the test outputs (depending only to qc) while the inaccurate records creates more error in shear strength parameters. But, three parameters qc, fs and u are dependent on friction angle in presented procedure and lead to prorate the error cases. The current procedures do not contain any recommendation for soil cohesion and it is one of the advantages in the proposed procedure. The presented procedure differs from common procedure results by increasing fine grains in soil. Comparison with 25 data sets of C and Ԅ from laboratory tests and predicted by the proposed method indicate good agreement and consistency. 7 REFERENCES Briaud J.L. and Gibbens R.M. 1994. Test and Prediction results for Five large Footing on Sand, FHWA prediction Symp, ASCE Spec, Publ. 41, 255-262. Campanella R.G. Robertson P.K. and Gillespie D. 1983. Cone Penetration testing in deltaic soils. Canadian Geotechnical Journal 20(1) ,23-35. Durgunglu H.T. 1975. Penetration tests of cohesion soils. Proceedings, ASCE, Speciality Conference on In-Situ Measurements of Soil Parameters Eslami A. and Fellenius B.H. 1997. Pile capacity By Direct CPT and CPTu Methods Applied to 102 Case Histories. Canadian Geotechnical Journal 34( 6), 880-898. Eslami, A. and Fellenius, B.H. 2004. CPT and CPTu Data for Soil Profile Interpretation: Review of Methods and a Proposed New Approach. Iranian Journal of Science and Technology, Transaction B 28(1), 6986. Gottardi G. and Tonni, L. 2009. Analysis and interpretation of piezocone data from the Treporti test site for the evaluation of compressibility characteristics of silty soils. DISTART Technical Reort NO. 226, University of Bologna. Jamiolkowski M. and Robertson P.K. 1988. Closing Adress: Future Trends for Penetration Testing. Geotechnology Conference Penetration Testing in UK, Birmingham.321-342. Janbu N. and Senneset K. 1974, Effective stress interpretation of in situ static penetration tests. Proceedings of the European Symposium on Penetration Testing, ESOPT, Stockholm. 22, 81-93. Kulhawy F.H. and Mayne. P.H. 1990. Manual on estimating soil properties for foundation design. Electric Power Research Institute, EPRI. Lunne T. Robertson. P.K. and Powell J.J.M. 1997. Cone penetration testing in geotechnical practice, Blackie Acad. Chapman and Hall/Routledge Press, London, Mandro. Consulting Engineers Final Report. 2012. Site Investigation and Geotechnical Survey for Narenjestan Hotel Babolsar located in Southern Caspian Sea, North of Iran.

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Maple. Maplesoft, A Division of Waterloo Maple Inc, 1981-2010. Mayne P.W. 2007. Cone penetration testing. NCHRP Synthesis 368, Transportation Research Board, National Academies Press, Washington, D.C Mayne P.W. Peuchen J. and Bouwmeester D. 2010. Estimation of soil unit weight from CPT. Proc., 2nd International Symp. On Cone Penetration Testing, CPT'10, Huntington Beach, CA. Meyerhof G.G. 1983. Scale effects of pile capacity. Journal of the Geotechnical Engineering Division, ASCE. 108(GT3). 195-228. Mitchell J.K. and Durgunoglu. H.T. 1983. Cone resistance as measure of sand strength”. Journal of the Geotechnical Engineering Division, ASCE. 104(GT7),. 995-1012. Na Y.M. Choa V. The C.I. and Chang M.F. 2004. Geotechnical parameters of reclaimed sandfill from the cone penetration test. Canadian Geotechnical Journal. 42(1), 91-109. Robertson P.K. 2009. CPT interpretation – a unified approach, Canadian Geotechnical Journal. 49 (11), 1337-1355. Robertson P.K. and Campanella R.G. 1988. Guidelines for geotechnical design using CPT and CPTu. University of British Columbia, Vancouver, Department of Civil Engineering, Soil Mechanics Series 120. Robertson P.K. Woeller D.J. and Finno W.D.L. 1992. Seismic Cone penetration test for evaluating liquefaction Potential under cyclic loading. Canadian Geotechnical Journal. 29(4), 685-95. Senneset K. and Janbu N. 1985. Shear strength parameters obtained from static cone penetration tests. Strength Testing of Marine Sediments; Laboratory and In Situ Measurement. Symposium, San Diego, 1984, ASTM Special technical publication, STP 883, 41-54. Sham-e Consulting Engineers, Final Report. 2012. Site Investigation and Geotechnical Survey for Narges Hotel sari located in Southern Caspian Sea, North of Iran. Tavenas F. and Leroueil S. 1987. State of the art on laboratory and insitu stress- strain-time behavior of soft clay. Proc. Intl. Symp. on Geotechnical Engineering of Soft Soils, Mexico City, 1-146. Vesic A.S. 1972. Expansion of cavities in infinite soil mass. Journal of the soil Mechanics and Foundations Division, ASCE, .265-290.

th


Use of penetration testing for determination of soil properties in earth dam Emploi des essais de pénétration pour déterminer les propriétés de sol pour barrages en terre Mulabdic M.

University in Osijek, Croatia ABSTRACT: The Paper describes a case of a small earth dam for which remediation work was planned, due to bad construction and possible damage to the dam that could have occurred during filling of the retention. In order to assure relevant data for the remediation design solution it was necessary to determine the current state of the co mpacted dam and properties of the clay fill in the dam. Investigation work consisted of drilling boreholes and performing in situ test on the dam, and of laboratory testing of soil samples. CPT and DMT in situ tests were carried out nearby the boreholes on the crest. Potential of these in situ tests in describing physical and mechanical properties of the clay was analysed, since standard methods of interpretation of these tests are based on natural soils, while the dam was constructed by compacting clay. It has been shown that CPT and DMT tests are useful in describing properties of a compacted clay embankment, but also that one should be cautious in using common methods of interpretation of their test results in case of earth fill embankments. RÉSUMÉ : L’article décrit le cas d'un petit barrage en terre, pour lequel une remise en état est planifiée à cause de mauvaise réalisation et du danger potentiel d’endommagement au cours de remplissage de la retenue. Afin d’avoir des données pertinentes pour les techniques de confortement il a été nécessaire de déterminer l’état actuel du barrage et les propriétés de l’argile utilisée dans la construction du barrage. Les travaux de reconnaissance ont compris les forages et les essais in situ sur le barrage, ainsi que les essais en laboratoire. Les essais de pénétration au cône (CPT) et les essais au dilatomètre (DMT) in situ ont été faits auprès des trous de forage dans la crête du barrage. Le potentiel de ces essais dans la description des propriétés physiques et mécaniques d’argile est analysé, étant donné que les interprétations de ces essais sont basées sur les sols naturels tandis que l’argile a été mise en œuvre dans le barrage par compactage. It est démontré que les essais CPT et DMT sont utiles pour l’analyse d’état du sol compacté, mais qu’il faut être très attentif dans l’emploi des procédés standard d’interprétation des résultats de ces essais quand il s’agit des essais pour les ouvrages en remblai. KEYWORDS: earth dam, compacted clay, piezocone test, flat dilatometer test, interpretation MOTS-CLÉS : barrage en terre, argile compactée, essai au piézocône, essai au dilatomètre plat, interprétation 1. INTRODUCTION A small earth dam was built as a part of a future irrigation system. The dam was about 10-meter high at the deepest point in depression, and was constructed of the clay from its vicinity. During the construction of the dam it was noticed that the construction company didn't fully follow the design requirements and criteria related to zoned construction, replacement of foundation soil and degree of compaction of the lifts of clay. During the filling of the lake, when only few meters of dam slopes were covered with water, problems with bottom discharge were observed and filling of water had to be stopped. It was decided that the dam should be checked for safety against sliding and deformability, for which geotechnical properties of compacted clay in the dam should have been should have been checked in detail. The site testing program consisted of drilling boreholes for getting samples for laboratory testing of clay, of penetration testing – CPT and a flat dilatometer test (Marchetti dilatometer – DMT). This paper presents the results of analysis of the properties of clay in the dam based on in situ (CPT and DMT) and laboratory testing. Only boreholes in the crest were used for the analysis, see Fig. 1.

Figure 1. Position of in situ tests and boreholes on the dam; most important work was done on the crest (line B2-C4)

From the samples taken during drilling boreholes specimens were formed for the laboratory testing program, which comprised the testing of physical and mechanical properties of clay from the dam. Fig. 2 shows the plasticity of clay from the dam, determined on samples from the B2 and B5 boreholes.

595



Figure 2. The clay from the dam was of low plasticity; there were zones of silty clay at some depths

It was determined that generally clay compaction degree was under 95 % of Proctor value, that water content was a bit higher than wopt according to Proctor test and that clay of somewhat lower plasticity was used than that which was defined by the design solution. 2. CPT AND DMT TESTS Four CPT test-boreholes and three flat dilatometer (DMT) test-boreholes were realised, along the crest of the dam. General interpretation of test results of these two test types is established for natural soils, and in this case there is compacted clay – human made soil. Therefore it was necessary to check the applicability of standard interpretation methods to compacted clay, for both tests. Both tests were conducted according to relevant standards (EN 1997 – Part 2:2006). Glycerine was used as fluid in porous stone in CPT cone. It should be noted that there is not much experience presented in literature covering CPT and DMT testing in compacted clay. All empirical and theoretical expressions for the interpretation of test results of these two tests are based on natural soils (Larsson and Mulabdic, 1991, Lunne et al, 1996, Marchetti, 1980). 2.1. Soil identification Clay in the embankment was never under water, except for the part deeper than 9 m as measured from the crest. That required careful cone filter saturation with glycerine. CPT soil-type identification was done according to a widely used chart (Robertson, 1990), and in doing so clay of low plasticity was identified in most cases, with some thin layers of silty clay (see plasticity chart in Fig. 2). Pore pressures measured behind the cone (u2) were almost zero, or slightly negative, in all depths.

Figure 3. DMT (M1, M2, M3) and CPT tests (C1, C2), over the dam height (cross-section along the crest). Both, CPT and DMT tests revealed inhomogeinity in the clay embankment – it seems that almost every lift of clay can be spotted over the dam height; DMT test illustrate interpreted versus required Mv, and CPT test interpreted versus required cu

On the other hand, DMT test detected a sandy-silty to silty-sandy soil type, with very rare clayey-silty thin layers. Therefore there were almost no data for undrained strength in DMT interpretation. According to Marchetti (1980), soil type in DMT test interpretation is related to Id = (p1-p0) / (p0-u0), and for clay soil-type it should satisfy 0.1 < Id < 0.6. Since the value of Id in compacted clay of the dam was found to be about or higher than 2 (suggesting a sandy or sandy-silty soil type), and there was no in situ pore pressure in soil, it could be concluded that p0 was too small, due to structure of compacted soil and absence of in situ pore pressures. 2.2. Undrained shear strength by CPT Undrained shear strength from CPT test is calculated according to common expression (Lunne et al, 1997)

su 

qc   v 0 Nk

(1)

Value for Nk=15 was used in this case, which is the mean value of proposed values for natural soils (suggested values are Nk=11-19), and it was confirmed to be applicable for compacted clay as well (Fig. 4).

596


th


would be respected, then bigger portion of the embankment would show lower values then required compared to situation illustrated in Fig 3 (M proj) .

250

Su (kPa)

200 150

Mv (MPa)

0

100

40

80

120

160

200

1

50 0

0

2 0

1000

2000 qT - σvo (kPa)

3000

4000

3 4

Figure 4. Nk=15 Nk = 15 for compacted clays was found to be applicable, based on comparable UU and CPT test results

5

2.3. Modulus of vertical deformation 6

Comparison of modulus of vertical deformation was made for relevant results for oedometer and CPT test. Lunne et al (1997) critically analyse expression for modulus of vertical deformation from CPT test when determined as

 M 8,25( qT   v 0 )

7 8

Mv DMT M3 Mv CPT C2

9

(2)

Mv lab

10 d (m)

In this case it seems that this value should be devided by factor of two (Fig. 5). This might be due to the fact that this general expression has limitations, and becouse oedometer tests were performed on submerged specimens while CPT and DMT tests were performed on clay fill in the embankment that was not submerged. Values of Mv from DMT test were the highest of these three (Fig. 6) (Fig. 6).

Figure 6. Modulus of vertical deformation from oedometer (on submerged specimens) was much smaller than from CPT interpretation (equation (2)) or even lower if compared to DMT standard interpretation values (performed on clay layers that were not submerged)

Fig. 7 presents the sets of CPT and DMT tests with a view to illustrate soil resistance in relation to depth. It seems that the results of tests from different locations are very similar throughout the depth of testing. Mv (MPa) 0

0

100

200

2

2

4

4

6

6

8

8

10

Figure 5. Relationship between laboratory determined modulus of vertical deformation and corrected tip resistance for CPT test, around B5 borehole

0

0

qc [MPa] 10

10 M3

12

M2

12

M1 14 d (m)

Based on a limited number of available test results, the expression Mv=4, 3 (qt-σvo) seems to better fit test results than the equation (2). Modulus seems to be half of the value suggested by that commonly used equation. If relationship between DMT-Mv and LAB Mv from Fig 6

14 d (m)

Figure 7. CPT and DMT tests in cumulative presentation

597

20

C1 C2


th


these two in situ tests was to determine compacted clay condition and its physical and mechanical properties in a continuous profile by depth and at different positions on dam crest. The tests and the interpretation of their results were performed according to accepted standards. Based on analyses of all test results – from in situ and laboratory tests – the following conclusions can be drawn from this case: (1) CPT and DMT detected inhomogeneous clay conditions very clearly along the depth, both in static testing and in seismic testing (SDMT), (2) common interpretation of CPT and DMT test results should be used with caution, allowing for appropriate corrections when tests are performed in compacted clays, since they are developed for natural clays, and here we deal with the compacted – man made soil; (3) it is of importance for the analysis and perception of clay properties whether the embankment is dry or submerged at the time of performing in situ tests; (4) CPT and DMT tests showed remarkable repeatability and proved to be valuable aid in characterizing embankment quality, both in terms of inhomogeinity and physical and mechanical properties; SDMT results also proved to be particularly useful; (5) local correlations between laboratory and in situ test results should always be used, in order to properly account for effects of the presence of water (submerged or non-submerged), specific structure of compacted soil, specific stress distribution and limited experience in using in situ tests for the characterization of compacted soils.

The influence of presence of water on DMT test results interpretation in terms of modulus Mv was discussed in Mulabdic and Bruncic (2000) for natural soils. They concluded that error in water depth assumption had limited influence on interpreted Mv values. Here we are dealing with compacted clay, never being submerged, and obviously soil would be softer if it were submerged. That is, it is difficult to predict soil modulus Mv for the state of a submerged embankment fill from an in situ test performed on a non-submerged embankment fill. Only comparison as shown in Fig. 6 can be used as a guide for correcting in situ evaluated parameters to laboratory values, but even then correction would not be constant with depth. Tests marked as M1, M2 and M3 (DMT-tests) were performed in one run as standard tests and seismic tests (SDMT), using a special seismic probe installed above blade (Cavallaro et al, 2006). Fig. 8 shows wave velocities measured in 0,5-meter depth intervals. Since velocity is a „measure“ of soil structure and its rigidity, variability of those two parameters should be regarded as a basic indication of the variability of soil mechanical properties. These variabilities are more pronounced in wave velocity diagrams than in CPT and DMT standard diagrams. Although velocities generally increase with depth, there are weaker and stronger intervals at certain depths in M2 and M1 boreholes. The M3 location shows constant increase in shear wave velocity by depth. 0

100

v (m/s) 200

300

4. ACKNOWLEDGEMENTS

400

0,0 0,5 1,0 1,5 2,0 2,5 3,0 3,5 4,0 4,5 5,0 5,5 6,0 M1 6,5 M2 7,0 M3 7,5 8,0 d (m) Figure 8. Measured shear wave velocity at different SDMT locations, depth intervals 0,5 m

Cooperation with designing company Elektroprojekt, Zagreb, Croatia, in the planning, execution and analysis of investigation work is highly appreciated. 5. REFERENCES Design solution documentation for Opatovac dam, Elektroprojekt, Zagreb, 2006. EN 1997-2:2006. Eurocode 7: Geotechncial design - Part 2Ground investigation and testing. Larsson, R. and Mulabdić, M. 1991. Piezocone tests in clay. Swedish Geotechnical Institute, Report No. 42, Linköping, pp 240. Lunne, T., Robertson, P.K., Powell, J.J.M. 1997. Cone Penetration Testing in Geotechnical Practice, E & FN Spon, pp 312. Marchetti S. 1980. "In Situ Tests by Flat Dilatometer", ASCE Journal GE, Vol. 106, No. 3, March 1980, pp 299-321. Mulabdić, M. and Brunčić, A. (2000.). Prilog analizi primjene dilatometra Marchetti, Građevinar, Vol 52, No. 1, pp 9-17 (in Croatian). Robertson,P.K. 1990. Soil classification using the cone penetration test. Canadian Geotechnical Journal, 27 (1), 151-8. Cavallaro, A., Grasso, S. and Maugeri, M. 2006. Clay Soil Characterization by the New Seismic Dilatometer Marchetti Test (SDMT), Proc. 2nd international flat dilatometer conference.

3. CONCLUSIONS The paper presented the case of an earth dam of a poor construction quality. In order to characterize clay fill in the embankment in terms of its physical and mechanical properties, CPT and DMT tests were performed in addition to borings and laboratory testing. The purpose of

598

Diagnosis ofof earth-fills and and reliability-based design design Diagnosis earth-fills reliability-based Diagnostic de remblais de terre et conception basée sur la fiabilité Diagnostic de remblais de terre et conception basée sur la fiabilité Nishimura S., Shuku T.

Graduate School of Environmental Science, Okayama University

Suzuki M.

Institute of Technology, Shimizu Co.

ABSTRACT: This research deals with the maintenance strategy of geotechnical structures such as earth-fill dams and river embankments. To determine the soil parameters, the standard penetration test (SPT) N-values are frequently used. Firstly, a statistical model for the N-values is determined from sounding test results. In this research, Swedish Weight Sounding (SWS) tests, simpler than SPT, are employed. Secondly, an indicator simulation is conducted to interpolate the spatial distribution of the N-values, and the results are utilized to find degraded areas inside the embankments and to maintain the embankments. Based on the statistical model for the N-values, the shear strength parameter is derived through the empirical relationships, and then a reliability analysis of the embankments is conducted considering the variability of the internal friction angle. Finally, the effect of improving the embankments is discussed, comparing the calculated risks of the original state with the improved and restored state. RÉSUMÉ : Cette recherche porte sur la stratégie de maintenance des structures géotechniques comme les barrages en remblais de terre et les digues fluviales. Les valeurs N du test de pénétration standard (SPT) sont fréquemment utilisées pour déterminer les paramètres du sol. Premièrement, le modèle statistique de N valeurs est déterminé à partir des résultats des essais de sondage. Dans cette recherche, on utilise le sondage par poids suédois (SWS), plus simple que le SPT. Deuxièmement, une simulation indicatrice est effectuée pour interpoler la distribution spatiale de N valeurs, et les résultats sont utilisés pour trouver les zones dégradées à l'intérieur des remblais, pour l'entretien des remblais. Basés sur le modèle statistique de N valeurs, les paramètres de résistance au cisaillement sont déduits des relations empiriques, ensuite, l'analyse de fiabilité des remblais est effectuée en tenant compte de la variabilité de l'angle de frottement interne. Finalement, l'effet de l'amélioration du remblai est discuté, en comparant l’analyse de risque calculée à partir de l’état initial et de l’état des remblais améliorés et restaurés. KEYWORDS: earth-fill dam reliability-based design, indicator simulation, statistical model of N-value 1

INTRODUCTION

There are many earth-fill dams for farm ponds in Japan. Some of them are getting old and decrepit, and therefore, have weakened. Making a diagnosis of the earth-fills is important for increasing their lifetime, and an investigation of the strength inside the embankments is required for this task. In the present research, firstly, the spatial distribution of the strength parameters of decrepit earth-fills is discussed, and an identification method for the distribution is proposed. Although the strength of earth-fills is generally predicted from the standard penetration test (SPT) N-values, Swedish Weight Sounding (SWS) tests are employed in this research as a simpler method of obtaining the spatial distribution of the N-values. SWS tests are advantageous in that they make short interval exams possible, because of their simplicity. To mitigate disasters, improvement works are conducted on the most decrepit earth-fill dams. Since there is a recent demand for low-cost improvements, the development of a design method for optimum improvement works at a low cost is the final objective of this research. A reliability-based design method is introduced here in response to this demand. Generally, the identification of the spatial correlation of soil parameters is difficult, since the usual sampling intervals are greater than the spatial correlation. Therefore, sounding tests are convenient for determining the correlation lengths. Tang (1979) determined the spatial correlation of a ground by cone penetration tests (CPT). Cafaro and Cherubini (1990) also evaluated the spatial correlation with CPT results. Uzielli, et al. (2005) considered several types of correlation functions for CPT

599

results. Firstly, statistical models for the N-values are determined from the SWS test results. Secondly, the relationship between the SPT and the SWS N-values is modeled, including the transformation error term. The N-value distributions derived from SWS are spatially interpolated with the indicator simulation (Journel and Huijbregts 1978), which is one of the geostatistical methods. The simulated spatial distribution of the N-values can be used for the health monitoring of the inside of an embankment. To evaluate the risk to earth-fill dams, due to the earthquakes, the circular slip surface (CSS) method is used as the stability analysis method along with the soil-water coupling finite element method. The finite element method is used to estimate the normal and the shear stress values on the slip surfaces. In this study, the Monte Carlo method (MCM) is combined with the CSS method to obtain the probability of failure. The procedure for the CSS method, combined with the MCM, has also been conducted by Shinoda, et al. (2006) and Yoshida, et al. (2005). The strength parameter, namely, internal friction angle , derived from SWS tests, is considered to be the probabilistic variable in this research. Additionally, two transformation error terms, namely, the error terms from the SWS N-value to the SPT N-value, and from the N-value to the internal friction angle, are introduced to the MCM. Finally, the risk to an earth-fill dam is calculated from the costs that would be incurred due to embankment failure and probability failure. In this study, the effect of improving an embankment is evaluated as a reduction in risk between the original and the improved states.


INSITU TEST RESULTS

Although high-density sampling is required in order to evaluate the spatial distribution of soil parameters, the amount of data is not sufficient in the general sampling plans. In such cases, sounding is a convenient way to identify the spatial distribution structure of soil parameters. In this research, an embankment at Site H is analyzed, for which SWS tests were conducted at 9 points, at 5-m intervals, along the embankment axis, as shown in Figure 1. The soil profile of the embankment is categorized as intermediate soil. Generally, the strength parameters are assumed based on standard penetration tests (SPT) with the use of empirical relationships. In this research, Swedish weight sounding tests, which are simpler than SPT, are employed instead of SPT. Inada (1960) derived the relationship between the results of SPT and SWS. Equation (1) shows the relationship for sandy grounds, and the relationship is shown in Figure 2.

NSPT  1 0.354

r

 NSWS

(2)

Top of embankment Plan view of test points

x

Figure 1. Plan view of embankment and testing interval.

30 25 20

J

J J

15

J

J

10

JJ

0

50

J

C  Cij 





The random variable function, s(X), is discretized spatially into a random vector s=(s1,s2,...,sM), in which sk is a point estimation value at the location X=(xk, yk, zk). The soil parameters, which are obtained from the tests, are defined here as S=(S1,S2,..., SM). Symbol M signifies the number of test points. Vector S is considered as a realization of the random vector s=(s1,s2,...,sM). If the variables s1, s2,...,sM constitute the M - variate normal distribution, the probability density function of  can then be given by the following equation. C

1 2

 1  t exp s  m C -1 s  m  2 

(4)

in which m=(m1,m2,...,mM) is the mean vector of random function s=(s1,s2,...,sM) and is assumed to be the following regression function. In this research, a 2-D statistical model is considered, namely, the horizontal coordinate x, which is parallel to the embankment axis, and the vertical coordinate z are introduced here, while the other horizontal coordinate y, which is perpendicular to the embankment axis, is disregarded.

k  a0  a1 x k  a2 z k  a3 x k2  a4 zk2  a5 x k z k



(5)

600

J J

J J

J

J J

150

200

250

(a)

 2 exp  x i  x j  lx2  zi  z j  lz2

(3)

J

in which (xk, zk) means the coordinate corresponding to the position of the parameter sk, and a0, a1, a2, a3, a4, and a5 are the regression coefficients. C is the M×M covariance matrix, which is selected from the following four types in this study.

2

 

x

 2 exp

A representative variable for the soil properties, s is defined by Equation (3) equation as a function of the location X=(x, y, z). Variable s is assumed to be expressed as the sum of the mean value m and the random variable U, which is a N(0,1) type normal random variable in this study.

M 2

J

Nsw

Wsw(N)

Determination method

f S s  2 

J

J

J

J

Figure 2. Relationship between SWS results and SPT N-values.

STATISTICAL MODEL OF N-VALUES

sX  mX  U X

J

100

 2 exp  x i  x j lx  zi  z j lz 3.1

J

J J J JJ

J

JJ

J J J J J J J J JJ J J J J JJ JJ J J J J JJ J J J J JJ J J J JJ JJ J JJ J J J J JJ JJJ JJ

5

in which r is an N(0,1) random variable. 3

No.5

No.4

No.3

5m

(1)

in which NSWS is the N-value derived from SWS, NSW is the number of half rations and WSW is the total weight of the loads. Based on this data, the variability of the relationship is evaluated in this study, and the coefficient of variation is determined as 0.354. The determined σ-limits are also shown in Figure 2 with broken lines. Considering the variability of the relationship, the SPT N-values are derived by

No.2

0 250 500 750 1000

NSWS  0.67NSW  0.002WSW

No.1

N

2

2

i



(b)

  x j  lx2  z i  z j  lz2  (c)  2

2



N e 2 exp  x i  x j lx  z i  z j lz



(d)

(6)

i, j  1,2,  , M in which the symbol [Cij] signifies a i-j component of the covariance matrix, is the standard deviation, and lx and lz are the correlation lengths for x and z directions, respectively. Parameter Ne is the nugget effect. The Akaike’s Information Criterion, AIC (Akaike 1974) is defined by Equation (7), considering the logarithmic likelihood.

AIC  2  maxln f S S 2L





 M ln2  min ln C  S  m C1 S  m  2L t

(7)

in which L is the number of unknown parameters included in Equation (4). By minimizing AIC (MAIC), the regression coefficients of the mean function, the number of regression coefficients, the standard deviation, , a type of the covariance function, the nugget effect parameter, and the correlation lengths are determined. 3.2

Determination of statistical model of SWS N-values

The mean function and the covariance function of the SWS Nvalues, NSWS, are determined with MAIC, and the mean and the σ-limits are exhibited in Figure 3. Although the covariance functions given by Equation (6) were examined, the available correlation lengths were not identified. Therefore, additional mean functions are examined. Since the periodic tendency,


whose period is about 10 m along the horizontal axis, is found, the term sin{(x/5-1/2)π} was added to Equation (5). The determined mean function is x 1   m  1.98  0.816 sin    0.157z 5 2  

(8)

The covariance function is determined by





Cij  0.75 exp  x i  x j 6.14  zi  z j 0.63 Cij  1.24

2

i  j 

(9)

i  j 

The horizontal correlation length is identified to be approximately ten times of the vertical one. Since this rate is similar to the values published previously (e.g. Soulie et. al. 1990), the correlation lengths identified here are judged to be appropriate. The boundary between the base ground and the embankment is determined based on the SWS results. The N-distribution predicted based on the determined statistical models with aid of the indicator simulation method (Deutsch and Journel 1990), which is one of the geo-statistical methods, and interpolates the point-estimated N-values, is exhibited in Figure 4. The horizontal periodicity of the Nvalues is presented according to the figure. 4 RELIAIBILITY-BASED DESIGN OF A FILLEMBANKMENT 4.1

Figure 3. Spatial distribution of NSWS and statistical model. 0



2



i  j 

0.5

N1  N SPT  v ' /98

0.5

0 5

5 10

7 9

15 20

i  j 

25

15

Figure 4. Predicted spatial distribution of N-value. Bs

Ac

As Gr

(a) Original embankment.

Rigid soil

Core

(12)

Bs

Ac

Block As

Gr

(13)

The analytical sections of the original embankment, and the improved and restored embankment are exhibited in Figure 5. The embankment is improved by constructing an inclined core, and by covering the original embankment with the additional soil for reinforcement. The material properties are given in Table 1. The soil parameters are determined from the SPT Nvalues and the laboratory soil tests. The Bs means the embankment material; it is determined from the N-values based on the SWS results to consider the spatial distribution. The effective internal friction angle '=d, is obtained from the conversion, namely, Equation (14) (Hatanaka and Uchida 1996). In the equation, 5.3f is the conversion error, in which f is an N(0,1) type normal random variable, and the ratio of 5.3 is the standard deviation.

 ' 20N1   20  5.3 f

40

13

Cij  0.75 exp  y i  y j 6.14  hi  h j 0.63 Cij  1.24

35

11

A stability analysis is conducted and the risk is evaluated for an earth-fill dam at Site H to analyze the transversal section, the mean of the equation. As a mean function, Equation (12) is proposed by averaging Equation (8) along the x –axis, while the covariance function is defined as Equation (13), in which coordinate x is replaced by y of Equation (9), and depth z is replaced by elevation h. This assumption is based on the reason why the embankments are compacted horizontally in the construction, and the correlation structure at the same elevation is homogeneous. 2

Horizontal Coordinate x (m) 10 15 20 25 30

3

Statistical model of an embankment

m  1.89  0.157z

5

1

Depth (m)

2

(14) (15)

in which v' is the effective vertical stress.

(b) Restored embankment. Figure 5. Cross sections and critical slip surfaces of embankments.

4.2

Reliability analysis

In the stability analysis, the pore water pressure is required; it is calculated with a saturated-unsaturated seepage finite element analysis (e.g., Nishigaki 2000). In the restored embankment, the water table level is dramatically reduced by the existence of the impermeable zone. Consequently, this reduction can make the embankment stable. The circular slip surface method is employed as the stability analysis in this study. For uncertain factors, the random numbers are assigned, and the stability of the embankments is evaluated as the probability of failure with the use of the Monte Carlo method. For the reliability analysis, Equation (16) is defined as a performance function, in which the internal friction angle is a probabilistic parameter. As the load of the earthquake, the design earthquake intensity of 0.15 is considered. n

g    fi   si li i1

601

(16)


Table 2. Result of reliability analysis.

Table 1. Parameters of embankment materials.

Fs

Pf

Cf

CF

Original

1.04

0.197

2,000

3,940

Restored

1.62

0

2,000

0

CF = C f × Pf

l

original embankment, and the probability of failure is nearly 20%, which seems very high. For the restored embankment, on the other hand, the probability of failure is nearly zero and the evaluated reduction in risk is drastic, at a value of 39,400,000 JPY. The reduction value means the effect of the improvement work for the embankment.

σn′ τf

τs

5

θ Figure 6. Slip surface across an element

where f and s are the shear strength and the shear force on the slip surface exhibited in Figure 6, which shows a slip surface across a finite element. In the figure, li is the length of the slip surface of element i, and n is the number of elements, which a slip circle crosses. The strength, f is defined by the MohrCoulomb law of Equation (17). Normal stress n and shear force s are defined in Figure 6, and calculated with the soil and water coupling finite element method in this study. In the finite element analysis, the pore pressure is estimated in the saturated zone identified with the saturated - unsaturated seepage analysis, and the negative pore water pressure in the unsaturated zone is disregarded. This assumption can simplify the analysis and make an evaluation for the stability that is on the safe side.

 f  c' n 'tan  '  n '

 ' '   ' 'cos2   z

y

z

y

2 2  z ' y 'sin2   cos2 s  yz 2

yz

sin2

(1) A method to determine the statistical models of the soil strength was presented. The indicator simulation, which is one of the geostatistical methods, was employed. With the proposed procedure, a detailed spatial distribution of the N-values was exhibited. (2) Based on the determined statistical model of the internal friction angle, including the spatial distribution of the N-values, the two conversion errors, from the SWS N-value to the SPT Nvalue, and the SPT N-value to the internal friction angle, the reliability analysis was conducted for an earth-fill embankment, and the probability of failure was evaluated for the original state of the embankment and the restored state of the embankment. By comparing the risks between the original state and the restored one, the effect of the improvement work of the embankment was evaluated 6

(18)

This work was partly supported by JSPS KAKENHI (23248040).

(19)

7

(20)

For the internal friction angle ' of the embankment material Bs, is dealt with as a random variable. Firstly, the random numbers considering the spatial distribution derived from Equations (12) and (13) are assigned to the NSWS. Secondly, the random variable NSPT is evaluated by Equation (2) by considering the conversion error r, and then the ' is obtained with Equation (14), including the conversion error term 5.3f. The Monte Carlo method is iterated 1000 times. 4.3

CONCLUSIONS

(17)

in which c' is the effective cohesion, ' is the effective internal friction angle, z' and y' are the vertical and the horizontal stresses, yz is the shear stress, and  is the angle between a horizontal plane and a slip surface. The probability failure is evaluated with Equation (20) through the use of the Monte Carlo method. Pf  Pr obabilityg  0

Unit (10000 JPY)

Risk evaluation

Two cases of the original embankment and the restored one are compared, whose cross sections are shown in Figures 5(a) and (b). In the figures, the representative slip surfaces, which give the minimum safety factors, are exhibited. In Table 2, the results of the reliability analysis are shown, in which Fs is the average factor of safety, Pf is the probability of failure, Cf is the failure cost, including the damage to houses, agricultural facilities, and farm lands, and CF is the value of the expected failure cost. The average factor of safety is almost 1.0 for the

602

ACKNOWLEDGEMENTS

REFERENCES

Akaike H. 1974. A new look at the statistical model identification. IEEE Trans. on Automatic Control, AC-19 (6), 716-723. Cafaro F. and Cherubini C. 2002. Large sample spacing in evaluation of vertical strength variability of clayey soil. Journal of Geotechnical and Geoenvironmental Engineering 128 (7), 558-568. Deutsch C.V. and Journel A.G. 1992. Geostatistical Software Library and User’s Guide, Oxford University Press. Inada M. 1960. Usage of Swedish weight sounding results. Tsuchi-toKiso, J. of JSSMGE 8 (1), 13-18 (in Japanese). Journel A.G. and Huijbregts Ch.J. 1978. Mining geostatistics, Academic Press. Hatanaka M. and Uchida A. 1996. Empirical correlation between penetration resistance and internal friction angle of sandy soils. Soils and Foundations 36(4), 1-9. Nishigaki M. 2001. AC-UNSAF3D User's Manual. (in Japanese). Shinoda M., Horii K., Yonezawa T., Tateyama M. and Koseki J. 2006. Reliability-based seismic deformation analysis of reinforced soil slopes. Soils and Foundations 46 (4), 477-490. Soulie P., Montes P. and Silvestri V. 1990. Modelling spatial variability of soil parameters. Canadian Geotechnical Journal 27. 617-630. Tang W.H. 1979. Probabilistic evaluation penetration resistances. Journal of the geotechnical engineering, ASCE, 105(GT10). 11731191. Uzielli M., Vannucchi and Phoon, K. K. 2005. Random field characterization of stress-normalized cone penetration testing parameters. Geotechnique 55(1), 3-20. Yoshida, I., Arakawa, T., Kitazume, T. and OOtsu H. 2005. Study on seismic probabilistic safety assessment of a slope, Journal of geotechnical engineering, JSCE, No.785, 27-37.

th


Correlation between cone penetration rate and measured cone penetration parameters in silty soils Corrélation entre le taux de pénétration d‘un cône et des mesures de paramètres de pénétration au cône dans les sols limoneux. Poulsen R., Nielsen B.N., Ibsen L.B.

Aalborg University, Department of Civil Engineering, Aalborg, Denmark

ABSTRACT: This paper shows, how a change in cone penetration rate affects the cone penetration measurements, hence the cone resistance, pore pressure, and sleeve friction in silty soil. The standard rate of penetration is 20 mm/s, and it is generally accepted that undrained penetration occurs in clay while drained penetration occurs in sand. When lowering the penetration rate, the soil pore water starts to dissipate and a change in the drainage condition is seen. In intermediate soils such as silty soils, the standard cone penetration rate may result in a drainage condition that could be undrained, partially or fully drained. However, lowering the penetration rate in silty soils has a great significance because of the soil permeability, and only a small change in penetration rate will result in changed cone penetration measurements. In this paper, analyses will be done on data from 15 field cone penetration tests with varying penetration rates conducted at a test site where the subsoil primary consists of sandy silt. It is shown how a reduced penetration rate influences the cone penetration measurements e.g. the cone resistance, pore pressure, and sleeve friction. RÉSUMÉ: Dans cet article, on montre comment un changement dans le taux de pénétration d’un cône affecte les mesures de pénétration de cône, d'où la résistance du cône, la pression interstitielle et la friction manche en sol limoneux. Le taux normal de pénétration est de 20 mm/s, et il est généralement admis que la pénétration se produit dans de l'argile non drainée alors que la pénétration se produit dans le sable drainé. Lors de l'abaissement du taux de pénétration, l'eau interstitielle du sol commence à se dissiper et un changement de l'état de drainage est vu. Dans les sols intermédiaires, tels que les sols limoneux, le taux de pénétration de cône standard peut conduire à un drainage des conditions qui pourraient être non drainées, partiellement ou totalement déchargée. Cependant, l'abaissement du taux de pénétration dans les sols limoneux a une grande importance en raison de la perméabilité du sol et seulement un petit changement dans le taux de pénétration se traduira par des mesures de pénétration au cône changé. Dans ce document, les données de 15 essais sur le terrain de pénétration au cône, avec différents taux de pénétration menées sur un site d'essai où le premier sous-sol se compose de limon sableux, sont analysés. L’influence d’une réduction du taux de pénétration sur les mesures de pénétration d’un cône, par exemple la résistance du cône, la pression de pore, et la friction manchon, est démontrée. KEYWORDS: Silt, CPT, penetration rate, cone resistance, pore pressure, sleeve friction, drainage, in situ testing. 1

INTRODUCTION

The Cone Penetration Test (CPT) is an in situ testing method that today’s geotechnical engineers often make use of when determining soil parameters, and classifying soil type. The standard rate of penetration is 20 ± 5 mm/s, (ASTM 2007), and while the cone is pushed into the ground the cone resistance, (qc), pore pressure (u2), sleeve friction (fs), and depth (d) are measured. During the penetration, the pore water starts to dissipate, and the dissipation for sands occurs so quickly that the penetration appears as fully drained, whereas the dissipation happens over time for clays, for which reason the penetration is undrained in clays. For intermediate soil, such as silty soils, the penetration is somewhat in between; that is partially drained. According to several researchers (Silva and Bolton 2005, Lehane et al. 2009, Kim et al. 2008, Schneider et al. 2008, Chung et al. 2006, House et al. 2001), the drainage is dependent on the soil permeability, compressibility and penetration rate. The soil permeability and compressibility are both connected to the soil type. However, the penetration rate is regardless of soil type 20 mm/s. When the penetration rate is lowered, the pore water dissipates (change in drainage condition) which results in an increased cone resistance (Lehane et al. 2009, Kim et al. 2008, Chung et al 2006, House et al. 2001). For this reason, the largest cone resistance that could be obtained corresponds to a fully drained penetration. This effect has been shown by several

6031

researchers (Chung et al. 2006, House et al. 2001 and Randolph and Hope 2004) from laboratory tests in clay. Kim et al. (2008) also conducted laboratory as well as field cone penetration tests in cohesive soil and found that the soil behaves undrained for a penetration rate of 20 mm/s and partially drained for a penetration rate of 0.05 mm/s. According to Poulsen et al. (2011a), the change in penetration rate and hence drainage condition has a greater impact in silty soils where the standard rate of penetration often induces a partially drained penetration. This paper analyses data from 15 field cone penetration tests conducted with a penetration rate varying from 60 to 0.5 mm/s. Only a short description of the method for the cone penetrations tests will be given. The results and the interpretation of how a change in the penetration rate affect the measured parameters, hence the cone resistance, pore pressure, and sleeve friction will be given. 2

DESCRIOTION OF EXEPERIMENTAL PROGRAM

The aim of the research is to examine how a change in the cone penetration rate affects the measured cone penetration parameters when conducting cone penetration tests (CPT). The research was carried out at a test site located in the northern Jutland in Denmark, more specifically at a field near the town Dronninglund.


3.1

Test site soil stratigraphy

The soil stratigraphy was at the test site identified by means of two geotechnical boring results. In addition, soil samples were collected and laboratory tests were executed in order to classify the soil. Both test results show that the subsoil consists of sandy silt with clay stripes from approx. 4.5 to 11.4 m below ground level. Above 4 m, the soil consists of silty sand, and below 11 m the soil consists of clay with sandy silt stripes. In addition, the groundwater was encountered at approximately 0.2-0.6 m below the ground level. Generally, the soil is much layered and inhomogeneous which makes the soil difficult to classify. However, in Poulsen et al. (2012a), the soil was overall classified as sandy silt with clay stripes. 2.2

In Figure 2, the smoothed pore pressure from 4.5 to 11.4 m for the CPTs conducted with a penetration rate of 60 and 0.5 mm/s can be seen.

6 7

North coordinate (m)

In order to examine how a change in the cone penetration rate affects the measurements, various cone penetration tests have been conducted. A total of 15 CPTs with five different penetration rates were conducted; these were 60, 20, 5, 1, and 0.5 mm/s. All CPTs were conducted with a distance of approximately 3 m. This was done to make sure that the drainage of each CPT would not affect the drainage of the other CPTs. The location of the 15 CPT’s can be seen in Figure 1.

6336540

6336530 577640

577670

Figure 1 Location of the 15 CPTs with penetration rates of 60, 20, 5, 1 and 0.5 mm/s. The coordinates are given UTM coordinates.

During the execution of the CPT’s, the cone resistance, (qc), pore pressure (u2), sleeve friction (fs), depth (d), and the penetration rate (v) were measured. A more detailed description of the test site, experimental programme and the validity of the tests can be found in Poulsen et al. (2012b). Because of the layered and inhomogeneous soil, the measured CPT parameters are very fluctuating and hence difficult to interpret. As a result, the data has been smoothed for every 50 cm, which was concluded acceptable in Poulsen et al. (2012b). 3

8 9

10 11 -200

0

200

400 u2 (kPa)

600

800

1000

Figure 2. Comparison of the smoothed pore pressure conducted with a penetration rate of 60 and 0.5 mm/s. The figure contains results from 3 CPTs test for each penetration rate. The plotted u0 is an average value.

60mm/s 20mm/s 5mm/s 1mm/s 0.5mm/s Borings

577650 577660 East coordinate(m)

60 mm/s 0.5 mm/s u0

5

Cone Penetration Tests

6336550

Pore pressure

Depth (m)

2.1

EFFCT OF PENETRATION RATE IN SILT LAYER

The soil layer classified as sandy silt with clay stripes was located between 4.5 to 11.4 m below ground level. Only this layer has been analysed since it is considered to be the silt layer where the effect of the penetration rate is clearest. As a result, the following graphs only contain results from 4.5 to 11.4 m. In the following, it is analysed how a change in cone penetration rate affects the measured cone resistance, pore pressure and sleeve friction respectively. As described in Poulsen et al. (2012b), the soil layer consists of many stripes, which gives a very fluctuating result for the measured cone penetration parameters. In order to clearly visualise the effect of a change in the penetration rate, only the penetration rates of 60 and 0.5 mm/s have been included. This is done as it is the extreme points corresponding to undrained and fully drained that are of interest, and the penetration rates of 60 and 0.5mm/s are closest to these conditions. Consequently, the data from the CPTs conducted with a penetration rate of 20, 5 and 1 mm/s have been excluded in the figures.

604

Figure 2 shows that changing the penetration rate from 60 to 0.5 mm/s results in a decreased pore pressure. This is because the drainage conditions change when the penetration rate is decreased. From Figure 2, it seems as though the CPT conducted with a penetration rate of 0.5 mm/s corresponding to fully drained penetration, since the measured pore pressure is close to or equivalent to u0. However, it is not possible to conclude if the CPT conducted with a penetration rate of 60 mm/s corresponds to undrained or partially drained penetration. Nevertheless, by lowering the penetration rate, the penetration changes from undrained or partially drained to fully drained, which results in a lower pore pressure. 3.2

Cone resistance

In Figure 3, the smoothed cone resistance from 4.5 to 11.4 m for the CPTs conducted with a penetration rate of 60 and 0.5 mm/s can be seen. Figure 3 likewise shows that changing the cone penetration rate from 60 to 0.5 mm/s results in a change in the cone resistance. However, a decreased penetration rate results in an increased cone resistance. The changes observed in the cone resistance are like the pore pressure caused by changes in drainage conditions where the penetration changes from undrained or partially drained to fully drained. This results in a higher cone resistance. 3.3

Sleeve friction

In Figure 4, the sleeve friction from 4.5 to 11.4 m for the CPTs conducted with a penetration rate of 60 and 0.5 mm/s can be seen. Contrary to the pore pressure and cone resistance, Figure 4 does not show any correlation between the sleeve friction and cone penetration rate.


140

60 mm/s 0.5 mm/s

5

Mean values 130

6 120

fs (kPa)

Depth (m)

7 8 9

110 100 90

10

80

11 0

5

10 qt (MPa)

15

70 0.1

20

Figure 3 Comparison of the smoothed cone resistance conducted with a penetration rate of 60 and 0.5 mm/s. The figure contains results from 3 CPTs test for each penetration rate.

60 mm/s 0.5 mm/s

300

6

Fitting curve Mean values

250

u2 (kPa)

7 Depth (m)

100

Figure 5. The mean sleeve friction plotted against the mean penetration rate from 4.5 to 11.4 m below ground level. No correlation seems to exist. The standard rate of penetration of 20 mm/s has been marked with a dotted line. 350

5

1 10 Penetration rate (mm/s)

8

200 150

9

100 10

50 11 20

40

60

80 100 fs (kPa)

120

140

160

0 0.1

180

Figure 4. Comparison of the smoothed sleeve friction conducted with a penetration rate of 60 and 0.5 mm/s. The figure contains results from 3 CPTs test for each penetration rate.

4

CORRELATION BETWEEN PENETRATION RATE AND MEASURED PARAMETERS


12

The order of the change in the cone penetrations parameters that can be anticipated is however difficult to read from Figure 2, Figure 3 and Figure 4. As a result, the mean value of the entire silt layer from all CPTs (CPTs with penetration rate of 60, 20, 5, 1 and 0.5 mm/s) can be plotted in a semi logarithmic plot. This has been done for the sleeve friction in Figure 5. Just as Figure 4, Figure 5 does not show any correlation between the mean sleeve friction and the mean penetration rate. According to Lunne et al. (1997) the sleeve friction does not give consistent results during cone penetration. The results shown in Figure 4 and Figure 5 substantiate this, for which reason caution must be taken when using the sleeve friction to analyse CPT data. The mean value for the pore pressure and cone resistance plotted against the mean penetration rate in a semi logarithmic plot is seen in Figure 6 and Figure 7. It can be seen that a correlation between the pore pressure and the penetration rate (Figure 6) and cone resistance and the penetration rate (Figure 7) exist.

605

100

Figure 6. Correlation between the mean pore pressure and the mean penetration rate from 4.5 to 11.4 m below ground level. The standard rate of penetration of 20 mm/s has been marked with a dotted line.

Fitting curve Mean values

10

8 qt (MPa)

0

6

4 2

0 0.1


100

Figure 7. Correlation between the mean cone resistance and the mean penetration rate from 4.5 to 11.4 m below ground level. The standard rate of penetration of 20 mm/s has been marked with a dotted line.


The correlations appear to be linear, however this cannot be true since there must exist an upper and lower boundary for the cone resistance and pore pressure corresponding to fully drained and fully undrained. The mean values can instead be fitted by an expression from Chung et al. (2006), which has been modified by Poulsen et al. (2012b). The expression is given in (1) and (2)

where

��

is the pore pressure (kPa),

(2) is the cone resistance

is the penetration rate (mm/s), is the reference (MPa), penetration rate equal to 20 mm/s and au, bu, cu, mu, aq, bq, cq, and mq are fitting constants. The value of corresponds to undrained penetration, corresponds to a fully drained penetration. whereas From Figure 6 and Figure 7, it is not possible to see when the penetration is undrained or fully drained from the mean values. For this reason, the constants a and b must be assumed. However, in Figure 2 the penetration is close to fully drained for a penetration rate of approximately 0.5 mm/s. This gives an estimate of the constants au + bu and aq + bq. The value of the other constants can be seen in Table 1, and the fitting curves for the pore pressure and cone resistance can be seen in Figure 6 and Figure 7.

6

a

b

c

m

Pore pressure, u

350

-290

1.2

1.1

Cone resistance, q

5.3

3.8

3.1

0.9

By lowering the penetration rate so that the penetration occurs as drained, the cone resistance increases. This can be expressed as (3) (Poulsen et al. 2012b): (3) Where is the cone resistance corresponding to is the measured cone drained penetration (MPa), resistance determined with a penetration rate of 20 mm/s (MPa), and is a coefficient of drainage. The coefficient of drainage, can for the Dronninglund silt be set to 1.0-1.7 depended on is undrained ( =1.7), fully drained whether ( =1.0), or how close to fully drained it is. CONCLUSIONS

This paper has shown how a change in the penetration rate affects the measured cone penetration parameters in silty soil. When using cone penetration tests (CPT) with the standard rate of penetration of 20 mm/s, the penetration will appear as fully drained in sandy soils and undrained in clayed soils. However, for silty soils the standard rate of penetration of 20 mm/s results in a partially drained penetration. In order to examine which affect a changed penetration rate has in silty soils on the measured cone penetration parameters (cone resistance, pore pressure, and sleeve friction), 15 CPTs with varying penetration from 60 to 0.5 mm/s have been conducted. Results from the cone penetration tests conducted with a penetration rate of 60 and 0.5 mm/s were compared for the cone resistance, pore pressure and sleeve friction. It was shown that

606

ACKNOWLEDGEMENTS

The project is funded by DONG Energy and associated with the EUDP programme “Monopile cost reduction and demonstration by joint applied research” funded by the Danish energy sector. The funding is sincerely acknowledged. 7

Table 1. Derived value for fitting constants.

5

both the pore pressure and cone resistance gave different results for a penetration rate of 60 and 0.5 mm/s. The pore pressure measured with a penetration rate of 0.5 mm/s corresponded to drained penetration, which resulted in the highest cone resistance. For the sleeve friction, no correlation was seen. In addition, a correlation between the mean pore pressure and mean penetration rate, and mean cone resistance and mean penetration rate was however seen when plotting the mean penetration rate in a semi logarithmic plot. Compared to the normal penetration rate of 20 mm/s, a decrease in the penetration rate leads to an increase in the cone resistance due to drainage. The increase can be expressed by a coefficient of drainage, , that is equal to 1.0 for fully drained penetration and 1.7 for undrained penetration. The increase depends on whether the normal penetration rate of 20 mm/s has been conducted under undrained, partially drained or fully drained conditions. For that reason, it can be concluded that a correlation between the cone penetration rate and the cone resistance and pore pressure exists. It is an important factor that the cone resistance is dependent on drainage condition and consequently the penetration rate. Particularly if a project requires knowledge of both the undrained soil parameters and the drained soil parameters. In this case, it can be useful to know when the penetration is partially drained and how to convert it to a fully drained penetration or undrained penetration.

REFERENCES

ASTM. 2007. Standard Test Method for Electronic Friction Cone and Piezocone Penetration Testing of Soils. D5778-07, ASTM, Pennsylvania. Chung, S. F., Randolph, M.F., and Schneider, J.A. 2006. Effect of Penetration Rate on Penetrometer Resistance in Clay. J. Geotech. Geoenviron Eng.. 132(9), 1188-1196. House, A. R., Oliveira, J. R. M. S., and Randolph, M.F. 2001. Evaluating the Coefficient of Consolidation using Penetration Tests. Int. J. of Physical Modelling in Geotechnics. 3, 17-26. Kim, K., Prezzi, M., Salgado, R., and Lee, W. 2008. Effect of Penetration Rate on Cone Penetration Resistance in Satuated Clayey Soils. J. Geotech. Geoenviron Eng.. 134(8), 1142-1153. Lehane, B. M., O’Loughlin, C. D., Gaudin, C., and Randolph, M. F. 2009. Rate effects on penetrometer resistance in kaolin. Géotechnique. 41(1), 41-52. Lunne, T., Robertson, P. K., and Powell, J. J. M. 1997. Cone Penetration Testing in Geotechnical Practice. 1st ed., Spon Press, New York, NY, USA Poulsen, R., Nielsen, B. N., Ibsen, L. B. 2011. Effect of Drainage Conditions on Cone Penetration Testing in Silty Soils. Proc. 64th Canadian Geotechnical Conference and 14th Pan-American Conference on Soil Mechanics and Engineering. Toronto, ON, Canada Poulsen, R., Ibsen, L. B., Nielsen, B. N. 2012a. Difficulties Regarding Determination of Plasticity Index of Silty Soils by use of Casagrande and Fall Cone Methods. Proc. of Nordic Geotechnical Meeting. Copenhagen, Denmark Poulsen, R., Nielsen, B. N., Ibsen, L. B. 2012b. Field Test Evaluation of Effect on Cone Resistance Caused by Change in Penetration Rate. Proc. of Nordic Geotechnical Meeting. Copenhagen, Denmark Schneider, J. A., Randolph, M. F., Mayne, P. W., and Ramsey, N. R. 2008. Analysis of Factors Influencing Soil Classification Using Normalized Piezocone Tip Resistance and Pore Pressure Parameters. J. Geotech. Geoenviron Eng.. 134(11), 1569-1576. Silva, M. F., and Bolton, M. D. 2005. Interpretation of centrifuge piezocone tests in dilatants, low plasticity silts. Proc., Int. Conf. on Problematic Soils. Eastern Mediterranean University, Famagusta, N. Cyprus.

Sampling method and pore water pressure measurement in the great depth (-400m) Méthode de mesure de pression interstitielle de l'eau d'échantillonnage en grande profondeur (– 400m) Rito F.

OYO Corporation

Emura T.

Kansai International Airport CO.,LTD

ABSTRACT: Pleistocene clay and sand layers are deposited in the great depth under Holocene soft clay layer at Kansai international airport area. Since the weight of the reclamation soil is heavy because of its depth of sea water which is reached -20m, it has become the very important issue that the characteristics of Pleistocene clays are investigated correctly. For this reason, the new type sampling method which has been called ‘Koken wire line system’ was developed and the undisturbed samples were obtained by this sampling system. Sample quality which obtained from great depth was estimated using the range of the strain which was re-consolidated to insitu effective stress by constant strain rate consolidation test. As a result, it was confirmed that the sample quality of these samples had good quality. On the construction phase of reclamation, cone type measuring equipment of pore pressure for Pleistocene clay and new type measuring equipment of pore pressure for sand were developed and the excess pore water pressure was measured. As a result of the examination of these data, the measured value has been had high accuracy. Therefore, the consolidation characteristic of Pleistocene deposit of Kansai international airport area has been estimated more correct by these useful data.

RÉSUMÉ : L’argile Pléistocène et couches du sable sont déposées dans la grande profondeur sous Holocene couche de l'argile douce à Kansai région aéroportuaire internationale. Depuis le poids du sol de la réclamation est lourd à cause de sa profondeur d'eau de mer qui en est atteinte -20m, il est devenu la question très importante que les caractéristiques d'argiles Pléistocène sont enquêtées sur correctement. Pour cette raison, la nouvelle méthode de l'échantillonnage du type qui a été appelée 'Koken installent le système de la ligne' a été développé et les échantillons non dérangés ont été obtenus par ce système de l'échantillonnage. Goûtez la qualité qui a obtenu de grande profondeur a été estimée utiliser la gamme de la tension qui a été réconsolidée à in-situ stress efficace par épreuve de la consolidation du taux de la tension constante. En conséquence, il a été confirmé que la qualité de l'échantillon de ces échantillons avait la bonne qualité. Sur la phase de la construction de réclamation, matériel de la mesure du type du cône de pression du pore pour argile Pléistocène et nouveau matériel de la mesure du type de pression du pore pour le sable a été développé et la pression de l'eau du pore en excès a été mesurée. Par suite de l'examen de ces données, la valeur mesurée a été eue la haute exactitude. Par conséquent, la caractéristique de la consolidation de dépôt Pléistocène de Kansai que la région aéroportuaire internationale a été estimée plus correct par ces données utiles. KEYWORDS: Pleistocene clay, Koken wire-line system, Pore water pressure measurement. 1

2 SAMPLING FROM GREAT DEPTH USING KOKEN WIRE LINE METHOD

INTRODUCTION

Kansai international airport has constructed in the Osaka bay area. In this area, Pleistocene clay and sand layers are deposited into the great depth under a Holocene soft clay layer. Since the weight of the reclamation soil is heavy because of its depth of sea water which is reached 20m in depth, it has become the very important issue that the characteristics of Pleistocene clays are investigated correctly. Therefore, it has been required high quality sampling and high-precision consolidation test for the samples deposited such great depth of 400m in depth.

Port and Airport Research Institute has tried to improve wire line boring method for the investigation method at the port and harbor area (Matsumoto K., et al.1981). As a result, new wire line method called Koken wire line method has developed. The characteristic of this method is to be able to obtain undisturbed samples which are stiff clay and sand in great depth. Koken wire line method has applied for boring and sampling method of Kansai international reclamation project. The system of Koken wire line method is shown in Figure 1 (Okumura T., et al.1982). Three types of specific samplers have made for Koken wire line method. The structure of these samplers is shown in Figure 2 and Table 1. Thin-walled tube sampler with fixed piston is used for soft and stiff clay whose unconfined compressive strength is under 2MN/m2. Denison

On the construction phase of reclamation, the measurement of pore water pressure for Pleistocene clay and sand layers has become a important issue to improve settlement analysis in addition to the measurement of settlement. As the target depth of the measurement of pore water pressure reaches 300m in depth, the piezometer and the permeability test equipment which are usually used in shallow depth can not use such great depth. The cone type measuring equipment of pore water pressure for Pleistocene clay named GD-CONE and the new type measuring equipment of pore water pressure for sand named H-MHT have been developed. We have been able to measure the pore water pressure in great depth using these new equipment.

sampler which is rotary double-tube sampler is used for more stiff clay whose unconfined compressive strength is over 2MN/m2. Rigid sampler which is double-tube sampler fixed outer tube and inner tube is used for stiff sand and gravel.

607


what kind of method. Therefore, the estimation of sample quality has to be examined quantitatively to interpret the results of constant strain rate consolidation test (CRS test).

The volumetric strain which is consolidated to the in-situ effective stress, εv0, can be used as

①Pump ②Mud screen ③Mixer for mud water ④Boring machine ⑤Tank for clean water ⑥Tank for sludge ⑦Suction tank ⑧Tank for mud water ⑨Winch ⑩Casing pipe (Φ=10”,8”,6”) Sea level

an indication of sample quality. The relationship between change of εv0 and quality of samples having various degree of sample disturbance is shown in Table 2 (Andersen A. and Kolstad P. 1979). The relationship between Δe/e0 and sample quality, where Δe is change in void ratio in recompressing a sample to in-situ effective stress and e0 is initial void ratio, is shown in Table 3 (Lunne T., et al. 1997). According to this figure, when the range of Δe/e0 is within 0.07, the sample can regard good quality. The change of εv0 andΔe/e0 profile of Pleistocene clay samples with recompression to the corresponding in-situ effective vertical stress is shown in Figure 3. With few exceptions, the range of εv0 varies within 2% to 4%. It is apparent that the majority of samples of Pleistocene clay are good quality. The values of Δe/e0 vary in a narrow band of 0.04 to 0.07, and are not sensitive with depth. It may be recalled that if Δe/e0 is within 0.07, the soil samples can be considered to be good quality. The relation between εv0 and OCR, Δe/e0 and OCR of Pleistocene clay samples is shown in Figure 4. It is obvious that OCR is almost constant with depth, and is independent of εv0 and Δe/e0. As the consequence, it is clear that the samples obtained from great depth in the Osaka bay have good and uniform quality.

Sea bottom Boring rod (Φ=135mm) Outer tube Rigid sampler or Denison sampler Wire rope (Φ=9mm) Water pressure Thin-walled tube sampler with fixed piston Sample Sampling tube

Figure 1. System of Koken wire line method. Guide ring Bit

Piston

Outer cube Inner cube

Piston rod

Sampling tube Drain hole

Ball cone clamp Shoulder ring

Wire

Valve for air extraction

Thin-walled tube sampler with fixed piston Cutting wedge Outer tube Sampling tube Ball check

Bit Wing

Metal crown

Bit

Inner tube

Spring

Shoulder ring Latch

Table 2. Relationship between volumetric strain (εv0) and sample quality.

Thrust bearing

εv0(%) ＜1 1～2 2～4 4～10 ＞ 10

Denison sampler Outer tube Inner tube Sampling tube

Shoulder ring Latch assembly Spring

Sample quality Very good Good Fair Poor Very poor

Table 3. Relationship between changing rate of void ratio (Δe/e0) and sample quality.

Ball check Guide ring

Rigid sampler Figure 2. Samplers for Koken wire line method.

Overconsolidation Ratio 1-2 2-4

3 ESTIMATION OF SAMPLESQUALITY OBTAINED FROM GREAT DEPTH

Δe/e0 Good Poor to fair 0.04-0.07 0.07-0.14 0.03-0.05 0.05-0.10

Very good to excellent ＜0.04 ＜0.03

Very poor ＞0.14 ＞0.10

For particular clay multiply Δe/e0 by e0/(l+e0) to get the criteria in terms of εV0

The samples obtained from great depth which is up to 400m in depth are influenced not only mechanical disturbance but also stress release. As mechanical disturbance can avoid improving sampling technic, stress release cannot avoid even Table 1. Specification of samplers. Sampler Applicated soil property

Sampler Name

Sampling tube

Outer Length diameter (mm) (mm)

Remarks

Material

Inner diameter (mm)

Thickness (mm)

Ratio of Angle of inside Length edge diameter (mm) (°) (%)

Soft clay

Thin-walled tube sampler with fixed piston

108

4358

Hydraulic type sampler

Stainless steel (SUS-304)

90

2

6

0

1250

Stiff clay

Thick-walled tube sampler with fixed piston

108

4358

Hydraulic type sampler

Same as the above

81.1

4

6

0

1170

Stiff clay

Denison sampler

108

2850

Projection length of edge blade (20-50mm)

Same as the above

81.1

4

6

0.5

1000

Stiff sand and gravel

Rigid sampler

108

2875

-

Same as the above

90

2

-

-

1000

608

Remarks

Exchange of edge blade is possible Exchange of edge blade is possible -


0

1

εvo（％） 2

3

4

Δe ／e o

5

4 MEASURMENT OF PORE WATER PRESSURE IN GREAT DEPTH

0.00 0.02 0.04 0.06 0.08 0.10 0

6

0

- 50

- 50

4.1 Cone type measuring equipment of pore water pressure for Pleistocene clay (GD-CONE)

Δe / e 0=0.07

The measurement of pore water pressure for clay layer is used a push-in type piezometer (JGS 1313-2003) .This type of piezometer has a merit which can seal the measuring section completely and can measure a correct pore water pressure. Therefore, this is usually used in shallow depth and cannot use such great depth which is up to 350m in depth because of the capacity of sensor and the penetrating power of cone. The cone type measuring equipment of pore water pressure for Pleistocene clay in great depth called GD-CONE has been developed. The structure of this equipment is shown in Figure 5. The characteristics of this cone are as follows: The tip part which is the penetrating part is very thin in order to decrease the penetrating resistance and promote the dissipation of pore water pressure. Its diameter is only 15mm to 20mm. The upper part of the tip becomes thicker gradually. Its diameter is 41mm to 56mm. This part is penetrated into the small borehole, which is drilled in advance, to seal the testing section completely. GD-CONE is connected with AQ rod whose outer diameter is 44.5mm and installed into the borehole. During installation of GD-CONE, the center riser fixed to AQ rod is used in order to install into the pre-drilled small borehole correctly.

εvo=4% - 100

- 150

Dept深度( h ( mm) )

Dept h (( m ) 深度 m)

- 100

- 200

Ｍａ１２Ｍａ１１Ｍａ１０Ｍａ９Ｍａ８Ｍａ７Ｍａ４Ｍａ３

- 250

- 300

- 350

- 150

- 200

- 250

- 300

- 350

Over consol i dat i on rat) i o ( OCR ) 過圧密比( OCR

Figure 3. Result of εv0 and Δe/e0 obtained by CRS test.

2. 5 2. 0


1. 5 1. 0 0. 5

4.2 New type measuring equipment of pore water pressure for sand (H-MHT)

0. 0 0. 0

Over consol過圧密比( i dat i onOCR rat i) o ( OCR )

The pressure gauge of GD-CONE has used a crystal oscillation sensor which has wide pressure range and high sensibility. The maximum pressure range is 5MPa and the sensibility has 0.01%FS. The compensation of atmospheric pressure has been done by using another pressure gauge on the ground.

1. 0

2. 0

3. 0

εvo(%)

4. 0

5. 0

The measurement of pore water pressure for sand layer in great depth is used a new type measuring equipment of pore water pressure called H-MHT. The structure and test procedure of HMHT is shown in Figure 6. The characteristic of H-MHT are as follows: As the principle of measurement is simple, the reliable measurement is possible easily. As the pressure gauge of H-MHT has used a crystal oscillation sensor too, the high pressure caused in the great depth can be measured highly precise.

6. 0

2. 5 2. 0


1. 5 1. 0 0. 5 0. 0 0. 00

0. 02

0. 04

0. 06

Δｅ／ｅo

0. 08

H-MHT can obtain equilibrium water table in a short time because the specific air valve which is joined to the AQ rod can shut the test section in order to promote dissipation of pore water pressure rapidly.

0. 10

Figure 4. Relationship between OCR and εv0, Δe/e0. Metal filter

Crystal oscillation sensor

Tip part

Connection to AQ rod (Outer diameter: 44.5mm, Inner diameter: 34.9mm)

(300mm) Penetrating part

Pore water pressure measuring part

Sealing part Length 2,540mm

Figure 5. Structure of cone type measuring equipment of pore water pressure for Pleistocene clay (GD-CONE).

609


Three-way Pressure regulation ·Distributor Converter

Water table in measuring pipe is fallen by being given gas pressure

Wire line rod

Computer

Converter Computer

Nylon tube for packer pressurization

Gas tube Air valve AQ rod Wire line rod Underwater connector

Packer

Packer (Expanded state) Underwater connector Drainage pipe Crystal oscillation sensor Guide for water flow

Testing section

Testing section

Pressure sensor code

Figure 6. Structure of new type measuring equipment of pore water pressure for sand (H-MHT).

4.3 Results and discussion 4.3.1 Estimation of pore water pressure using hyperbolic method The diameter of the tip part of GD-CONE is designed of thin size because pore water pressure is needed to dissipate rapidly. The standard method to GD-CONE continue to measure for three hours after penetrating. The adoption value of pore water pressure is calculated by hyperbolic method using the data measured after three hours from penetration. In order to accurate the adoption value, the long hour’s measurement, 38 hours, has carried out. The results of these data are shown in Table 4. Since the results of the long hour’s measurement and the adoption value by hyperbolic method using three hours’ measurement are almost same, the accuracy of hyperbolic method has been confirmed. Table 4. Application of hyperbola method for measuring result of pore water pressure.

11260 12110

Pore water pressure of last measuring time (kPa) 2760 2800

Pore water pressure by hyperbola method (kPa) 2700 2702

138917

2700

2702

No.

Altitude (CDL-m)

Investigation case

Measuring hours(sec)

1 2

263.47 263.47

Standard measurement

3

264.02

Long measurement

4.3.2 Reliability of sealing of measurement section of GDCONE Three patterns of penetration of GD-CONE, which is varied from 30cm, 60cm to 90cm length, have carried out. The results are shown in Figure 7. These data are almost same despite the penetrating length. In addition, since the result of pore water pressure is not same to the mud water pressure of the bore hole, the seal of measuring section is regarded completely. 6000

Note 凡例 Pushing length: 30cm 1 回目押し込み結果 Pushing length: 60cm 2 回目押し込み結果 Pushing length: 90cm 3 回目押し込み結果

間隙水圧（kPa） Pore water pressure (kPa)

5000 4000

Mud water pressure=2,960kPa 泥水圧＝2960kPa

3000 2000 1000 0

0

2000

4000

6000

8000

10000 12000 14000 16000 18000 20000

経過時間（sec） Time (sec)

Figure 7. Difference of test result of pore water pressure by penetrating length.

610

4.3.3 Comparison examination of the results by GD-CONE and H-MHT for Pleistocene clay The comparative experiments used both GD-CONE and HMHT has been carried out in same depth, which is 170m in depth. The results are shown in Table 5. Since these data are almost same, the appropriate of GD-CONE and H-MHT method for measuring equipment for great depth can be confirmed. Table 5. Comparison between measuring result of pore water pressure using GD-CONE and H-MHT. Measuring method GD-CONE H-MHT

5

Altitude (CDL-m) 172.77 169.12～172.77

Pore water pressure p (kPa) 2006 1984

Excess pore water pressure Δu (kPa) 256 257

CONCLUSIONS

Kansai international airport which has been constructed in the Osaka bay far from 5km from the land area in order to solve noise pollution has been able to take off and landing of airplane using two runways whose length is about 4,000m. The consolidation settlement of 2nd runway, which is related to this paper, is almost the same like the consolidation analysis. For the future, the management of consolidation settlement shall be important for the operation of the airport while the consolidation settlement will continue for long times. It is important that the continuing study for the settlement of Pleistocene clays using the in-situ observation data. 6

REFFERENCES

Matsumoto K., et al. 1981. Undisturbed sampling method using wire line sampler (in Japanese). Sampling symposium. Okumura T., et al. 1982. Soil investigation at Kansai international airport –The investigation for great depth- (in Japanese). Mechanization for construction. Andresen A. and Kolstad P. 1979. The NGI 54-mm Samplers for Undisturbed Sampling of Clays and Representative Sampling of Coarser Materials. State of the Art on Current Practice of Soil Sampling, Progressing of the International Symposium of Soil Sampling. Singapore. 13-21. Lunne T., et al. 1997. Sample disturbance effect in soft low plastic Norwegian clay. Symposium on Recent developments in Soil and Pavement Mechanics. Rio de Janeiro. 81-102. Japanese geotechnical standard (JGS 1313-2003). 2004. Method for Measuring Pore Water Pressure using Electric Transducer (in Japanese). Japanese Standards for Geotechnical and Geoenvironmental Investigation Methods –Standards and Explanations- . 368-376.

Une méthode de classification de la sensibilité des sols au moyen du piézocône Soils sensibility classification method from piezocone data Serratrice J.-F.

CETE Méditerranée, Aix en Provence, France

RÉSUMÉ : Une méthode de classification des sols naturels à partir des mesures au piézocône est proposée. La méthode procède en deux étapes et en référence aux résistances drainées et non drainées mesurées à l'appareil triaxial sur les sols carottés au préalable dans le même site. La méthode est présentée puis deux exemples d'interprétation sont proposés et commentés à partir de mesures recueillies en sondages. Les tendances d'évolution des mesures au piézocône avec la profondeur dans les massifs argileux homogènes récents sont examinées ensuite. ABSTRACT: A method of classification of natural soils from piezocone measurements is proposed. The method proceeds in two steps with reference to drained and undrained strength provided by triaxial testing on soil previously sampled into the same site. The method is presented and two examples of interpretation are proposed and discussed from data collected in-situ. Evolution trends of piezocone responses with depth in recent homogeneous clayey deposits are then examined. KEYWORDS: Soil, penetrometer test, piezocone, triaxial testing, shear strength, soil classification 1

résistances triaxiales drainées et non drainées ; 2) évaluation de la sensibilité des sols. Ainsi, les trois mesures du piézocône sont utilisées directement, sans recours à des variables normalisées. Cette méthode d'interprétation par analogie à l'essai triaxial est présentée tout d'abord, puis deux exemples d'illustration sont commentés ensuite.

INTRODUCTION

Une méthode de classification des sols est proposée ici sur la base des mesures au piézocône et par analogie avec les résistances déterminées à l'appareil triaxial. Cette méthode s'inscrit dans la lignée des méthodes de classification proposées dans le passé. Senneset et al. (1982) sont les premiers à avoir introduit la variable Bq définie comme le rapport entre la pression d'eau nette u2 – u0 et la résistance nette qt – v0, où v0 et la contrainte verticale totale à la profondeur considérée dans le massif, u0 la pression hydrostatique, qt la résistance de pointe et u2 la pression d'eau mesurée en position "u2". Parez et Fauriel (1988) ont proposé un guide de classification (et non pas un abaque) basé sur la représentation de qt en fonction de Bq, qui s'inspire de celui proposé par Senneset et Jambu (1984). Parez et Fauriel (1988) rappellent à ce titre que le guide qu'ils proposent " … ne dispense pas de réaliser, sur chaque chantier, un forage carotté.". Par la suite, Robertson et al. (1986), puis Robertson (1990), ont proposé une classification qui fait intervenir les trois composantes mesurées par l'intermédiaire de variables normalisées de la résistance de pointe qt et du frottement latéral unitaire fs en accompagnement de la variable Bq. Fellenius et Eslami (2000) ont proposé un abaque donnant la résistance effective qE (qE = qt – u2) en fonction du frottement latéral unitaire fs. Cette classification présente l'avantage d'utiliser directement les mesures pénétrométriques ou une combinaison linéaire de celles-ci. Cette résistance "effective" avait été introduite par Senneset et al. (1982). Schneider et al. (2008) proposent un cadre de classification des sols d'après les données du piézocône, qui associe à la fois la résistance de pointe qt et la pression d'eau u2 sous formes normalisées notées Q et U respectivement. Les mesures sont représentées dans trois diagrammes qui combinent les deux variables Q et U avec Bq, chacun de ces diagrammes étant plus pertinent qu'un autre, selon la nature du sol, pour établir la classification. La méthode par analogie à l'essai triaxial, proposée ici, préconise une utilisation directe de la mesure u2. Elle procède en deux étapes : 1) classification des sols en référence à leurs

2 2.1

PRÉSENTATION DE LA MÉTHODE Présentation

La méthode de classification des sols d'après les mesures au piézocône se fonde sur l'analogie qui peut être établie entre les comportements des matériaux observés en laboratoire à l'appareil triaxial et le fonçage d'un piézocône dans ces mêmes matériaux. La méthode procède en deux étapes, en partant des mesures brutes qt, fs et u2. A l'étape 1, la résistance de pointe est décomposée en une partie isotrope et une partie déviatoire en tenant compte de la pression d'eau u2. Pour cela, il est fait référence aux résistances drainées et non drainées mesurées préalablement en laboratoire à l'appareil triaxial. Cette décomposition permet de classer les sols, en distinguant les sols argileux dans lesquels se développent de fortes pressions d'eau, des sols sableux dans lesquels ces pressions sont égales à la pression hydrostatique ou sont négatives. L'interprétation se fonde sur les caractéristiques de résistance mesurées à partir d'échantillons carottés dans le site. L'étape 2 consiste à identifier les sols sableux sensibles, de faible compacité et peu résistants, exposés au risque de liquéfaction notamment. La méthode se fonde sur les variations relatives de fs et qt induites par la densification d'un sol. Le principe de la classification à l'étape 2 s'appuie sur des données de la littérature et sur des données pénétrométriques recueillies dans différents sites en France. L'intérêt de la méthode réside dans l'utilisation simultanée des trois mesures fournies par le piézocône et qui portent en elles l'effet de la profondeur sur la résistance (effet du poids des terres en tant que pression de confinement), pour des sols qui

611


peuvent être considérés comme normalement consolidés dans leur état naturel. 2.2

Classification à l'étape 1

Des variables équivalentes sont définies pour le piézocône par analogie aux variables p, pT et q de l'essai triaxial. Pendant l'essai triaxial, p est la pression effective moyenne, pT la pression totale moyenne et q le déviateur. La résistance du sol est définie dans le plan des contraintes effectives (p, q) par une droite de Coulomb de caractéristiques Cqc et Mc (ordonnée à l'origine et pente) dont découlent les propriétés effectives c' et ' (cohésion et angle de frottement). En contraintes totales et dans le plan (pT, q), les caractéristiques de résistance sont Cq cu, Mc cu, ccu, cu. (l'indice cu signifie consolidé non drainé).

contraintes totales équivalentes autour du cône. Parmi les trois droites définies en contraintes totales pour représenter les résistances non drainées des sols, la droite la plus proche de ce point permet de classer le sol. Un exemple est donné sur la figure 1. Trois mesures effectuées au piézocône dans trois faciès différents du même sondage y sont reportées (figurés pleins). Les pressions p'c sont d'abord recherchées sur la droites des résistance effectives (en trait épais). Puis, après adjonction de la pression u2, la classe de sol est déterminée par la droite de résistance totale la plus proche (en traits fins, figurés creux). Après classification, les figurés représentant les classes de sols sont reportés sur les diagrammes donnant qt, fs et u2 en fonction de la profondeur, comme sur les figures 3 et 4, puis sur tous les autres diagrammes dérivés. Il apparaît clairement sur ces exemples que les plus fortes pressions u2 sont attribuées aux limons argileux, alors que les plus faibles pressions sont attribuées aux sables et sables limoneux. 2.3

Recherche de la sensibilité des sols à l'étape 2

Après cette première étape de classification des sols, l'étape 2 consiste à détecter les sols sensibles. Les sols sont considérés comme sensibles s'ils appartiennent au quart inférieur gauche du diagramme (lgfs, lgqE) délimité par les deux courbes d'équations qE = 2000 fs 2 et qE = 2 fs –0,16 (qE et fs en MPa).

Figure 1. Principe de la classification des sols à l'étape 1 d'après les données du piézocône et les résistances triaxiales.

Dans la méthode par analogie à l'essai triaxial, les variables équivalentes définies à partir des mesures du piézocône sont p'c, pc et qt, une pression moyenne effective équivalente p'c, une pression moyenne totale équivalente pc et la résistance de pointe qt. Les résistances des sols mesurées au piézocône sont supposées s'exprimer au moyen des même droites de résistance dans le plan des contraintes effectives équivalentes (p'c, qt) et des contraintes totales équivalentes (pc, qt), comme indiqué sur le graphique de la figure 1. Tableau 1. Caractéristiques des résistances non drainées des sols. Sols

ccu (kPa)

cu (°)

Argiles, limons argileux

24

19,5

Limons, limons sableux

33

25,4

Sables, sables limoneux

50

36,9

Figure 2. Identification des sols sensibles dans le diagramme (fs, qE).

Ces courbes sont indiquées sur le diagramme (fs, qE) de la figure 2, en échelles arithmétiques. Les sols sensibles sont symbolisés par des figurés en rapport avec les croix représentant les classes de sol identifiées à l'étape 1. 3

EXEMPLES

Ce paragraphe présente deux exemples d'interprétation des données recueillies au piézocône. 3.1

Dans cet exemple et d'après les essais triaxiaux, la droite définie en "contraintes effectives" possède les caractéristiques Mc = 1,5 et Cqc = 0, soit ' = 36,9 ° et c' = 0. Les trois autres droites représentent les résistances "consolidées non drainées" dont les propriétés cu et ccu sont données dans le tableau 1 et qui représentent les résistances des "argiles et limons argileux", des "limons et limons sableux" et des "sables et sables limoneux". Ainsi, dans le graphique (p'c, qt), la mesure de qt permet de déterminer la contrainte moyenne effective équivalente p'c générée par la pointe sur la droite effective (Mc, Cqc). La contrainte moyenne totale équivalente pc s'obtient alors en ajoutant la mesure de u2 dans la direction isotrope, ce qui donne les coordonnées (pc, qt) du point représentatif de l'état de

612

Exemple 1

Le premier exemple concerne des données recueillies pendant une campagne de sondages au piézocône. Trois familles de sols apparaissent, qui sont indiquées dans le tableau 1. Leurs caractéristiques de résistance ont été mesurées au triaxial à partir des échantillons carottés dans le site. Les trois familles de sols sont frottantes, avec des résistances effectives communes (c' = 0 et ' = 36,9 °). Le tableau 1 indique les résistances non drainées correspondantes. La classification des sols à l'issue des étapes 1 et 2 de la méthode par analogie est représentée sur la figure 3 pour l'un des six sondages. Des sols sableux sensibles apparaissent entre 4 et 6,5 m de profondeur et entre 7,5 et 9,5 m, puis en des points isolés plus en profondeur.


apparaissent aussi et souvent près de la surface. La frange superficielle comprend un à deux mètres de remblais, qui sont souvent détectés comme sensibles.

Figure 3. Classification des sols d'après les données d'un sondage au piézocône et à l'issue des étapes 1 et 2 de la méthode de classification par analogie à l'essai triaxial.

D'après la description des sols établie à l'issue des carottages, les sols argileux représentent 16,8 % de l'ensemble des échantillons, les sols limoneux 8,5 % et les sols sableux 74,7 %, pour un linéaire effectif de 26,5 m de carottage. Ces pourcentages sont à comparer aux proportions des trois familles de sols fournies par les six piézocônes, qui sont : 9,4 % pour les sols argileux, 12,1 % pour les sols limoneux et 78,5 % pour les sols sableux. Une bonne correspondance apparaît ainsi entre les carottages et les sondages au piézocône interprétés en référence aux caractéristiques triaxiales mesurées au laboratoire sur les échantillons carottés. 3.2

Figure 4. Classification des sols d'après les données du sondage au piézocône SF5, site de Soccer Field, Gölcük (Turquie), à l'issue des étapes 1 et 2 de la méthode de classification par analogie à l'essai triaxial.

Exemple 2

Cetin et al. (2004) proposent une revue de différents cas de rupture de pentes qui se sont produites sur le rivage de la baie d'Izmit lors du séisme de Kocaeli du 17 août 1999 en Turquie. Plusieurs sites de cette région ont été explorés en 2000 au moyen de différentes techniques parmi lesquelles figurent des sondages au piézocône. Tableau 2. Nature des sols dans le site de Soccer Field à Gölcük (D'après Cetin et al., 2004). profondeur (m) 0 à 0,5 ou 1,5 0,5 ou 1,5 à 2,5 2,5 à 16,5

sol remblai d'argile silteuse brune sable silteux et silts argile silteuse molle très plastique

Les données numériques de quelques sondages au piézocône ont été importées directement d'une base de données (http://peer.berkeley.edu). Les mesures du sondage SF5 réalisé dans le site de Soccer Field à Gölcük sont utilisées ici à titre d'illustration. Le tableau 2 indique le profil des terrains donné par Cetin et al. (2004) jusqu'à 16,5 m de profondeur. Les sols rencontrés sont des sédiments fins récents de nature variée. Des sols sableux apparaissent dans tous les sites à différentes profondeurs. Mais des sols silteux et argileux

613

Figure 5. Extrait du profil de la figure 4 pour le sondage au piézocône SF5, entre 0 et 2,6 m de profondeur.

La classification des sols du sondage SF5 à l'issue des étapes 1 et 2 de la méthode par analogie à l'essai triaxial est représentée sur la figure 4. Les résistances triaxiales drainées et non drainées du tableau 1 ont été adoptées pour effectuer cette interprétation des données. Il apparaît des sols sableux sensibles entre 0,8 et 2,5 m de profondeur, des sols limoneux et argileux jusqu'à 16,5 m avec des pressions u2 positives, puis, au-delà, des sols sableux résistants et non sensibles. La figure 5 montre un extrait du profil pénétrométrique du sondage SF5 en surface et marqué par des sols sableux sensibles entre 1 et 2 m de profondeur. 4

ÉVOLUTIONS AVEC LA PROFONDEUR

Bon nombre d'enregistrements pénétrométriques obtenus dans des argiles molles et publiés dans la littérature font apparaître une augmentation quasi-linéaire des mesures avec la profondeur. Des droites représentent l'évolution moyenne de la résistance qt et de la pression u2 en fonction de la profondeur (et


parfois du frottement latéral unitaire fs). De tels profils apparaissent plus rarement dans les dépôts sableux, dont la lithologie et l'état initial sont caractérisés généralement par une forte variabilité spatiale. L'expérience acquise dans différents sites en France confirme ces tendances. Les gradients d'augmentation de qt avec la profondeur z varient autour 30 à 50 dans les argiles, ceux de u2, autour de 25 à 40. Pour la résistance de pointe, ce résultat s'explique par la relation empirique donnant l'évolution de la cohésion non drainée avec z et la relation donnant l'évolution de la contrainte verticale effective 'v0 avec z : cu = 'v0 / 3 qt = 15 cu 'v0 = ( - w) z

(1)

soit :

6

qt = 5 ( - w) z

(2)

avec  le poids volumique du sol et w le poids volumique de l'eau. Avec  = 18 kN/m3 et w = 10 kN/m3, il vient : qt = 40 z

(qt en kPa et z en m)

(3)

La formulation basée sur le modèle Cam-Clay de Chang et al. (2001) aboutit à des résultats semblables. En admettant par exemple que Mc = 1,2 pour le critère de Coulomb (' = 30 °), un degré de surconsolidation OCR = 1 (argile normalement consolidée), un indice de rigidité Ir = 100 pour un sol mou (rapport entre le module élastique de cisaillement G et la cohésion non drainée cu, Ir = G / cu) et le rapport de compressibilité  = 0,9 (rapport des coefficients de compressibilité Cc et de gonflement Cs,   1 – Cs/Cc), les relations suivantes apparaissent : cu= 0,322 'v0 qt = 12,2 cu + p0 u2 = 5,68 cu + p0

(4)

où p0 est la pression moyenne effective (p0 = ('v0 + 2'h0)/3, 'h0 contrainte effective horizontale). Puis, en admettant un coefficient des terres au repos K0 = 'h0/'v0 = 0,5, p0 s'écrit p0 = 2 'v0 / 3 + u0. Enfin, avec  = 18 kN/m3 et w = 10 kN/m3 : cu= 2,5 z qt = 47 z u2 = 30 z

(cu en kPa et z en m) (qt en kPa et z en m ) (u2 en kPa et z en m)

(5)

Tableau 3. Pentes des profils pénétrométriques dans les sols argileux. Exemple

qt/z (kPa/m)

fs/z (kPa/m)

u2/z (kPa/m)

1

31

0,35

37

2

38

0,55

31

Ces relations donnent des ordres de grandeur des gradients compatibles avec l'observation. Le tableau 3 indique les pentes évaluées dans les niveaux argileux des profils pénétrométriques des figures 3 et 4. 5

La méthode proposée s'accorde aux méthodes en usage en matière de reconnaissances pénétrométriques, en comprenant à une étape d'identification des sols (profiling), puis une étape de recherche des sols sensibles. Cette méthode cherche à tirer parti des essais triaxiaux pour interpréter les données pénétrométriques, ce qui suppose que les reconnaissances géotechniques prévoient à la fois la réalisation de sondages carottés et de sondages pénétrométriques, pour aboutir à une analyse dédiée du site. Elle ne vise pas à revêtir un caractère universel, en utilisant une classification unique des sols. Concernant les enregistrements au piézocône recueillis dans des dépôts argileux homogènes et récents, l'expérience fait apparaître des gradients d'évolution des mesures qt et u2 avec la profondeur qui peuvent être encadrés par des ordres de grandeur répétitifs.

CONCLUSION

Une méthode a été proposée d'identification des sols à partir des données mesurées au piézocône. Cette méthode procède en deux étapes : étape 1, classification des sols ; étape 2, identification des sols sensibles. L'exploitation des données fait référence aux résistances drainées et non drainées mesurées préalablement à l'appareil triaxial sur les sols carottés dans le site.

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RÉFÉRENCES

Cetin O.K., Youd T.L., Seed R.B., Bray J.D., Stewart J.P., Durgunoglu T., Lettis W., Yilmaz M.T. (2004) Liquefaction-induced lateral spreading at Izmit Bay during the Kocaeli (Izmit)-Turkey earthquake. J. of Geotech. and Geoenv. Engng., vol. 130, n° 12, 1300-1313. Chang M.F., Teh C.I., Cao L.F. (2001) Undrained cavity expansion in modified Cam clay: II Application to the interpretation of the piézocône test. Géotechnique, vol. 51, n° 4, pp. 335-350. Fellenius B.H., Elsami A. (2000) Soil profile interpreted from CPTu data. "Year 2000 Geotechnics", Geotech. Engng. Conf., Asian Institute of Technology, Bangkok, Thailand, 27-30 novembre 2000, 18 p. Parez L., Fauriel R. (1988) Le piézocône. Améliorations apportées à la reconnaissance des sols. Revue Française de Géotechnique, n° 44, pp. 13-27. Robertson P.K., Campanella R.G., Gillepsie D., Greig J. (1986) Use of piezometer cone data. Use on in situ tests in geotechnical engineering; Proc ASCE Speciality Conference In Situ '86, Blacksburg, pp. 1263-1280. Robertson P.K. (1990) Soil classification using the cone penetration test. Canadian Geotech. J., vol. 27, n° 1, pp. 151-158. Schneider J.A., Randolph M.F., Mayne P.W., Ramsay N.R. (2008) Analysis of factors influencing soil classification using normalized piezocone tip resistance and pore pressure parameters. J. Geotech. and Geoenv. Engng., vol. 134, n° 11, pp. 1569-1586. Senneset K., Jambu N., Svano G. (1982) Strength and deformation from cone penetration tests. Proc. 2nd Euro. Symp. on Penetration Testing, ESOPT-2, Amsterdam, vol. 2, pp. 863-870. Senneset K., Jambu N. (1984). Shear strength parameters obtained from static cone penetration tests. Proc. on Strength Testing on Marine Sediments. Laboratory and In-situ Measurements. ASTM Special Technical Publication 883, Symp. San Diego, pp. 41-54. Serratrice J.F. Identification des sols argileux, limoneux et sableux du plateau deltaïque du Var à partir de sondages au piézocône, Soumis au Bulletin des Laboratoires des Ponts et Chaussées.

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Correction of soil design parameters for the calculation of the foundation based on the results of barrettes static load test Correction des paramètres de conception du sol pour le calcul sur la base des résultats de test de barrettes de charge statique Shulyatiev O., Dzagov A., Bokov I., Shuliatev S.

Gersevanov institute for soil bases and underground structures, e-mail:[email protected].

ABSTRACT: Geotechnical investigations and design works were being performed in 2008-2010 for the construction of Okhta-center high-rise tower in St. Petersburg. Unique static load tests of 80 m deep barrettes were performed in 2010 as part of design process. 3 barrettes were tested simultaneously as a group and two were tested individually. The tests were planned in such a way as to get the standard values of bearing capacity of barrettes, and to clarify the parameters of soil needed for the calculation of the barrette foundation.The comparison of the bearing capacity values with the values calculated using Russian and foreign building codes is provided. Limitations of currently used codes are shown. RÉSUMÉ : Les études géotechniques et les travaux de conception ont été réalisées en 2008-2010 pour la construction du centre Okhta haute tour à Saint-Pétersbourg. Uniques essais de charge statique de 80 m de profondeur barrettes ont été réalisées en 2010 dans le cadre du processus de conception. 3 barrettes ont été testés simultanément en tant que groupe et deux ont été testés individuellement. Les tests ont été planifiés de manière à obtenir les valeurs standard de la capacité portante des barrettes, et de préciser les paramètres du sol nécessaires pour le calcul de la fondation barrette. La comparaison des valeurs de capacité portante avec les valeurs calculées à l'aide des codes de construction russes et étrangères est fournie. Limites des codes actuellement utilisés sont affichés. KEYWORDS:piles, barrettes, static load test, shaft friction, FEM, bearing capacity, high-rise building. 1

modulus needs to be adjusted to plate loading test modulus, and if that is not possible soil anisotropy factor needs to be determined for conversion of soil modulus in the horizontal direction to the modulus in the vertical direction. Trial Barrette static test was scheduled as part of the pile foundation design process. Given the high testing load "Osterberg" method were considered economically effective. Given specifics of the method, in addition to pile bearing capacity assesment, one can provide design engineer with the possibility of making “deep plate loading test”.

INTRODUCTION

In recent decades, in Russia there is a steady increase in the number of tall buildings being built, of which a substantial part is the building higher than 150 m. Building heigher then 150m need a special approach to design. Existing bulding codes in Russia and other countries as well, can not fulfill the requirements of modern day high-rise construction. For foundation constructions existing codes are limited by relatively small depth of ground investigation and testing loads. In the current RF building codes plate loading test is considered as the reference method for soil Young modulus estimation. According to codes soil modulus determined by other methods should be adjusted to plate loading test modulus. It is not always possible, given the great depth of the soil used as the bearing layer of high-rise building foundation. This paper discusses the engineering properties of Vendian clay as a bearing layer of Okhta tower high-rise building in St. Petersburg. According to building design it’s pile foundation will be embedded in Vendian clay layer lying deeper then 45 m from ground surface. Building design The project has a device for high-rise building with pile foundation bearing on Vendian clay layer, lying with a mark of -45 m B.S.V. Laboratory tests on odometer and triaxial schemes were made during ground investigations to study the properties of Vendian clays. Given the depth of bearing layer pressuremeter test were selected as in-situ test method. Laboratory testing of soil extracted from great depth usually complicated by disturbanceof soil samples, caused by stress relief and preparation of samples for testing, and by the complexity of high-precision measurements of deformation of the sample (especially true for high stiffness soils). Pressuremeter test, in turn, has no alternatives for soil testing in-situ at greater depths. Design value of pressuremeter Young

Figure1 .Location of test barretes.

Trial Barrettes were made from the surface of the soil. The working part of a 65-m barrette was made of B40 grade reinforced concrete. Barrettes were constructed by the conventional technology - in the trenches, excavated under the protection of bentonite slurry.

615 1



2

GEOLOGICAL CONDITIONS

4

Geological conditions of the construction site can be described as unfavorable for civil engineering and especially for high-rise buildings. Weak water-saturated soils lies to the depth of 30 ... 35 m. Underneath these soils is a layer of moraine deposits of small thickness. From a depth of 45 m liesVendian clay. Rock, commonly used as a bearing stratum for high-rise buildings are located at depths of over 200 m. Considering the aforementioned facts Vendian clay was selected as the bearing layer for the Okhta tower pile foundation. Vendian clay is relatively strong soil and classified as hard clay and weak rock at the same time. Despite the relatively high strength properties Vendian clay exhibits long-term development of deformations in time under load. It should be noted that engineering properties of these soils in Saint-Petersburg is mostly unstudied. 3

TESTING SETUP

The test program was design in such a way as to achieve the following goals:Determination of the bearing capacity of barrette and it’s individual fragments;Determination of loadsettlement characteristics for "top-down" loading scheme;evaluation of the Young modulus for the underlying the barrette base;Evaluation of interface strength on the shaft of the barrette. Three of the five tested group barrettes were equipped with loading device installed in two levels, two single piles in one level. Single-level and two-level testing scheme and barrette part nameingare shown in Fig. 2

4.1

TEST RESULTS Test of the lower parts of barrettes B1...B5.

The value of the load reached during first stage of testing was 40MN and 48 MN for second stage. Bearing capacity of the lower parts of the test group barrette was 90% of the bearing capacity for a single barretes test, due to the group effect. Load-settlement characteristics for barrettes B1…B5 shown on Figure 3 and shows that settlement of Barrette B2, located between Barrette B1 and B3 is 15% more than that of Barrette B1. This effect is referred to well-known concepts of group effect in pile groups. Pile settlement in the group always exceeds the settlement of single pile, and the settlement of central pile is highest. With the growth of the number of piles in the group this effect expected to increase. By means of mathematical modeling of group testing and achieving the same group effect, which was observed in the trial, one can confirm the accuracy of the model input parameters, and to validate its use for the calculation of the entire foundation. The elastic component of the Barrette B1 ... B5 base settlement is 13 ... 20%, and the residual inelastic component reaches 79 ... 87.6% (Fig. 3), i.e. much of the ground under the base of Barrette undergone plastic deformation. In the analysis of Fig. 3 it may be seen that load-settlement characteristics can be divided into several stages. In the first phase, with a load values up to 5MN, load held by the shaft friction on the surface of Barrette part, and movement up to 1 mm recorded. At the 2nd stage of loading barrete part is moved and load being transferred to barrete base. Soil underneath the barrete disturbed by drilling began to compact under load. Settlement of barrete base increases linearly with load until 20…40 MN load value is reached. As the barrette part movement increases, shaft frictions on its side reaches a maximum value and remain constant to the rest of stage 1. Due to this effect further increment of load transferred directly to the barrette base. The final stage is characterized by an increase of settlement increment per unit increment of load, indicating that the transition of the ground under the base of barrette to the plastic state.

Figure2. Scheme of barrete parts namings Barrettes with one level jacks were tested in one phase, the pile with two levels - in two stages. In the first stage the lower part of the pile is loaded with the lower level of the jack. In the second stage the upper jack level creates load on the middle part of barrette. During a first stage of testing upper level jacks are closed and load transfer through them is not different from a solid barrette section. During loading of the upper level, the hydraulic system of lower jacks is open into atmosphere, making them closing or opening freely. During the testing of the upper level when the lower level is open and jacks are retracted, the entire load of the upper level of loading is transmitted to the shaft of the barrette middle fragment. As the criteria for test advancement standard RF deformation stabilization criteria 0.1 mm/h was used. It is 2.5 times more rigorous then the standard European 0.25 mm/h. Results comparison with two different stabilization criteria showed, that application of the criterion of 0.25 mm/h underestimates the magnitude of barrete base displacement by 30%. Choice of stabilization criteria is especially important when the testing jack located near the barrette base in clay soils, as in this case, due to soil consolidation deformation process is much slower.

616

Figure 3. Stage 1 test results. In order to clarify the shaft bearing capacity for bottom part of Barrette B4, the loading increments in the first stages of the load testing has been reduced from 5MN to 2.5 MN, which led to an increase in the number of stages in the load range of up to 20 MN from 4 to 8 . An interesting finding was the fact that, regardless of the number of stages loading time spent on testing barrette B4 and B5, was similar and was 277 and 259 hours, respectively. Concluding the analysis of bottom level testing one can mention high repeatability of results, which indicates the


th


homogeneity of the subsoil under the Barrette base and the good quality of their production. 4.2

convergence criteria between the experimental and calculated values.

Test of the middle parts of barrettes B1...B5.

At the end of the first stage of testing bottom level jacks are retracted, and their hydraulic system is open. In this configuration, the lower level of jacks do not transfer the load from a upper jacks level on the base. In this case, loading of the upper jack level resisted only by shaft resistance of the middle barrette part, allowing to accurately determine shaft friction value.

Figure 5. Comparative charts of data obtained from the experiment and the results of the calculation. Comparison of modeling results with the experimental data shown in Figure 4. The graph shows the results of calculations with adjusted characteristics of soil is almost equal to the results of the test. It should be noted that non-compliance of Barrette behavior during unloading caused by incorrect rapid unloading procedure. Figure 4. Stage 2 test results.

Table 1.Adjusted by FEM calculations soil properties.

GE Soil number classification 11 Hard clay 12 Hard clay 12а Hard clay 13 Hard clay

Compared with the test of the first stage (lower jack level) results show significantly greater variation in the ultimate shaft friction values. Load-settlement curves for the second-stage tests are shown in Fig. 4. Load-settlement curves shown on Fig. 3 characterized by initial almost flat part, with a slightly longer length for Barrette B2. The angle of the load-settlement curves for B1 and B3 began to increase after the load value of 20 ... 25 MN, and for B3 35 MN. The presence of a longer horizontal part on load-settlement curve for B2 may be due to heterogeneity of soil conditions along the tested barrette part, or, more likely, due to its central position in the group. 4.3

5

Density, g/cm3 21.3 22.2 21.1 22.3

Е, Poisson ratio MPa ν 50 0.25 200 0.2 105 0.22 252 0.18

φ, ° 17 25 18 27

С, kPa 150 330 200 491

COMPARISON OF OBTAINED SHAFT UNIT FRICTION WITH BASIC CALCULATION METHOD RESULTS.

The main purpose of the second phase of the test was to determine the specific shaft resistance values for middle parts of Barrette B1 ... B3. Resistance value is determined by dividing the applied load on the shaft surface area of the middle part of barrette.

Back-analysis of test results performed by FEM.

One of the most effective tools for the analysis of load test made by the Osterberg scheme is the reverse calculation method with regard to elastic-plastic soil properties by means of FEM. The reverse calculation has several objectives: 1) Calibration of design parameters of adopted soil model 2) evaluation of the bearing capacity of single pile in the top down loading conditions 3) assessment of the applicability and adequacy of the chosen soil model. The starting point for the reverse calculations is the soil properties obtained by laboratory testing. By varying individual soil model parameters one can identify the most important of them, and then achieve convergence between experimental and calculated results. The first iteration of calculation based on laboratory determined soil properties showed that the calculated values of barrete upward movement is 6 times larger than the experimental values, and downward movement is overestimated by 2 times. This suggests that the characteristics of soils, provided through laboratory testing are very different from the characteristics of the soil in-situ. Taking into account observed discrepancy the objective was to find such soil characteristics, which would have shown the best convergence of calculation with the experiment. Barrette movement and stress along its body were chosen as

Figure 6. Shaft friction-movement curves. As can be seen from Fig. 5, for barrette movements of up to 20 mm shaft friction increasing drastically to 190 ... 290 kPa. A further increase in displacement to 60 mm results in a small (about 60 kPa) monotonic increment of resistance. The peak (maximum) value of the shaft resistance was not clearly

617


th


observed as the resistance monotonically increases up to 100 mm movement. By analyzing the obtained movementsresistance curves it can be noted that values of shaft resistance for Vendian clays can be taken as corresponding barrette shaft resistance for 40 mm (according to RF building codes) movement. Design shaft resistance values provided for assessment of pile bearing capacity by SP24.13330.2011 at depths greater than 40 m are assumed constant and equal to 50 kPa, which is 4.6 ... 6.4 times less than the values obtained in static Barrette test. For comparison, ultimate values of shat friction were also calculated for the most common approaches worldwide: the α-method (Skempton 1959) using undrained strength parameters and Mohr-Coulomb law. In calculation with the alpha method lower (α=0.3) and upper bound (α=0.6) values of alpha were examined. Resulting specific values of shaft friction values were 250 and 500 kPa respectively. For calculation by MohrColumb earth pressure at rest coefficient K0 were taken for nonconsolidated soil by well-known Jaky equation and as for overconsolidated soil with OCR=2,5. Resulting K0 values were 0.66 and 0.99 respectively. Factor of 0.5 for interface strength also applied according to SP24.13330.2011. Specific shaft friction values obtained by this calculation method were 270 kPa for K0=0.66 and 460 kPa for K0=0.99. Thus, the lower limit of the specific shaft friction, calculated using the mechanical properties of soils were within the values obtained by the results of static Barrette tests, and the upper limit value – was higher on average of 1.7 times. One of the reasons for this discrepancy may be that the central parts of barrette during the first stage of test (lower part testing) has the 5...7 mm upward movement, during which partial mobilization of shaft friction forces in opposite direction were observed. 6

CONCLUSIONS

Trial works confirmed the technical feasibility of barrettes construction with cross-section of 1.5 x 3.0 m length of 85 m by the standard "slurry wall" technology in difficult sub-soil conditions of St. Petersburg. It is advised to implement Osterberg testing technique (by cast in pile submerged jack) for the deep foundation of high-rise buildings. The method allows to use pile parts as an anchor system and to clearly determine the values of unit shaft friction and base resistance. It is recommended to install two levels of jacks in a pile: one near the base of the pile, and the second in the middle of the main bearing layer. It is critical to install several levels of strain gauges in the pile along its length. Pile testing at construction site should be seen not only as method to determine pile bearing capacity but as an effective method to calibrate design parameters of adopted soil model, and to assess its applicability and adequacy. Soil parameters provided by ground investigation can by checked and adjusted if necessary. During design process of tower foundation, obtained results of unit shaft friction and base resistance should be used as the control values, against which the calculation results are checked. The calculations made on the basis of experimental data showed what the values of the mechanical properties of soils determined by the laboratory testing has severely underestimated soil strength and deformability parameters due to sample disturbance, the influence of the scale factor & etc. As a result of the tests it was found that the Vendian clays can provide high values of shaft friction and base resistance. The experimental values of shaft friction and base resistance exceed the ultimate values provided by codes by 4 ... 6 and 1.6 timesrespectively.

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7

REFERENCES

Aspects of pile testing for high-rise buildings on the example of ODTS «Okhta» tower.High-rise buildings journal 6 (2012).9699.PetrukhinV.P., ShuliatevO.A., BokovI.A., ShuliatevS.O. Cast-in-situ Bored Piles in London Clay, Geotech,Vol.9.Skempton, A.W., and Northey, R.D. (1952)

Characterization and Settlement Modeling of Deep Inert Debris Fills Caractérisation et modélisation du tassement de dépôts épais de gravats inertes Somasundaram S., Khilnani K., Shenthan T.

Advanced Earth Sciences, Irvine, California, USA

Irvine J.

Irvine Geotechnical, Pasadena, California, USA ABSTRACT: Inert debris fills are difficult to characterize and model by normal geotechnical methods, due to their inherent heterogeneity, very large particle size, and nested and voided structure. The approach taken to characterize a 54 m deep inert debris fill, model its settlement behavior under seismic loading and groundwater level rise, and develop remedial measures to render it suitable for development is presented. Fines migration into open cavities and collapse of nested structure were determined to be the primary settlement mechanisms for this material. An upper bound estimate of cavity volume vulnerable to fines migration and collapse was made based on the results of large scale in-situ density and gradation tests. Settlement was estimated for various percentages of cavities becoming filled, and compared to case histories of dry fill settlement from the San Fernando and Northridge earthquakes. The proposed remedy involved partial removal of the debris fill and replacement as a compacted fill cap to attenuate the surface expression of differential settlement occurring in the underlying debris fill. Surface manifestation of settlement was simulated using FLAC. Charts were developed relating cap thickness to surficial manifestation of differential settlement. RÉSUMÉ : Les dépôts de gravats inertes sont difficiles à caractériser et à modéliser par les approches géotechniques usuelles, en raison de leur hétérogénéité intrinsèque, de la grande taille des particules qui les constituent, et de leur structure lacunaire et emboîtée. On présente une approche utilisée pour caractériser un dépôt de gravats inertes de 54 m d'épaisseur, modéliser son comportement de tassement sous chargement sismique et sous l'effet d'une montée du niveau de la nappe phréatique, et développer des mesures de remédiation en vue de le rendre propre à l'utilisation. On a pu montrer que la migration des fines dans les cavités ouvertes, et l'écrasement des structures emboîtées, constituent les mécanismes principaux responsables du tassement pour ce matériau. Une estimation par excès du volume des cavités vulnérables par la migration des fines et écrasement a été établie sur la base d'essais à grande échelle de densité in-situ et de granulométrie. Le tassement a été estimé pour divers proportions de remplissage de cavités, et comparé à des observations historiques de tassement de remblais secs suite aux séismes de San Fernando et de Nothridge. Le remède proposé implique un retrait partiel du dépôt de gravats et son remplacement par une couche de remblai compacté, en vue de minimiser l'expression en surface des tassements différentiels survenant dans le dépôt de gravats sous-jacent. Le déplacement en surface a été simulé en utilisant le logiciel FLAC. La relation entre l'épaisseur de la couche de protection et l'incidence en surface du tassement différentiel a été exprimée sous forme d'abaques. KEYWORDS: inert debris landfills; debris fills; seismic settlemen 1

INTRODUCTION

Inert debris landfills in urban areas are increasingly becoming potential sites for industrial / commercial redevelopment due to scarcity of vacant land and a desire by local communities to turn blighted areas into revenue sources. These fills, generally placed in abandoned mine pits, could be over 50 m deep and typically consist of uncontrolled fills of construction and demolition (C&D) debris. Due to their inherent heterogeneity and very large particle size they are difficult to characterize and model by normal geotechnical methods. This case study presents the approach taken to characterize a deep inert debris fill, model its settlement behavior under seismic loading and groundwater level fluctuations, and develop remedial measures to render it suitable for development. The inert debris fill, located in the City of Irwindale in southern California, consists of over 8 million cubic meters of C&D waste placed over a period of 15 years within a 54 m deep abandoned open pit gravel mine covering a footprint of 22 hectares. The lower 2 to 12 m of the pit was filled with hydraulically placed silt, a by-product of aggregate mining operations. Review of placement records indicated that the inert debris fill above the silt layer consists of a succession of 1 to 3 m thick lifts of rubble consisting mostly of broken concrete, brick, tile and asphalt capped with 15 to 30 cm thick lifts of sandy and silty soils. The soil layers were generally placed and compacted above each rubble lift to provide a suitable surface for rubber tired traffic. The entire inert debris fill is capped with a 3 m thick layer of compacted soil to allow for utility excavation and structure foundation at the finished surface.

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Placement records indicate that initially the rubble fills were placed with some degree of material processing (crushing of oversize concrete clasts) and compaction. However, much of the inert debris fill was loosely end dumped with little or no control of lift thickness, particle size or compaction. The groundwater level was approximately 36 m below the ground surface during filling, but could rise by about 12 m based on historic records. An idealized profile of the fill stratigraphy is shown in Figure 1. Compacted Fill Cap (3 m)

150 to 300 mm

1 to 3 m

Rubble Fill (Thickness varies)

Potential highest groundwater level (24 m below ground surface)

Previous highest groundwater level in fill (36 m below ground surface)

Silt Deposit (Thickness varies)

Figure 1. Debris Fill Stratigraphy


The site is vulnerable to relatively high levels of seismic loading, with a design peak ground acceleration of approximately 0.53g per the building code. Deaggregation analysis indicated the corresponding moment magnitude to be 6.7. The area is zoned for industrial or commercial development. The owners are evaluating remedial measures to make the site suitable for building development. There are no industry-accepted standards or case histories to predict settlements of inert debris fill containing significant oversize fragments and significant open cavities. Case histories of seismic settlements of unsaturated fills are generally limited to earthfill/rockfill dams and compacted soil fills. Laboratory cyclic simple shear test data relating cyclic shear strain to volumetric strain, that may be used to estimate the settlement of unsaturated fills under seismic shaking, are limited to sands (Silver and Seed 1971, Pyke et al 1975), and finer grained compacted fills (Stewart et al 2002). Charles (2008) documents case histories of long-term settlement and collapse potential of uncontrolled opencast mining backfills in Britain. The City of Irwindale is currently conducting a laboratory study to evaluate the potential for wetting induced settlements (hydrocollapse) in inert debris fills. 2

FIELD INVESTIGATIONS

Field investigations for this site included Becker hammer borings, surface and downhole geophysical surveys, downhole video logging, test excavations and large scale in-situ density and grain size distribution tests. Neither the Becker penetration tests (BPTs) nor the surface and downhole seismic surveys, proved to be suitable to characterize the heavily nested and voided nature of the fills. The presence of very large size fragments appear to significantly skew the measured Becker blow counts and shear wave velocities, making these methods incapable of adequately differentiating between well compacted, grading code - compliant fills (derived from the same debris materials), and the loose debris fills with voids/cavities. This conclusion has been confirmed by studies performed by the City of Irwindale at other debris fill sites (Geomatrix, 2007).

A qualitative evaluation of the voided / nested structure of the inert debris fill was performed by measuring the rate of water percolation in large diameter test holes. After completing the large diameter in-situ density tests, the plastic sheeting used to line the test hole was pulled out and the water level drop was monitored. The water levels dropped very rapidly (emptied in a matter of minutes) in test holes in debris fills, while the water levels stayed full for several days in the MAD tests holes, confirming the presence of significant voids / cavities in the debris fill. Field bulk gradation tests performed on the bulk samples excavated from the density test pits showed the following distribution: Table 2. Summary of Field Gradation Test Results Material Size Boulders (>300 mm) Cobbles (>75 mm) Gravels (>19 mm) Finer than (19 mm)

Range (%) 3 to 23 10 to 25 6 to 20 44 to 66

Average (%) 11 18 14 57

Visual observations of the materials removed from the test excavations suggest that the oversize fraction is greater than the amounts listed above, since representative amounts of very large concrete clasts could not be included in the material from 1.5 m diameter test holes. The actual boulder size fraction (> 300 mm) was estimated to be in excess of 20 percent by weight. 3

SETTLEMENT MODEL

The settlement model used in the analysis considered the layered nature of the debris fill consisting of a succession of 1 to 3 m thick voided and nested rubble lifts capped by 15 to 30 cm thick loose to medium dense soil lifts. The total debris fill may be considered to consist of nested oversize clasts (defined as materials lager than 19 mm for purposes of this analysis), infill soils (minus 19 mm fraction that partially fills the cavities between clasts and also caps individual layers of rubble), and cavities (Figure 2).

Mapping of two deep test excavations to 21 m depth in the poorly controlled debris fill, confirmed the layered filling pattern consisting of thick rubble fill lifts capped by thin soil layers. The layered filling pattern was also apparent in the BPT logs. The rubble fill consisted of concrete clasts and blocks up to 2 m in size (with abundant rebar), mixed with brick, tiles, asphalt concrete, crushed glass and variable amounts of soil infill. Large voids, cavities and nesting were very common. Eight large diameter ring density tests (1.8 m diameter x 1.5 m deep) performed as per ASTM D5030 in the inert debris fill at various depths (ranging from 5 to 15 m below ground surface) in the test excavations, and eight sand cone tests performed on soil layers or soil rich fills gave the following results. Table 1. Results of In-situ Density Tests Material Inert Debris Fill Soil Layers

In-situ Dry Density (gm/cc) Range Average 1.22 – 2.03 1.77 1.45 – 1.86 1.64

Average Void Ratio 0.43 (et ) 0.62 (es)

The in-situ densities of the inert debris fill were compared to field maximum achievable density (MAD) tests performed on inert debris materials placed in 30-cm thick lifts and compacted by 50 passes of heavy earthmoving equipment (combination of Caterpillar 820 front end loader and 825 compactor). The corresponding MAD dry densities ranged from 2.03 to 2.13 gm/cc.

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Figure 2. Debris Fill Structure

When subjected to seismic loading and/or saturation due to groundwater rise, the predominant mechanisms of settlement in the debris fill are considered to be partial filling of the cavities by fines migration (cap soils migrating into the underlying nested rubble), and collapse of the nested structure. Volumetric compression of the infill soils and soil lifts will also take place, but they are considered to be significantly smaller than the two dominant settlement mechanisms. The volume of cavities between the nested clasts, as a percentage of the total volume of fill, will, therefore, form an upper bound of the potential volumetric strain / settlement of the fill. The volume of cavities in the fill (Figure 2) as a ratio of the total fill volume, was estimated as shown below, based on the void ratio of the entire debris fill, et (calculated from large diameter ring density tests), void ratio of the infill soils, es (calculated from the sand cone density tests), the ratio of weight of clasts to weight of infill


groundwater rise, the two components of settlement (seismic and hydrocollapse) are not considered to be cumulative.

soils, R (from particle size distribution tests), and the specific gravity of the clasts (Gc) and infill soils (Gs). The ratio of volume of cavities to volume of solids, ec, and the relative volume of cavities with respect to the total volume for debris fill (Pc) may be expressed as: ec = et – es / (1+R.Gs/Gc)

(1)

Pc = ec / (1+ et)

(2)

0

0

Average Total Seismic Settlement (mm) 100 200 300 400

500

5

Based on the above equations, and using the average values of et (=0.43), es (=0.62), R (43%/57% = 0.75) and specific gravity (Gs = 2.65; Gc = 2.4), the average volume of cavities within the poorly controlled debris fills was calculated at 6.6% of fill volume. The calculated volume of cavities agrees well with field experiment estimates of cavity volume made at other inert debris fill sites in Irwindale with similar materials and filling practices. Those evaluations included a controlled in-situ pilot grouting test which resulted in a grout take of 4.4 to 7.2% of total volume, and an in-situ dynamic compaction test which resulted in a volume reduction of 5 to 7% of total fill volume (AMEC, 2008).

Depth (m)

10 15

% of Cavities Filled (p) 0%

20

10% 20%

25

30% 30

However, not all of the calculated cavity volume is available for fines migration / collapse. Actual volumetric strain and the resulting settlement is proportional to the volume of cavities that are closed or filled with fines in the event of an earthquake or hydrocollapse caused by rise in groundwater level. This is a function of many factors including the grain size distribution of the oversize clasts, accessibility of cavities to overlying infill soils, cohesion of infill soil and intensity and duration of seismic shaking, and cannot be reliably estimated in the absence of material-specific physical modeling. Therefore, a parametric settlement evaluation considering various percentages (p) of total cavity volume becoming filled was performed. The results are summarized as average settlement versus depth plots (Figure 3). The settlements shown in Figure 3 for each value of p, represent the average of the calculated settlements at six BPT locations across the site. Although the total thickness of debris fill was similar at each location (approximately 33 m), the thickness of the poorly controlled, layered rubble fill vulnerable to fines migration/collapse was variable (ranging from 15.6 to 25.0 m).

Figure 3 Distribution of Seismic Settlement with Depth

4

REMEDIAL MEASURES

The remedial measures recommended for limiting settlement at the site to within agency-defined guidelines or structural tolerances, consisted of partial removal of the existing debris fill and replacement with a properly processed and compacted fill cap. The required cap thickness could also be achieved by a shallower removal and replacement combined with in-situ ground improvement of the lower part of the debris fill by dynamic compaction. With increasing thickness of cap, the fill thickness left in place that is vulnerable to settlements would decrease. The cap will also help attenuate the differential settlement taking place at depth as it manifests at the surface of the fill cap. The surface manifestation of settlement was simulated by numerical modeling using FLAC. A representative twodimensional cross section across the entire site was considered. The fill cap was modeled as a non-linear elastic – perfectly plastic Mohr-Coulomb material. The initial shear modulus for the cap was based on the average shear wave velocity of 268 m/sec measured in the compacted fill. The modulus degradation curve was based on the Seed-Idriss relationship for sand. The calculated seismic / hydrocollapse settlement of the debris fill underlying the fill cap, was applied as nodal displacement boundary conditions at the base of the cap. Since the thickness of poorly controlled rubble fill and the corresponding settlements are variable across the site, the nodal displacements were specified as randomly varying over the range of settlements calculated at the 6 BPT locations.

The average settlement corresponding to 20% of cavities filled (p = 20%), was computed at 28 cm (approximately 1.32% of poorly controlled debris fill thickness or 0.85% of total debris fill thickness). The latter value compared favorably with some case histories of dry compacted fills in southern California which settled by 0.6 to 0.9 percent of fill thickness during the M6.6, 1971 San Fernando, and the M6.7, 1994 Northridge earthquakes, under ground accelerations comparable to the design ground motions for the site. Considering the significant heterogeneity of the debris fills, the seismic settlements could be higher or lower than that predicted for p = 20%. To bracket this uncertainty, seismic settlements under the design earthquake were calculated for ‘p’ ranging from 10% to 30%. The resulting settlements ranged from 0.4 to 1.1 percent of total debris fill thickness.

The nodal displacements (ρn) were generated as follows: ρn = ρmin + r. (ρmax - ρmin) where, r is a random number between 0.0 and 1.0 (determined by a random number generator for the numerical analyses) and ρmin and ρmax are the minimum and maximum values, respectively, of calculated seismic/hydrocollapse settlements, for a given value of p. The specified random nodal displacements were applied at 1.5 m horizontal intervals along the base of the cap. The modeling was performed for p = 10%, 20% and 30%.

A 12 m thick zone of debris fill immediately above the current groundwater level could become saturated if the groundwater level was to rise to the historic high groundwater level. This zone has not been saturated since the time of placement. Settlement due to groundwater saturation was considered to result from the same mechanisms of fines migration and collapse, and was assumed to be of the same order of magnitude as the seismic settlements. These settlements, estimated to range from 75 mm to 150 mm, occur approximately 24 m below ground surface (the depth of the high groundwater level below ground surface). Because the same mechanisms (migration of sands into open voids and collapse) apply to both seismic settlement and settlement due to

Typical FLAC analysis results as illustrated in Figure 4, show the original and deformed shape (grid) of a segment of the fill cap as a result of the random differential settlement applied at the base of the cap, for cap thicknesses of 12, 18 and 24 m, respectively. As the fill cap thickness increases, the magnitude of the total and differential settlement of the material left in

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place decreases, and the attenuation of the surface manifestation of differential settlement increases. For the case illustrated in Figure 4, the differential settlement at the base of the cap decreases from 122 mm to 43 mm as the cap thickness increases from 12 m to 24 m. The corresponding maximum differential settlement at the surface (over a 9-m horizontal distance) decreases from 56 mm to 8 mm. 9m Max. ρ

Max. ρ

= 55 mm

diff

diff

Max. ρ

= 120 mm (a) 12 m Compacted Soil Cap

diff

Max. ρ

= 33 mm

= 105 mm

diff

Figure 5. Surface Manifestation of Settlements (b) 18 m Compacted Soil Cap

Max. ρ

Max. ρ

diff

diff

 A settlement model based on partial filling of cavities by fines migration and collapse of nested structure was developed. Parametric analyses of various degrees of cavities filling were performed to account for heterogeneity of the debris fill and to obtain a range of likely settlements. Estimated settlements due to seismic shaking ranged from 0.4 to 1.1 percent of total fill thickness with an average of 0.85%.

= 8 mm

= 43 mm

(c) 24 m Compacted Soil Cap 75 m

90 m

105 m

 The predicted settlements from this model were compared to published case histories of seismic settlement of unsaturated fills under earthquake ground motions similar to the design ground motions.

120 m

ρdiff - Diffential settlement over horizontal distance of 9 m - Original grid - Deformed grid (distortion magnified by a factor of 20)

Figure 4. Sample Results from FLAC Analysis

The results of the surficial manifestation analyses, presented as plots of surficial total and differential settlements versus thickness of fill cap (for a range of assumed values of cavities filled by migration of fines and collapse, p), are plotted in Figure 5. This chart was used to select a suitable thickness of removal and replacement based on the differential and total settlement tolerance of the proposed structures. 5

CONCLUSIONS

The seismic and hydrocollapse settlement potential of uncontrolled inert debris fills containing significant oversize clasts could not be evaluated by conventional means. Laboratory testing of representative material was not feasible because of particle size limitations. BPTs and seismic shear wave velocity surveys were ineffective in differentiating well compacted fills from uncontrolled fills. An alternative approach consisted of the following steps:  Based on the results of large scale in-situ density and grain size distribution tests, an upper bound estimate of cavity volume was made (approximately 6.6% of total debris fill volume).

The proposed remedy for rendering the site suitable for building development was partial removal of the uncontrolled debris fill and replacement as a properly compacted fill cap. Based on numerical modeling, charts were developed relating thickness of fill cap to estimated surficial differential settlement. To meet local building code requirement of maximum 25 mm differential settlement over a 9-m length, 22 m of removal and replacement will be necessary. The depth of removal and replacement may be reduced, provided the differential settlement tolerance of the structure is increased by structural improvements such as stiffened foundation systems including mat foundations, post tensioned slabs and grade beams. The reliability of predictions by this approach may be increased by physical modeling of debris fill settlement under the effects of seismic shaking and saturation, and developing a database of observed settlements under moderate seismic events. 6

REFERENCES

AMEC Geomatrix, Inc., 2008. Closure geotechnical report, Reliance II landfill improvements,Vulcan Materials Company, Irwindale. Charles J.A. 2008. The engineering behavior of fill materials: the use, misuse and disuse of case histories. Géotechnique 58 (7), 541-570. Geomatrix, 2007. Documentation of Becker penetration tests, Reliance landfill improvement, Azusa and Irwindale.. Pyke R, Chan C.K. and Seed H.B. 1975. Settlement of sands under multidirectional shaking. J .Geotech. Engrg. Div., ASCE, 101(4), 379-398 Silver M.L. and Seed H.B., 1971. Volume changes in sands during cyclic loading. J. Soil Mech. and Found. Div., ASCE, 97(9), 11711182 Stewart J.P., Bray J.D., McMahon D.J,. Smith P.M., and Kropp A.L. 2001. Seismic performance of hillside fills. Journal of Geotechnical and Geoenvironmental Engineering, 127(11) 905-919

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Site Characterization for the HZM Immersed Tunnel Caractérisation du site pour le tunnel immergé HZM Steenfelt J.S., Yding S., Rosborg A. COWI, Copenhagen, Denmark

Hansen J.G.

Ben C. Gerwick, COWI Group Company, Oakland, USA

Yu R.

COWI China, Beijing, Peoples Republic of China

ABSTRACT: The 36 km long HZM Link, crossing the Pearl River estuary between Hong Kong in the east and Macao and Zhuhai in the west is rated one of the most important current infrastructure projects in China. It is slated for completion in 2016 and consists of a world record length of 6 km immersed tunnel, two artificial transition islands and some 30 km bridges with a dual three lane motorway. In order to provide the structural designers with the requisite input for proper soil structure interaction analysis a very extensive site characterisation was carried out comprising geotechnical boreholes, CPTUs and seismic testing with associated advanced laboratory testing. This paper describes the results and calibration of geotechnical boreholes, CPTUs and advanced laboratory tests to provide the requisite tool for inference of ground stratification and stiffness variation to be used in the structural modelling of the immersed tunnel, the design of piles and dredging slopes. RÉSUMÉ : La liaison HZM de 36 km de long qui traverse l’estuaire du fleuve Pearl entre Hong-Kong à l’est, Macao et Zhuhai à l’ouest, est considéré comme étant l’un des plus importants projets d’infrastructure en Chine. Le projet qui doit être achevé en 2016 est composé d’un tunnel immergé d’une longueur record de 6 km, de deux îles artificielles de transition et d’environ 30 km de pont autoroutier à deux fois trois voies. Afin d’obtenir les éléments essentiels pour l’analyses de l’interaction entre les fondations et les structures, une campagne de sondages géotechniques très détaillée a été menée comprenant des forages, des tests de pénétration au cône (CPTU) et des sondages sismiques ainsi que les études en laboratoire correspondantes. Cet article décrit les résultats obtenus et méthodes de calibration des forages, CPTU et des essais en laboratoire mis en œuvre afin d’obtenir les éléments de base nécessaire pour la détermination des caractéristiques mécaniques des sols à utiliser pour la modélisation des éléments du tunnel immergé, la définition des pieux de fondation et l’étude des pentes de dragage. KEYWORDS: Site characterization, immersed tunnel, CPTU, triaxial testing, undrained shear strength, settlements, spring stiffness. 1

INTRODUCTION

The Hong Kong-Zhuhai-Macao (HZM) Link crosses the Pearl River Estuary in south-eastern China in the Guangdong province connecting Hong Kong at Shek Wan, Lantau Island to the Pearl at Macau and to the district of Gongbei, Zhuhai in mainland China, see Figure 1.

Figure 1. Location of the HZM project in south-eastern China.

The link is 36 km in total length of which 6 km comprises the immersed tunnel. The remainder consists of two artificial transition islands and low bridges some 30 km in total length.

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The whole connection has the capacity of a dual three lane highway. Provisions for two possible future 570 m wide navigation channels are planned along the immersed tunnel alignment with proposed design dredging levels some 15-20 m below existing seabed level. The particular challenges for the design of the immersed tunnel are:  the presence of very soft clays requiring extensive dredging profiles and soil improvement,  very deep foundation level of the tunnel in order to allow for future navigation channels 570 m wide over the central part of the tunnel,  up to 23 m sedimentation load over the central part of the tunnel,  potential of differential settlements due to the highly varying loading and ground stiffness conditions,  the need for mixed foundation solutions with end bearing or settlement reducing piles near the artificial islands and direct foundation for the central part. In order to provide the structural designers with the requisite input for proper soil structure interaction analysis for Detailed Design, a very extensive site characterisation was required. The scope and findings of this site characterisation are described in this paper. The Project Owner is the HZM Bridge Authority, and the design and construction is being undertaken by a Joint Venture headed by the contractor China Communications and Construction Company (CCCC) Ltd.


Figure 2. Simplified geological model along the immersed tunnel alignment.

2

GEOLOGICAL CONDITIONS

The project area is located in the Pearl River drainage basin, which historically has been shaped as a result of the uplift of the Tibetan Plateau during the Tertiary and Quaternary Periods, forming the present-day Pearl River Delta with its network system and estuarine bays (see Figure 1). The river delta is one of the most important and complex large-scale estuarine systems in China. The Holocene development of the delta has been controlled and affected by the variations in the deposition of sediments, sea-levels and groundwater levels. The soil deposits in the present-day Pearl River delta overlying weathered basement rock can be traced back to the Late Pleistocene and Holocene periods. These deposits consist of three cycles of upward fining sequences of delta deposits, namely one Holocene and two Pleistocene delta cycle, which have been divided by two previously exposed and subsequently eroded surfaces. Based on the described regional geology and the findings of the site investigations carried out for the project, the soil deposits and rock formations encountered along the alignment of the immersed tunnel, and in the locations of the artificial islands, can be grouped into five main units for soil deposits, and two main units for rock formations:  Marine deposits of clays and sands formed during the Holocene period,  Continental deposits of clays and sand from a once exposed surface formed during the late Pleistocene period,  Marine alluvial deposits of clays and sands formed during the Mid to Late Pleistocene period,  Fluvial alluvial deposits of clays and sands formed during the Early to Mid Pleistocene period,  Residual soils formed during the Early Pleistocene period,  Highly to completely migmatic schists formed during the Sinian period,  Moderately to completely weathered migmatic granites formed during the Sinian period. A simplified geological model is shown in Figure 2. 3

SCOPE OF INVESTIGATIONS



Supplementary Soil investigations were carried out in 20102011: 80 Nos. boreholes, 364 Nos. CPTUs, 20 Nos. CPTUDs and seismic P-S suspension logging (in 6 Nos. boreholes) was carried out along the alignment of the immersed tunnel and at the locations of the artificial islands.

The Supplementary Soil investigations formed the main basis for Detailed Design, and the scope of and specifications for these investigations were defined by COWI as being a member of the design and construction Joint Venture. Site and laboratory works were followed closely by means of inspections carried out by COWI's geotechnical engineers, in order to ensure that all works were carried out in accordance with applicable standards. The boreholes for the Supplementary Soil investigations were split into two types of boreholes: the GITB-series where geotechnical in-situ testing was carried out and disturbed samples were retrieved, and the TCB-series that were used entirely to retrieve undisturbed samples of fine grained soils. Most of the boreholes were carried out in pairs, each pair consisting of one GITB borehole and one TCB borehole, and as a general rule the GITB and TCB boreholes were drilled within five meters of each other, in order to produce mirror boreholes displaying similar geological and geotechnical properties. The drilling depths varied from 29 to 107 m below existing seabed level. The general distance between boreholes (and borehole pairs) was on average approx. 200 m in the longitudinal direction. In general the CPTUs were carried out along three lines parallel to the tunnel alignment at distances of 0 m, +25 m and -25 m from the tunnel axis. The probing positions were staggered (cf. Figure 3), in order to effectively allow for one CPTU carried out at 25 m spacing along the projected centreline of the entire immersed tunnel alignment. Furthermore, additional CPTUs were carried out near the artificial islands. The CPTUs were carried out to penetration depths varying from 28 to 43 m below existing seabed level (basically to refusal in the fluvial alluvial sands and clays underlying soft deposits of marine clays). A typical arrangement of investigations along the immersed tunnel alignment is shown in Figure 3. The complete results of the Supplementary Soil investigations were provided by the geotechnical sub-contractors, Fourth Harbour Design Institute (FHDI) and Fugro, in native AGS 3.1 format.

Three geotechnical investigation campaigns have been carried out for the project:  Feasibility Study investigations carried out in 2004 and 2008: Only 16 Nos. boreholes were carried out in the vicinity of the immersed tunnel.



Preliminary Design investigations carried out in 2009: 151 Nos. boreholes were carried out for the artificial islands and 115 Nos. boreholes, 29 Nos. CPTUs and seismic P-S suspension logging (in 10 Nos. boreholes) was carried out along the immersed tunnel alignment.

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Figure 3. Typical arrangement of investigations along immersed tunnel alignment.


4

tests in an attempt to quantify and reduce the sample disturbance resulting from sample retrieval, transportation and extrusion. The application of the this initial branch of unloading/reloading conceivably improved the apparent sample quality significantly, as e.g. evaluated in accordance with NORSOK (2004), on average from poor to very good/excellent sample quality. Triaxial testing of fine grained undisturbed samples was carried out as Consolidated Anisotropic Undrained (CAU) triaxial tests in accordance with BS1(1990b). The triaxial tests allowed for site specific calibration of the Nkt cone factor for determination of realistic undrained shear strengths based on CPTUs. Secondly, they allowed the value of su/σ'pc for the normally consolidated condition (often referred to as the c/p ratio) to be determined. In this way a site specific SHANSEP relation could be established allowing determination of the undrained shear strength variation from actual unloading/reloading cycles as a consequence of construction activities.

DRILLING AND IN-SITU TESTING

Drilling for the Supplementary Soil investigations was carried out from five drilling barges equipped with suspended rotary top drive drilling rigs and passive heave compensation. Three different passive heave compensation systems were installed on the five drilling barges used for the investigations:  A strictly mechanical weight load system on one barge,  A spring loaded mechanical system on three barges, and  A hydraulic piston system on one barge. The above mentioned systems were able to be operated with good results (in terms of heave compensation) at maximum heave of approx. 0.7 to 1.0 m. Undisturbed samples (fine grained soils) were primarily retrieved with a 76 mm diameter thin walled stationary piston sampler with stainless steel seamless sampling tubes of length 1.0 m. Undisturbed samples were sealed with wax and taped-shut end caps immediately after retrieval. Storage and transportation were carried out vertically in wooden boxes filled with shock absorbing material (coarse sawdust). SPT testing in coarse grained soils was generally carried out at 1.5 m intervals, and the hydraulic head in the boreholes was as a minimum kept at a level corresponding to sea level. The SPT-N Energy Transfer Ratio (ETR) was determined by carrying out PDA tests of the equipment used from three different barges. In situ shear vane testing was performed at 1 m intervals in fine grained soils using the Chinese electrical vane equipment with cruciform vanes of dimensions 75 mm x 150 mm for the softer clays. CPTU testing was carried out using underwater seabed piezocone penetration systems deployed from barges where the position was maintained by means of 4 heavy anchors. Two different CPTU systems were used, the Wheeldrive Seacalf with 200 kN thrust and the ROSON system with a 100 kN thrust. All CPTU testing was carried out in accordance with the ISSMGE (2001) standard. 5

6

CPTU CORRELATIONS

For the purpose of establishing a detailed geological and geotechnical model of the subsurface conditions, a combination of cored boreholes and closely spaced CPTU soundings was selected as the primary method of investigating the project site. The CPTUs and boreholes were generally carried out as described. The locations of the boreholes were arranged to provide a total of 68 Nos. pairs of boreholes and CPTUs along the alignment. This allowed for a site specific correlation between the stratigraphy as encountered within the boreholes and the corresponding principal CPTU properties with respect to cone resistance, friction ratio and pore pressure. The boreholes and CPTUs carried out in pairs were generally positioned within a 5 m distance from each other. Initially, two approaches were investigated to find the most appropriate correlation model for the site investigation data, namely a conventional method developed by Robertson et al (Lunne et al 1997) and a site specific approach based on pairing the CPTU and borehole data. The depiction of the site CPTU results categorised into the different main geological units and using the Robertson classification chart is shown in Figure 4.

LABORATORY TESTING

Classification testing for the Supplementary Soil investigations consisted of natural moisture content, bulk and dry density, particle density, Atterberg limits, particle size distributions, maximum and minimum dry densities and organic content. Incremental loading (IL) oedometer testing was carried out on both undisturbed fine grained soil samples and reconstituted coarse grained soil samples in accordance with BSI (1990a). The specific schedule for the IL oedometer tests on fine grained samples was designed to take into account the in-situ and pre-consolidation stress together with the anticipated stress history imposed by the construction activities. The maximum net stress increments under the tunnel elements were not expected to lead to exceedance of the in-situ stresses neither along the middle part of the immersed tunnel alignment nor towards the artificial islands. In view of the above, special attention was paid to determine reliable estimates of the values of the pre-consolidation stress and the reloading stiffness. The IL oedometer tests carried out on fine grained samples were performed in two batches:  Batch I IL oedometer tests: Mainly carried out to provide an estimate of the pre-consolidation pressure (and the virgin compression index),  Batch II IL oedometer tests: Carried out to provide an estimate of the reloading stiffness from varying unloading stress levels below the pre-consolidation stress estimated from the Batch I tests. Initial unloading/reloading steps from/to the presumed in-situ stress were included for both the Batch I and II IL oedometer

Figure 4. CPTU results superposed on soils classification chart (Lunne et al 1997).

Instead, the CPTU data were analysed statistically, yielding representative ranges and frequency distributions of each geological unit with respect to cone resistance, friction ratio and excess pore pressure. In this way a unique "foot print" was produced for each geological unit as e.g. shown in Figure 5.

625


(1)

Figure 5. Example of qc frequency distribution "foot print" for Marine Clay.

Based on the frequency distributions, representative ranges were established for the three principal CPTU properties, which in turn were used as filter criteria for a template predicting the geological unit. The pore pressures varied greatly within each geological unit and were not used as a criterion for the geological interpretation, but merely as a guide when visually cross checking the results. The interpretation template set up in this way worked on the premise that if a data set fell within the established "foot print" criteria, the template would subsequently yield the associated geological unit. The criteria were regarded as a key to a typical geological interpretation, not an unambiguous analysis. The final geological interpretation was therefore based on additional factors such as the combined appearance of the qc, Rf and u2 distributions combined with cross referencing to nearby boreholes. Approximately 400 Nos. CPTUs (including those carried out during the Preliminary Design investigations) were interpreted using this method. This allowed for a 3D stratigraphical model to be set up for the geotechnical interpretation of the subsurface conditions surrounding the tunnel alignment, see e.g. Figure 6. 7

GEOTECHNICAL INTERPRETATION

The interpretation of the results of the oedometer tests carried out yielded the modulus number, m, recompression modulus number, mr, secondary compression index, C, secondary recompression index, Cr, coefficient of consolidation, cv and excess preconsolidation pressure, σ'pc (= σ'pc - σ'v0). The use of CPTUs was a key element in the evaluation of the settlement/stiffness variation along the alignment of the Having established the modulus number, m, for a range of soil deposits through laboratory oedometer testing, the modulus modifier, a, can be determined based on the formula:

where qtM is the stress-adjusted cone resistance and σr is a reference stress (=100 kPa). Based on the modulus number from the oedometer tests and the stress adjusted cone resistance from CPTU testing, the modulus modifier, a, was derived or each soil deposit from (1). The modulus modifier is plotted in Figure 7 assessing all oedometer results for fine grained samples. The results shown in this figure indicate relatively little data scatter and a general grouping of fine grained soils around 2 to 5 and 60 to 90 for the coarse grained soils (the latter values are not shown in Figure 7).

Figure 7. Modulus modifier, a, for selected geological units as derived from oedometer and CPTU testing results.

The recompression branch of the oedometer tests on fine grained soils indicated a linear correlation rather than a log-linear correlation. Further, the recompression modulus number, mr, resulting from the reloading branches was found to vary with load for the fine grained soils. A reasonable approximation was achieved by applying different mr values above and below an in situ stress of 100 kPa. The resulting recompression modulus modifier, ar, was therefore defined for in situ stress below and above 100 kPa. Relatively little data scatter was observed in the ar values, with a general grouping of ar values for fine grained soils around 14 to 25 and 14 to 33 for in situ stress above and below 100 kPa, respectively.

Figure 6. Example of contour plot generated based on the compiled 3D stratigraphical model showing top of Continental/Marine Alluvial deposits in the location of the East Artificial Island.

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The SHANSEP concept derives from the empirical observation that the ratio of the undrained shear strength, su, to the effective confining stress, σ'v, is approximately constant for a given Over Consolidation Ratio (OCR) and varies linearly with OCR:

(4) and Eq (5) for fine grained soils below and above the preconsolidation stress:

(2) where S is the proportionality constant (also referred to as the c/p ratio) and  is the memory exponent. These values were estimated from the CAU triaxial testing carried out on undisturbed samples. The S (or c/p-ratio) value was determined based on CAU tests loaded anisotropically to >150% of the assumed preconsolidation stress (as determined from the Batch I IL oedometer tests) and then sheared. The S-value thus determined was used for the determination of the  value for tests loaded anisotropically to below the assumed preconsolidation stress. Due to relatively high uncertainty with regards to the determination of the preconsolidation pressure, the memory exponent was found difficult to determine with accuracy. For the clay deposits found along the alignment of the immersed tunnel average S and  values shown in Table 1 were found.

(5) Here ε is the vertical strain, σ'v is the increase in effective vertical stress from the tunnel (σ'1-σ'0), σ'p is the preconsolidation pressure, σ'0 is the in-situ vertical stress prior to loading, σ'1 is the final vertical effective stress and σ'r is a reference stress of 100 kPa. The secondary settlement was calculated from (Terzaghi et al. 1996): (6) where C is the secondary compression index, and t/tp is the ratio between the lifespan of the structure and the time for primary consolidation (t/tp = 100 was conservatively assumed). When the final load was lower than the preconsolidation stress, the secondary recompression index, Cr, was used instead of C. The calculation of settlement was terminated at the top of rock, and due to the limited penetration of the CPTUs into the fluvial alluvial deposits of sand and gravel, the settlement calculations were based on SPT N data between the bottom of the CPTUs and the top of rock. An empirical qc/N correlation dependent on the grain size distribution was used (Kulhawy & Mayne 1990):

Table 1. Average values of S and  for clay deposits found along the immersed tunnel alignment. Soil deposit

Nos. of tests

Marine clay

2

0.31

0.7

Continental clay

2

0.40

NA

Marine alluvial clay

7

0.31

1.0

Marine alluvial clay with sand laminae

4

0.36

0.7

S (avg.)

avg.)

Notes: NA = Not Applicable

(7)

The results of the CAU triaxial tests were also used to provide a correlation to results of CPTU testing, and thereby for providing an estimate of the Nkt cone bearing factor as used in the following equation (e.g. Lunne et al 1997):

where pa is a reference stress of 100 kPa, d50 is the mean grain size in mm and qc is given in kPa. The spring stiffness was then calculated as: (8)

(3)

The settlement/spring stiffness calculations were carried out in purposefully set up Excel spreadsheets. The settlement/spring stiffness calculations were carried out for some 400 Nos. CPTUs, and considering that each CPTU could contain up to 6,000 measurement points, running the entire series of calculations could take up to 2 hours. The variation of calculated settlement and spring stiffness along the immersed tunnel alignment is shown in Figures 8 and 9, respectively.

where σv0 is the overburden pressure at the cone tip and qt is the cone resistance corrected for pore pressure. For the clay deposits found along the alignment of the immersed tunnel, the Nkt values were found to be 17 on average for the four deposits referenced in Table 1. 8

SETTLEMENT/SPRING STIFFNESS CALCULATION

Based on the geotechnical interpretation of the geology and settlement characteristics of soil deposits, the settlement and spring stiffness was calculated for each individual CPTU location. The settlement analysis was carried out using the Janbu (1963) tangent modulus method, which accounts for the general non-linear load deformation relationship of soils. The settlement equations differ between coarse grained (sandy) and fine grained (clayey and silty) soils, and whether or not the preconsolidation stress is exceeded. All in all four different equations were established. Eq (4) for coarse grained soils below and above the preconsolidation stress:

Figure 8. Calculated settlement along immersed tunnel alignment centre line and lines at 25 m distance from centreline.

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Figure 9. Calculated spring stiffness along immersed tunnel alignment centre line and lines at 25 m distance from centreline.

9

11 REFERENCES

CONCLUSION

The design of the 6 km world record long immersed tunnel with highly variable soil and loading conditions poses significant challenges to both the geotechnical site characterization and the soil-tunnel interaction. The structural tunnel design is very sensitive to differential settlements and rotations of individual tunnel elements and segments and thus to variation is soil stiffness along and across the tunnel alignment. Rather than resolving to empirical rules for handling the soil stiffness variation (Monte Carlo simulation or additional sinusoidal variation around the mean stiffness) the variation was handled directly by the tight mesh of CPTU probing points along and across the alignment. Thus, the CPTUs provided a strong tool for clear geological unit delineation and allowed for very detailed settlement and soil stiffness assessment along the entire tunnel. The CPTU data were correlated with results from oedometer and CAU triaxial test results to provide site specific correlations regarding stiffness and undrained shear strength. The geotechnical site characterization thus facilitated the tool for interaction between geotechnical and structural design of the tunnel elements and allowed for a robust and safe design.

COWI 2011. Hong Kong-Zhuhai-Macao Link Immersed Tunnel Detailed Design, Final Geotechnical Interpretative Report. ISSMGE 2001. International Reference Test Procedure for the Cone Penetration Test (CPT) and the Cone Penetration Test with pore pressure (CPTU). BSI 1990a. British Standard Methods of test for Soils for civil engineering purposes. Part 5. Compressibility, permeability and durability tests, BS1377:Part 5:1990. BSI 1990b. British Standard Methods of test for Soils for civil engineering purposes. Part 8. Shear strength tests, BS1377:Part 8:1990. NORSOK 2004. Standard. G-001. Rev. 2. Marine Soil Investigations. Lunne, T., Robertson, P.K., Powell, J.J.M. 1997. Cone Penetration Testing in Geotechnical Practice, First Edition. Massarsch, K.R., Fellenius,B.H. 2002. Vibratory compaction of coarse grained soils. Canadian Geotechnical Journal 39, 695-709. Janbu, N. 1963. Soil compressibility as determined by oedometer and triaxial tests. III European conference on soil mechanics and foundation engineering, Wiesbaden, Vol. 1, pp. 19-25 and Vol. 2, pp. 17-21. Terzaghi, K, Peck, R.B., Mesri, G. 1996. Soil Mechanics in Engineering Practice, Third Edition. Kulhawy, F.H. and Mayne, P.W. 1990. Manual on estimating soil properties for foundation design. EPRI EL-6800, Cornell University.

10 ACKNOWLEDGEMENTS The authors gratefully acknowledge the permission by COWI to publish the paper.

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris2013

Controversial and Contradictory Evaluations in Analyses of Ground Vibrations from Pile Driving Évaluations controversées et contradictoires par suite de l'enfoncement de pieux

dans

l'analyse

des

vibrations

de

terre

Svinkin M.R. VIBRACONSULT, Cleveland, USA

ABSTRACT: Pile driving operations are powerful and wide-spread sources of construction vibrations which may detrimentally affect adjacent and remote structures, make obstacles for operating sensitive processes and devices, and disturb people. A number of factors can affect ground vibration from pile installation. Wave propagation from pile driving is a complicated problem, and different approaches are utilized to analyze this phenomenon. A paper presents several controversial and contradictory issues in assessment of ground vibrations generated by pile driving such as connections between wave propagation in piles and ground vibrations, the relationship between pile impedance and intensity of ground vibrations, effects of the hammer energy on ground vibrations and a possible correlation between pile capacity and ground vibrations. Also, it is important to show the inadequate roles of condition surveys of structures and monitoring of ground vibrations and the necessity to properly assess crack changes in condition surveys. Analysis and clarification of various approaches are important for practical applications. RÉSUMÉ: Les opérations d’enfoncement de pieux sont des sources puissantes et très répandues de vibrations de construction qui pourraient affecter de façon nuisible des structures adjacentes et éloignées, faire obstacle à l’utilisation de procédés et d’appareils sensibles, et perturber des gens. De nombreux facteurs peuvent affecter la vibration du sol par suite de l'installation de pieux. La propagation d’ondes causée par l’enfoncement de pieux est un problème compliqué, et diverses méthodes sont utilisées pour analyser ce phénomène. Cet article présente plusieurs points controversés et contradictoires dans l’évaluation des vibrations de sol générées par l’enfoncement de pieux, telles que des connections entre la propagation des ondes dans les pieux et les vibrations de sol ; les rapports entre l'impédance des pieux et l’intensité des vibrations de terre ; des effets de l’énergie percutante sur des vibrations de sol, et une corrélation possible entre la capacité des pieux et les vibrations de sol. De plus, il est important de montrer le rôle inadapté des conditions de surveillance des structures, du contrôle des vibrations de sol et la nécessité d'évaluer les modifications des fissures par les opérations de contrôle. L'analyse et la clarification de diverses méthodes sont importantes pour des applications pratiques. KEYWORDS: pile driving, ground vibrations, stress wave theory, impedance, energy, survey correlation between them is possible. For sure, both variables are dependent on the hammer energy. On the one hand approximate calculation of expected ground vibrations and even vibration monitoring yield relative information on vibration effects on structures, and these results could be inconclusive. On the other hand condition surveys of structures before, during, and after pile driving provide complete information on structural responses to vibration excitations and this information can be much beneficial than vibration assessment and measurements. Clarifications of different ways used for analyses of pile driving as the source of construction vibrations, ground vibrations generated by pile driving and various effects of these ground vibrations on structures, people and sensitive devices are important to understand the problem and prevent harmful consequences of pile driving operations.

1. INTRODUCTION Installation of driven piles creates soil vibrations and displacements which may affect adjacent and remote structures, people and sensitive equipment. Therefore, various approaches are used for evaluation of vibration effects of pile driving. There is a trend to connect stress-wave propagation in piles during pile driving with prediction or calculation of the peak particle velocity (PPV) of ground vibrations from pile installation. However, there are ambiguous problems in using of this approach for assessment of ground vibrations. Pile impedance affects force and velocity at the pile head in opposite ways at the same time. Therefore, the pile impedance effect on the intensity of ground vibrations is not obvious. Pile driving generates ground vibrations due to the hammer energy applied to a pile, but some case histories demonstrate no correlations between the hammer energy and the maximum velocity of ground vibrations. Other factors such as the depth of pile penetration into the ground and soil resistance to pile driving should be taken into account. The relationship between pile capacity and ground vibrations is not clear. Moreover, pile capacity and ground vibrations are outcomes of pile driving and only an accidental

2. STRESS-WAVES PROPAGATION IN PILES AND GROUND VIBRATIONS For about forty years, the stress-wave theory is successfully used for driveability analysis of driven piles and also for determination of pile capacity at the time of testing, for example Proceedings of IS-Kanazawa 2012 (2012). In recent years, there is a trend to connect stress-wave propagation in

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piles during pile installation with prediction or calculation of PPV of ground vibrations generated by pile driving, Robertson (2006) and Massarsch & Fellenius (2008). The first such an attempt was made by Svinkin (1996) in favor of the Impulse Response Function Prediction (IRFP) method for prediction of ground and structure vibrations from pile driving; the method was developed toward prediction of complete time-domain vibration records on existing soils, buildings, and equipment prior to installation of impact machine foundations, Svinkin (2002). In the application of this method to pile driving, wave equation analysis was used to assign a movement of the pile top, but it’s necessary to underline that the top pile movement can be assigned arbitrarily, for example as a damped sinusoid, because ground vibrations at some distance from a dynamic source depend only on the dynamic force transmitted on the machine support and soil properties, Svinkin (2002). It is necessary to point out that a connection of the stresswave theory with ground vibrations from pile driving has few problems. First, there are several different programs for signal matching techniques which produce diverse results for the same piles and their outcomes depend on variety of soil conditions and pile types, Svinkin (2012). It is not clear what software should be used. Second, stress waves in piles obviously generate internal forces in driven piles. Third, according to Saint Venant’s principle, wave propagation in piles does not affect dynamic field at some distance from a driven pile. It is known that impact hammers for pile driving and forge hammers released comparable amounts of the energy and they generate similar vibration records of ground vibrations (Steffens 1974). Therefore, it is reasonable to compare both dynamic sources and their effects on ground vibrations. A forge hammer foundation is considered as a rigid body which transfers impacts loads from a hammer onto the ground. Dynamic forces in the machine foundation itself are internal forces generated by stress-waves propagated in the machine foundation under forge hammer impact. The duration of internal forces is substantially smaller than the duration of dynamic forces transferred from a machine foundation onto the ground, and these two kinds of dynamic forces work in different time frames. Consequently, internal dynamic forces in hammer foundations are not taken into account in determination of dynamic loads transferred from a hammer foundation on the ground and consideration of ground vibrations generated by oscillations of forge hammer foundations (Barkan 1962 and Richart et al. 1970). In prediction of ground vibrations from operating forge hammers, ground vibrations depend on the impulse dynamic load applied to a hammer foundation, the damping coefficient and the natural frequency of vertical foundation oscillations, and also the impulse response functions of the considered dynamic systems. The latter represent the soil medium where wave propagate from the hammer foundations to destination locations. The experimental studies showed that at some distances from the source, ground vibrations become dependent only on the impulse load transmitted to a hammer foundation and the soil medium where waves disseminate from the source (Svinkin 2002). These results are in agreement with a dynamic version of Saint Venant’s principle (Timoshenko & Goodier 1951and Karp & Durban 1997). A similar picture of a dynamic load transfer from a forge hammer on its foundation and the ground can be represented for pile installation. Piles also can be considered as rigid bodies in which stress-waves propagate from hammer ram impacts and generate internal forced in piles which are the causes of pile

movement and vibrations. Besides, a pile-soil load transfer is released by means of both concentrated loads from the pile toe and distributed loads generated along pile shaft. Similarly to hammer foundations, at some distances from a pile, as the dynamic source, ground vibrations become dependent only on the dynamic load applied to a pile and the soil medium where waves propagate from the source. It is known that velocities of wave propagation in piles are about 4000 m/s in concrete piles and about 5100 m/s in steel ones (PDA 1991). Velocities of shear wave propagation in the ground are shown in Table 1. Velocities of surface waves are equal about 0.92-0.96 of the velocities of shear waves, Barkan (1962). Table 1. Velocity of shear waves in soils, Savinov (1979) Soil Velocity m/s 120 – 150 Sand Sand with gravel 150 – 250 Loess with natural 130 – 160 moisture Plastic clay 150 - 400 It can be seen that that wave propagation in piles under impact load is much faster process in comparison to wave propagation in the ground. Therefore, dynamic loads transferred from driven piles onto the ground for practical purposes can be considered as the point impulse load at some distance from the source, Svinkin (2000). It can be expected that this conception is correct at distance derived from an assumption that the time of surface wave propagation with velocity, cs, in the ground at distance, D, from a driven pile is 5-10 times larger than the time of stress wave propagation with velocity, c, in the pile with length, L (Svinkin 2000).

 D (5  10)Lcs / c

(1)

Minimum distances from a driven concrete pile as the point vibration source are shown in Table 2 (coefficient 10 was used). Table 2. Minimum distance from pile as point vibration source Pile Length m 10 15 20 30 40

150/4000 Lcs/c 10Lcs/c m m 0.375 3.75 0.5625 5.63 0.8438 8.44 1.125 11.25 1.6875 16.88

cs/c Lcs/c m 0.75 1.125 1.5 2.25 3.0

300/4000 10Lcs/c m 7.5 11.25 15.0 22.5 30.0

It can be expected at distances determined by equation (1), that only dynamic forces transferred to piles during pile driving and soil medium where waves propagate from driven piles will affect ground vibrations generated by pile driving. It is important to point out that calculation of expected ground vibrations during the time of pile installation is irrelevant. For example, Massarsch & Fellenius (2008) tried to connect stress-wave propagation in piles under the hammer ram impact with ground vibrations, but they eventually suggested the old empirical equation to calculate attenuation of PPV of ground vibrations generated by surface waves, which contain more than 2/3 of the total vibration energy, from pile installation without any connection with the stress-wave theory. Ground vibrations have to be measured during pile driving operations.

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris2013

3. PILE IMPEDANCE

4. HAMMER ENERGY

During pile installation, an impulse load from the hammer ram is applied to the pile top, and dynamic longitudinal force in the pile is transferred to the surrounding soil. According to Peck et al. (1974) and Woods (1997), pile impedance affects the force transmitted down the pile. Pile impedance characterizes the pile ability to overcome the soil resistance to pile penetration and develop required capacity. One of pile impedance, Z, definition can be presented as

Pile installation generates ground vibrations due to the hammer energy applied to a pile. Obviously, PPV of ground vibrations have to be a function of the hammer energy transferred on a pile. However, some case histories demonstrate no correlations between the hammer energy and PPV of ground vibrations, Hope and Hiller (2000). It happens due to the effects of soil conditions, the pile penetration depth, and the soil resistance to pile penetration into the ground. Nevertheless, the hammer energy is the major cause of ground vibrations because without the hammer energy there are no pile penetration into the ground and ground vibrations.

Z  EA / c

(2)

where E is modulus of elasticity of pile material; A is pile cross-section area; and c is longitudinal stress wave velocity. It can be seen that impedance depends only on the pile material and dimensions. Recognizing the importance of pile impedance for assessment of to the ground vibration, Heckman and Hagerty (1978) proposed the equation for the peak particle velocity of ground vibrations from pile driving as a function of the rated hammer energy, Wr , and the distance, D, from a driven pile with the coefficient, k, which is dependent on pile impedance.

vk

Wr D

(3)

The coefficient, k, is inversely proportional to pile impedance. It means that driven piles with higher impedance generates lower PPV of ground vibrations and vice versa. Svinkin (2000) derived equations for PPV of pile vibrations, V, and the maximum force, F, measured at the pile head as

V

2cWt ZL

(4)

F

2cZWt L

(5)

and

where Wt is the energy transferred to a pile. Similarly to equation (3), equation (4) shows that the velocity triggered by the hammer ram impact is an inversely proportional function of pile impedance. However, equation (5) displays that the force is proportional to the root square of pile impedance. It means that pile impedance affects force and velocity at the pile head in opposite ways. Case histories presented in a number of publications, for example Svinkin (2000), demonstrate higher ground vibrations triggered by installation of high soil displacement piles (concrete piles and steel pipes with closed ends) in comparison with low soil displacement piles (H-piles and steel pipes with open ends). A practical experience is the evidence that pile impedance affects ground vibrations in the proximity of driven piles, but this pile property does not affect the dynamic field at some distance from driven piles in accordance with Saint Venant’s principle.

5. PILE CAPACITY AND GROUND VIBRATIONS Some authors, for example Robinson (2006), found enormous scatter of PPV of ground vibrations as a function of the hammer energy. For example, PPV of ground vibrations changed between about 0.4-21.6 mm/s at the rated energy of 135 kJ and between about 0.9-17.8 mm/s at the transferred energy of 40 kJ. It happened because other factors mentioned above affected ground vibrations and in consequence that data measured at various construction sites with different soil conditions, pile types and pile driving implementations were considered together. However, Robinson (2006) suggested a correlation between ground vibrations and pile capacity determined during pile driving. He believes that pile-soil interaction, not energy, is the major influence in the generation of ground vibrations from driven piles. Obtained conclusions are not accurate because ground vibrations and pile capacity are outcomes of the same pile driving process and only an accidental correlation between them is possible. It is necessary to say that ground vibrations and pile capacity for sure depend on the hammer energy because pile capacity cannot be mobilized without the sufficient hammer energy. Moreover, during pile driving, the static pile capacity is determined by signal matching software on the basis of force and velocity measurements at the pile head. Unfortunately, different software produces different results. It means that PPV of ground vibrations are dependant on signal matching technique used for analysis of testing data. Besides, during pile installation, ground vibrations should be measured not calculated because of possible detrimental effects of pile driving operations and also measured ground vibrations are more reliable than calculated ones. 6. CONDITION SURVEYS AND VIBRATION MEASUREMENTS Approximate calculation of expected ground vibrations and even vibration monitoring yield relative information on vibration effects on structures, and these results could be inconclusive. Moreover, there is uncertainty in application of the existing vibration limits for assessment of pile driving effects on soils and structures. Therefore, it is imperative to perform condition surveys of structures before, during and after pile installation which provide complete information on structural responses to vibration excitations. Obtained information can be much beneficial for analysis of causes of damage to structures than vibration assessment and measurements. Dowding (1996) pointed out the necessity of professional performance of a preconstruction survey. Condition surveys during pile installation and after the completion of pile driving are significant for analysis of possible causes of damage to structures. Each construction site

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is unique and even similarity of soil deposits does not mean the same condition of the dynamic settlement development. Physical evidences of damage to structures from dynamic sources are very important. Therefore, much attention is provided for measurement of crack width at condition surveys of structures during pile driving. Changes of crack dimensions are the major evidences of vibration effects on structures. Micrometers are used to determine changes of crack widths. It is necessary to keep in mind that each structure has its own “breathing” because of microseisms in the earth and human activities inside and outside structures. Hence, it is typical that crack widths may vary in time. If crack widths increase without increasing of crack lengths, it is a safe situation. However, if variations of crack widths trigger increasing of crack length, it becomes dangerous for structures. Thus, changes of crack widths alone are not the indicators of damage to structures from pile driving. Therefore, it is necessary to measure crack widths together with assessment of crack length enlargements. 7. CONCLUSIONS Ground vibrations from pile driving may harmfully affect structures, people and sensitive devices, and these effects should be evaluated before and during pile driving operations. The paper presents several controversial and contradictory issues in assessment of ground vibrations generated by pile driving. Analyses of various approaches are important for practical applications. A connection of the stress-wave theory with ground vibrations from pile driving has few problems. There is no unique solution of stress-wave propagation in the pile because different signal-matching software provides different outcomes. Internal forces in piles may somewhat affect ground vibrations in the proximity of the pile. However, according to Saint Venant’s principle, wave propagation in piles does not affect dynamic field at some distance from a driven pile. Pile impedance affects ground vibrations in the proximity of driven piles, but this pile property does not affect the dynamic field at some distance from driven piles in accordance with Saint Venant’s principle. Pile installation generates ground vibrations due to the hammer energy applied to a pile. Missing correlation between PPV of ground vibrations and the hammer energy in some case histories occurred on account of the effects of soil conditions, the pile penetration depth, and the soil resistance to pile penetration into the ground. Nevertheless, the hammer energy is the major cause of ground vibrations because without the hammer energy there are no pile penetration into the ground and ground vibrations. Pile capacity and ground vibrations are outcomes of the same pile installation and only an accidental correlation between them is possible. Condition surveys should be performed before, during and after pile driving. Assessment of crack length enlargements has to accompany measurements of crack widths because changes of crack widths alone are not the indicators of damage to structures from pile driving. Clarification of different views on the problems would be helpful in practice for assessment of pile driving effects on surrounding structures.

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8. REFERENCES Barkan, D.D. 1962. Dynamics of Bases and Foundations. New York: McGraw Hill Co. Dowding, C.H. 1996. Construction Vibrations. Prentice Hall, Upper Saddle River. Heckman, W.S. & D.J. Hagerty 1978. Vibrations associated with pile driving. Journal of the Construction Division, 104(CO4): 385-394. ASCE. Hope, V.S. and Hiller, D.M. 2000. The prediction of groundborne vibration from percussive piling. Canadian Geotechnical Journal, 37, 700-711. Karp, B and Durban, D. 1997. Towards a dynamic version of Saint Venant’s principle. Modern Practice in Stress and Vibration Analysis. M.D. Gilchrist (Ed.): 251-255. Rotterdam: Balkema. Massarsch, K. R., and Fellenius, B. H. 2008. Ground vibrations induced by impact pile driving. SOAP 3, Proceedings of the Sixth International Conference on Case Histories in Geotechnical Engineering: 1-38, Arlington, Virginia: OmniPress. PDA (1991). Pile Driving AmalyzerTM - Manual. Pile Dynamic, Inc. Cleveland, Ohio. Peck, R.B., Hanson, W.E. and Thornburn, T.H. 1974. Foundation Engineering, 2nd ed., New York: John Wiley & Sons, Inc. Proceedings of IS-Kanazawa 2012. Testing and Design Methods for Deep Foundations. Proceedings of the 9th International Conference on Testing and Design Methods for Deep Foundations, Kanazawa, Japan, 18-20 September 2012. Robinson, B.R. 2006. Models for Prediction of Surface Vibrations from Pile Driving Records. A thesis submitted in partial fulfillment of the Requirements for the degree of Master of Science, NC State University. Richart, F.E., Hall, J.R. and Woods, R.D. (1970). Vibrations of soils and foundations, Prentic-Hall, Inc., Englewood Cliffs, NJ. Savinov, O.A. 1979. Modern construction of machine foundations and their calculations. Second Ed. Stroiizdat, Leningrad. Steffens, R.J. 1974. Structural vibration and damage. Building Research Establishment Report, HMSO. Svinkin, M.R. 1996. Overcoming soil uncertainty in prediction of construction and industrial vibrations. Proceedings of Uncertainty in the Geologic Environment: From Theory to Practice, C.D. Shackelford, P. Nelson, and M.J.S. Roth (Eds.), Geotechnical Special Publications No. 58, ASCE, 2: 1178-1194. Svinkin, M.R., Roth, B.C. and Hannen, W.R. 2000. The effect of pile impedance on energy transfer to pile and ground vibrations. Proceedings of the Sixth International Conference on the Application of Stress-Wave Theory to Piles. S. Niyama &J. Beim (Eds.): 503-510, Rotterdam: Balkema Svinkin, M.R. 2002. Predicting soil and structure vibrations from impact machines. Journal of Geotechnical and Geoenvironmental Engineering., 128(7): 602-612. ASCE. Svinkin, M.R. 2012. Engineering evaluation of static capacity by dynamic methods. Proceedings of the 9th International Conference on Testing and Design Methods for Deep Foundations, Kanazawa, Japan, 18-20 September 2012: 179-186. Kanazawa University. Timoshenko, S.P. and Goodier, J.N. 1951. Theory of Elasticity. New York: McGrawHill Book Co. Woods, R.D. (1997). Dynamic Effects of Pile Installations on Adjacent Structures, NCHRP Synthesis 253, Transportation Research Board, National Research Council, Washington, D.C.

CPT/PCPT- Based Organic Material Profiling Matière organique - Le profilage basé sur le CPT/PCPT Tümay M.T.

Louisiana State University, Baton Rouge, LA, USA and Boğaziçi University, İstanbul, Turkey

Hatipkarasulu Y.

The University of Texas at San Antonio, San Antonio, TX, USA

Marx E.R.

Fugro Consultants, Inc., Baton Rouge, LA, USA

Cotton B.

Fugro Consultants, Inc., Kenner, LA, USA ABSTRACT: Cone and Piezocone Penetration Test (CPT and PCPT) based analysis and modeling is a popular and handy tool for geotechnical engineers for subsurface investigations and soil characterization. However, effective identification and extent of organic content proves to be a challenge based on traditional CPT and PCPT data and methodologies. This paper presents a comprehensive CPT/PCPT-based organic content identification method using Zhang and Tumay (1999) probabilistic soil classification method. The probabilistic method employs a non-traditional modeling approach that takes the uncertainty of correlation between the soil composition and soil behavior into account. The method is based the conformal mapping of the Douglas and Olsen (1981) classification chart which results in the soil classification index (U) and in-situ behavior index (V). The organic content identification method proposed in this paper uses the in-situ behavior index (V) in combination with the compositional soil classification index (U) to estimate the organic content. A detailed description of the proposed methodology and a discussion of effective applications are included in the paper. RÉSUMÉ : Le pénétromètre quasi-statique et le piézocône (CPT et PCPT) constituent des outils d’analyse populaires et pratiques pour la reconnaissance géotechnique des sites et la caractérisation des sols. Cependant, l’identification des sols organiques ainsi que l’évaluation de la teneur en matière organique à partir des données classiques obtenues au CPT et au PCPT se sont révélées être un challenge. Cet article présente une méthode d’évaluation complète de la teneur en matière organique basée sur le CPT et la méthode de classification des sols probabiliste de Zhang et Tumay (1999). Cette méthode probabiliste utilise une approche de modélisation non conventionnelle qui prend en compte l’aléa sur la corrélation entre la composition du sol et son comportement mécanique. La méthode est basée sur l’abaque de classification des sols de Douglas et Olsen (1981) qui permet de définir l’indice de classification des sols (U) et l’indice de comportement in situ (V). La méthode d’évaluation du contenu en matière organique proposée dans l’article utilise l’indice de comportement in situ (V) combiné à l’indice de classification (U) pour estimer la teneur en matière organique du sol. Une description détaillée de la méthodologie proposée et une discussion de ses applications sont aussi présentées dans l’article. KEYWORDS: CPT, PCPT, Cone Penetration, Soil Classification ,Organic Soils 1

analyses of two well-documented test sites to illustrate the effectiveness of CPT/PCPT-based profiling and their correlation to laboratory test results.

INTRODUCTION

Cone and Piezocone Penetration Test (CPT and PCPT) (ASTM D5778-12) based analysis and modeling is a popular and handy tool for geotechnical engineers for subsurface investigations and soil characterization. Since the 1960s, several modeling approaches have been developed for soil classification and evaluation of different soil properties such as the strength and consolidation characteristics of geomedia. However, effective identification and extent of organic content has proved to be a challenge based on traditional CPT and PCPT data and methodologies. Although some models identify organic materials as a separate soil class, they do not provide a continuous profile (for example, Schmertmann, 1978; Robertson et al, 1986; Robertson, 1990). Considering the likelihood of having different levels of organic content in any soil type, a continuous profile will provide additional understanding and evaluation of the subsurface. This paper presents a comprehensive CPT/PCPT-based organic content identification method using Zhang and Tumay (1999) probabilistic soil classification method. The organic content identification method proposed in this paper uses the Zhang and Tumay method’s in-situ behavior index (V) in combination with the compositional soil classification index (U) to estimate the organic content. A detailed description of the proposed methodology and a discussion of effective applications are included in the paper. The paper also presents

2

CPT-BASED PROBABILISTIC SOIL CLASSIFICATION

Unlike the traditional chart-based two-dimensional classification methods, the Zhang and Tumay method uses a probabilistic region estimation method to address the uncertainty in misclassifying the soil layers. This statistical based method provides a profile of the probability or the chance of having each soil type (clayey, silty, and sandy) with depth. This method is similar to the classic soil classification methods which are based on soil composition. The probability of incorrectly identifying soil type using the tradition CPT classification charts, especially in transition zones, motivated the development of the probabilistic region estimation method. This CPT classification method addresses the uncertainty of correlation between the soil composition and soil mechanical behavior. In the Zhang and Tumay probabilistic method, conformal mapping was performed on the Douglas and Olsen (1981) chart to transfer the chart axis from the CPT data (qc, Rf) to the soil classification index (U) and in-situ behavior index (V). The conformal transformation is accomplished using the following equations:

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x  0.1539R f  0.8870 logqc   3.35

(1)

 y 0.2957R f  0.4617 logqc   0.37

(2)

3

The soil classification index (U) and in-situ behavior index (V) are given as: U 

a1x  a2 y  b1 c1x  c2 y  d1  c1x  c2 y  d1 2  c2 x  c1 y  d 2 2 a2 x  a1 y  b2 c2 x  c1 y  d 2  c1x  c2 y  d1 2  c2 x  c1 y  d 2 2



 V 10 



(3)

(c1x  c2 y  d1)(a2 x  a1 y  b2 )

(c1x  c2 y  d1 )2  (c2 x  c1 y  d 2 )2 (a1x  a2 y  b1 )(c2 x  c1 y  d 2 )

(c1x  c2 y  d1 )2  (c2 x  c1 y  d 2 ) 2

ORGANIC MATERIAL PROFILING

The proposed profiling method utilizes a combination of the soil classification index and the in-situ behavior index values where the [(V-U) > 3.0] indicates significant organic content. The organic content indicator (V-U) makes it possible to profile the changes in organic content continuously while acknowledging the likelihood of having organic material in any given soil composition. To illustrate the proposed method, two well documented data sets from the Mississippi River Long Distance Sediment Pipeline study located near Barataria Waterway in Jefferson Parish, Louisiana are considered. Tables 1, 2, and 3 present the laboratory test results for data sets B-7 and B-28 including moisture content percentage (ASTM D2216-10), Atterberg Limits (ASTM D4318-10), and material content finer than No. 200 sieve (ASTM D1140-06). Table 4 presents the organic content percentages (ASTM D2974-07) for the same data sets. Table 1. Laboratory testing values for B-7 data set.

(4)

The coefficients in equations 3 and 4 are defined as: a1=-11.345, a2=-3.795, b1=15.202, b2=5.085, c1=-0.269, c2=-0.759, d1=-2.960 and d2=2.477. A statistical correlation was then established between the U index and the compositional soil type given by the Unified Soil Classification System (USCS) (ASTM D2487-11). A normal distribution of U was established for each reference USCS soil type (GP, SP, SM, SC, ML, CL, and CH). Each U value corresponds to several soil types with different probabilities. Boundary values were used to divide the U axis into seven regions as described in Figure 1. Soil types were further rearranged into three groups: sandy and gravelly soils (GP, SP, and SM), silty soils (SC and ML) and clayey soils (CL and CH). Figure 1 also gives the probability of having each soil group within each region. The original method gives constant probability of each soil type (represented by the step lines) regardless of the U value within the same region (R1 to R7 in Figure 1). This allows for the sudden drop in the probabilities as the U value crosses the border from one region to another. This method was further modified to allow smooth transition of probability (curved lines) with U values, and hence to provide a continuous profile of the probability of soil constituents with depth.

Depth (m) 1.2 3.0 4.9 9.1

Depth (m) 2.1 2.8 4.3 16.8

Data Set B-7 B-28

SC, ML CL, CH

Probability (%)

70

50

Data Set B-7

40 30 20

0

Plasticity Index % 98 16 67 58

Moisture Content % 72 81 80 50

Liquid Limit % 44 78 86 69

Plastic Limit % 20 24 28 23

Plasticity Index % 24 54 58 46

Depth (m) 1.2 9.1 2.1 3.7 4.2 6.1 8.8 10.7 16.5

Material Passing No.200 Sieve % 89 100 96 27 98 87 27 12 100

Table 4. Organic content values for B-7 and B-28 data sets.

60

10

Plastic Limit % 35 22 32 27

Table 3. Material passing No. 200 sieve for B-7 and B-28 data sets.

GS, SP, SM

80

Liquid Limit % 133 38 99 85

Table 2. Laboratory testing values for B-28 data set.

100 90

Moisture Content % 125 42 93 74

R6

R7 -0.14

R5 0.61

R4 1.33 U Value

R3 2.01

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Figure 1. Regional boundaries and the corresponding probabilities of each soil group.

In-situ behavior index (V) provides a profile of soil behavior and, in combination with the compositional soil classification index (U), estimates of soil organic content and of soil rigidity/stiffness (indirectly OCR) can be determined (Tümay et.al, 2012).

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Depth (m) 0.6-0.9 1.8-2.1 3.0-3.4 0.6-1.2 2.4-2.8 3.7-4.0 7.2-7.6

Ash Content % Organic Content % 85.42 14.58 96.85 3.15 96.85 3.15 85.95 14.05 84.78 5.22 91.73 8.27 97.10 2.90

Figures 2 and 3 illustrate the CPT sounding values (qc, fs and Rf) (ASTM D 5778-12), Zhang and Tumay (1999) probabilistic soil classification soil results with U and V index values, organic content indicator (V-U), and corresponding lithology obtained from the boring logs for B-7 and B-28 data sets.


TIP RESISTANCE (qc)

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-3.0 3.0 9.0 Inorganic | Organic

%

% Cl a y % Si l t % Sa nd

FAT CLAY (CH), very soft, dark gray

FAT CLAY (CH), soft to firm, gray, with silt seams and lenses

18.0 %0

%50

%100

%0

%50

%100

Figure 2. CPT-Based organic material profile and lithology – B-7 Data Set.

TIP RESISTANCE (qc)

SLEEVE FRICTION (fs)

FRICTION RATIO (Rf)

U INDEX

Organic Content (V-U)

V INDEX

ZT99 Probabilistic

Lithology

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ORGANIC CLAY (OH), % Cl a y very soft, dark gray, % Si l t with roots, peat, and % Sa nd shells LEAN CLAY (CL), w/organics FAT CLAY (CH), w/organics ORGANIC CLAY (OH)

4.0

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%50

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Inorganic | Organic

Figure 3. CPT-Based organic material profile and lithology – B-28 Data Set. As illustrated in Figures 2 and 3, the organic content indicator (V-U) identified the significant organic content in the sample data sets while providing a continuous profile. When this information is combined with the CPT-based soil classification, it provides a better understanding of the subsurface conditions. For example, laboratory testing values indicate a fine grained soil with high plasticity index (98%) for

the first meter of the B-7 data set where the organic content test resulted in 14.58%. The test values show a significant drop after 2.0 meters for the plasticity index (16%) and the organic content (3.15%). This profile change is clearly illustrated in Figure 2. Similarly, as shown in Figure 3, the test results show organic content over 5% for the B-28 data set between 2.5 and 4.0 meters. This value decreases to 2.90% at 7.2 meters.

635


The data capture frequency of the CPT soundings (2 cm/sec) also allows for the identification of the thinner seams of sandy and silty sand layers as well as the increases in the organic content. For example, the soil classification of the B-28 data set between 8.0 and 12.0 meters shows thin layers of clayey materials. At these layers, the organic content appears to be higher that the surrounding silty sand. 4

CONCLUSION

Although there are several CPT-based soil classification models effectively used in subsurface investigations, accurate identification of organic materials using CPT soundings provides a challenge. The organic content indicator (V-U) proposed in this study offers a continuous profile for the organic content based on the soil classification and in-situ behavior indexes as defined by the Zhang and Tumay’s probabilistic method. For the examples provided in this paper, the organic content indicator shows a strong agreement with the test results and [(VU) > 3.0] indicates an approximate level of 5% organic material. The CPT-based indicator may provide a useful tool considering the importance of identifying organic materials which could lead to problems concerning stability, compaction, compressibility and usability. Soil classification effectiveness and accuracy of the Zhang Tumay (1999) method has been evaluated with several data sets under different conditions and test sites, and compared with other classification charts, for subsoil stratigraphy (Tümay et. al., 2011), multi model comparison (Hatipkarasulu and Tümay, 2011(1)), tip resistance value correction (Tümay and Hatipkarasulu, 2011), effective area ratio assumption (Hatipkarasulu and Tümay, 2011(2)), and data reduction effects (Tümay and Hatipkarasulu, 2012). The method uses a nontraditional approach which takes into account the probability of having each soil type with depth. The use of soil classification (U) and in-situ behavior (V) indexes for organic profiling extends the same concept of accounting for the chance of having organic material at any given depth. The organic content indicator (V-U) profiling shows reliable performance in clayey materials and its further evaluation is imperative for non-clay organic soils. 5

ACKNOWLEDGEMENTS

This study rests on the interpretations of the field and laboratory investigations conducted for the Mississippi River Long Distance Sediment Pipeline Project designed and undertaken by the Louisiana Department of Transportation and Development (LADOTD). The contents of this paper reflect the views of the authors, who are responsible for the facts and the accuracy of the data presented herein, and do not necessarily indicate official assessments of the agencies, firms and institutions with which the authors are affiliated. 6

REFERENCES

ASTM D5778-12. 2012. Standard Test Method for Electronic Friction Cone and Piezocone Penetration Testing of Soils. ASTM International, West Conshohocken, PA. DOI: 10.1520/D5778-12. http://www.astm.org/Standards/D5778.htm ASTM D2487. 2011. Standard Practice for Classification of Soils for Engineering Purposes (Unified Soil Classification System). ASTM International, West Conshohocken, PA. DOI: 10.1520/D2487-11. http://www.astm.org/Standards/D2487.htm ASTM D2216-10.2010. Standard Test Methods for Laboratory Determination of Water (Moisture) Content of Soil and Rock by Mass. ASTM International, West Conshohocken, PA. DOI: 10.1520/D2216-10.

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http://www.astm.org/Standards/D2216.htm ASTM D4318-10. 2010. Standard Test Methods for Liquid Limit, Plastic Limit, and Plasticity Index of Soils. ASTM International, West Conshohocken, PA. DOI: 10.1520/D4318-10. http://www.astm.org/Standards/D4318.htm ASTM D2974-07a. 2007. Standard Test Methods for Moisture, Ash, and Organic Matter of Peat and Other Organic Soils. ASTM International, West Conshohocken, PA. DOI: 10.1520/D2974-07. http://www.astm.org/Standards/D2974.htm ASTM D1140-00. 2006. Standard Test Methods for Amount of Material in Soils Finer than No. 200 (75-μm) Sieve. ASTM International, West Conshohocken, PA. DOI: 10.1520/D1140-06. http://www.astm.org/Standards/D1140.htm Douglas, J. B., and Olsen, R. S. 1981. Soil Classification using Electric Cone Penetrometer, Symposium on Cone Penetration Testing and Experience, Geotechnical Engineering Division, ASCE, St. Louis, MO, USA, (1981), 209-227. HatipKarasulu, Y. and Tumay, M.T. 2011(1). Practical visual presentation approach for CPT-based soil characterization and modelling. Proceedings of the ASCE Geo-Frontiers Conference: Advances in Geotechnical Engineering, Dallas, TX, pp. 2387-2396 Hatipkarasulu, Y. and Tümay. 2011(2). Impact of effective area ratio assumption on PCPT-based soil classification. Proceedings of the Fourth International Conference on Site Characterization, ISC’4: Geotechnical and Geophysical Site Characterization, Porto de Galingas – Pernambuko, Brazil. pp. 275-282 Robertson, P. K., Campanella, R. G., Gillespie, D., and Greig, J. 1986. Use of Piezometer Cone Data. Proceedings of the ASCE Specialty Conference on In Situ’86: Use of In Situ Tests in Geotechnical Engineering, Blacksburg, Virginia, pp. 1263-1280. Robertson, P. K. 1990. Soil Classification using the Cone Penetration Test. Canadian Geotechnical Journal, Vol. 1, No. 27, pp. 151-158. Schmertmann, J.H. 1978. Guidelines for Cone Penetration Test, Performance and Design. Report No. FHWA-TS-78-209, U.S. Tümay, M. T. and HatipKarasulu, Y. 2011. Impact of Using Measured v. Corrected Tip Resistance Values in PCPT-Based Soil Characterization and Modeling, Proceedings, of the ASCE GeoFrontiers Conference: Advances in Geotechnical Engineering, ASCE Special Publication No. 211, Dallas, TX, pp. 2544-2553. Tümay, M. T., HatipKarasulu, Y., Młynarek, Z., and Wierzbicki, J. 2011. Effectiveness of CPT-Based classification methods for identification of subsoil stratigraphy. Proceedings of the 15th European Conference on Soil Mechanics and Geotechnical Engineering, Athens, Greece, pp.91-98. Tümay, M. T. and HatipKarasulu, Y. 2012 Effects of data smoothing and reduction on CPT-based probabilistic soil classification. Proceedings of the Fourth International Conference on Site Characterization, ISC’4: Geotechnical and Geophysical Site Characterization, Porto de Galingas – Pernambuko, Brazil. pp. 843850 Tümay, M.T., Hatipkarasulu, Y., Marx, E.R and Cotton, B. 2012. Multi Model Subsurface Evaluation for Louisiana I-10 Bridge Replacement Using Electronic CPT. Proceedings of the Fourth International Conference on Site Characterization, ISC’4: Geotechnical and Geophysical Site Characterization, Porto de Galingas – Pernambuko, Brazil. pp. 1281-1288 Zhang, Z., and Tumay, M.T. 1999. Statistical to Fuzzy Approach toward CPT Soil Classification. ASCE Journal of Geotechnical and Geoenvironmental Engineering, Vol. 125, No. 3, pp. 179 186.

Geotechnical Challenge for Total Cost Reduction related to Construction of Connecting Bridge with Pile Foundations Défi géotechnique pour la réduction totale des coûts liés à la construction du pont de liaison avec les fondations sur pieux Yasufuku N., Ochiai H.

Kyushu University, Fukuoka, Japan

Maeda Y.

West Nippon Expressway Company Limited, Osaka, Japan

ABSTRACT: Changes of geotecnical engineering profile are briefly mentioned based on the density of in-situ investigations and laboratory tests. Then, the method used for evaluating the vertical bearing capacity of driven piles in the actual design is presented.. The applicability is also verified by comparing the predicted results with the results from the full-scale pile load tests, whose results were linked with the reduction of the safety factor for design. Finally, the significance of geotechnical investigations including in-situ and laboratory tests and full scale pile load tests are discussed in terms of the cost performance of the construction of pile foundations for supporting the connecting bridge. It is concluded that in-situ and laboratory investigation with reasonable geotechnical considerations can reduce the total cost of the construction of the bridge with pile foundations for New-Kitakyushu airport. RÉSUMÉ : Dans ce papier, la politique de base et des concepts pour des études géotechniques et de conception fondation sur pieux du pont qui relie pour la Nouvelle-Kitakyūshū sont introduits. Les changements de profil géotechnique sont brièvement mentionnés basés d’après la densité du terrain (in-situ) et des essais au laboratoire. Ainsi, la méthode utilisée pour l’évaluation de la capacité portante des pieux battus conçu selon la méthodologie actuelle est présentée sur la base des considérations géotechniques. L'applicabilité est également vérifiée en comparant les résultats prédits avec les résultats des essais en vraie grandeur de chargement de pieux. Les résultats ont été comparés en termes de réduction du facteur de sécurité utilisé au dimensionnement. . Enfin, l'importance des études géotechniques y compris les essais in situ et en laboratoire et les essais en vrai grandeur de chargement de pieux sont discutés en termes de performance des coûts de la construction des fondations sur pieux pour soutenir le pont de liaison. KEYWORDS: cost reduction, field investigations, pile foundations design, bearing capacity 1

INTRODUCTION

A connecting bridge has been constructed on the sea as an access road for New Kitakyushu airport, which will be opened in 2005. The length of the bridge is about 2km and 24 piers are mounted for supporting the bridge. An overview of the connecting bridge under construction is shown in Figure 1. In order to clarify the geological and mechanical characteristics of the ground for supporting the bridge and the manmade airport island, a large number of in-situ and laboratory tests had been performed for five years from 1991 to 1995. In this paper, the basic policy and concepts for geotechnical investigations and design of this project are introduced. The changes of geotecnical engineering profile are briefly mentioned based on the density of in-situ investigations and laboratory tests. The process of producing a model ground for design is also made clear, which is used for estimating the bearing capacity of driven piles. Further the method used for predicting the vertical bearing capacity of driven piles is presented based on the geotechnical considerations. The applicability is also verified by comparing the predicted results with the results from the full-scale pile load tests, whose results are linked with the reduction of the safety factor for design. Finally, the significance of in-situ investigations and full scale pile load tests are discussed in terms of the cost performance of the construction of pile foundations for supporting the connecting bridge. 2

Connecting Bridge （2.1km, 24 piers）

Airport site

Aug. 2002

Figure 1. Overview of connecting bridge under construction Design of foundation (Reassessment)

(Feedback)

Assessments based on Geotechnical considerations Investigation

(Collaboration)

design

Select of possible models Decision of sort and number of field & lab. tests Implementation of site investigation

Verification by site investigations

• Full scale load tests • Field observations

Modeling of ground

• Careful selection of soil parameters • Determination of adequate model

Reconsideration of safety factors et al.

GEOTECHNICAL INVESTIGATIONS AND DESIGN

Figure 2 shows the policy and concept of geotechnical investigation and design for constructing the connecting bridge

637

Implementation of rational and Economical design in total

Figure 2. Collaboration of geotechnical investigations with design

Elevation (m)


TP 　0 -1 0 -2 0 -3 0 -4 0 -5 0 -6 0 -7 0 -8 0 -9 0

P12

P1 Ac D Lc

P22

A2 Ac

Ac

DUs

DUc

DLs

S chw S ch

(c ) 1 9 9 4 -9 5

D Lc D Ls

T tc

Unconformity

L a n d a rea

D Lc S ch

T tc

A irp o rt a r ea

S ea a r ea

T 0 -1 -2 -3 -4 -5 -6 -7 -8 -9

Embankment Ac

Dutf

DLs1 DLc1

DUs DUc DLs2-5

Figure 3. Final geotechnical engineering profile obtained DLc2-6 N-blow values 0 -10

Evelation,T.P.

-20

30 50

100

150

N-blow values 200 0 20 50

a) DLs layer

100

150

200

b) DLc layer

Bearing stratum

-50

Figure 5. Model geotechnical engineering map for design

Figure 3 clarified that 1) the investigated ground consists of alluvial clayey layers with 7-9m thickness and Pleistocene layers laminated by sandy and clayey soils with 20-60m thickness below the alluvial layers and also weathered crystalline schist as the base layer. The corresponding ground is therefore roughly divided into 3 layers. 2) The undulation of base layer is extremely high in which the difference becomes more than 45m. 3) The structure of Pleistocene layers is complicated and the continuities in horizontal direction are not so clear, and so the lens shape layers are found here and there. 4) The surface of unconformity in Pleistocene layers is clearly found from sea area to airport area of which inclination is about 15 degrees in the longitudinal direction.

Measured N-values

-30 -40

Sch

Averaged N-values Range of dispersion

-60

Figure 4. Distribution of N-values in DL layer against depth

for New-Kitakyushu airport. As shown in this figure, the field and laboratory investigations and the engineering design are conducted based on the clear policy, which includes that: 1) The strong collaboration between geotechnical investigators and designers should be made for a rational design and construction in pile foundations. 2) The design parameters should be determined based on the geotechnical considerations, which reflect the results obtained from the geotechnical investigations and laboratory soil tests. The model for estimating the bearing capacity of piles in design should be based on the geotechnical considerations. 3) A rational bearing stratum should be carefully selected based on the geological and geotechnical investigations. 4) The predicted performance in design should be checked by a full-scale model tests as much as possible. The results are reflected to the reduction of factor of safety for design. Such policy seems to be strongly linked with the performance based design, which may become the mainstream in foundation design near future. 3 GEOTECHNICAL ENGINEERING MAP FOR DESIGN REFLECTED THE SOIL PROPERTY 3.1 Geological profile with increases of site investigation Figure 3 shows the final geotechnical engineering profile mainly by the field investigations from 1992 to 1995, which covers the land, sea and airport areas. Figure 3 was drawn by adding the boring data in each pier of the access road, where the total number of borings became more than 65 with 3500m in total length, and the geological investigations on the diatom earth and also volcanic ash deposit with the results of the seismic exploration. The boring densities of each area in 1992, 1993 and 1995 are roughly 350m, 180m and 70m respectively. It is judged that the geotechnical engineering profile becomes more precise with the increasing boring density and quality of in-situ investigations. The accuracy of geotechnical investigations is believed to lead to the economical and rational design and construction, even if the percentage of investigation cost to the total one might be somewhat increased (see Table 2).

638

3.2 Model geotechnical engineering map for design When determining a good bearing stratum for pile foundation, Japanese design code by Japan Road Association recommends that the N-values of sandy or sand-gravel layers are grater than 30 blow counts, and also N-values of clayey layers are more than 20. Figure 4 shows the characteristics of N-values in Pleistocene sandy and clayey layers obtained from the SPT. The N-values of both layers tend to become more than 30 in average when the depth is roughly deeper than 30m T.P. level. Based on the results, the following guideline for pile foundation design was determined such that: 1) The layer at 30m T.P level was judged as an effective bearing stratum for driving the pile foundation. A steel pipe sheet-pile foundation was selected as a type of pile foundation in this project, where, all of pile tips are set up in Pleistocene laminated ground at around 30m T.P. levels. 2) As shown in Figures 3 and 4, the scatters of N-values seems not to be small and also it is not easier to distinguish from the sandy and clayey layers from N-values obtained because the site consists of the complicated laminated sandy and clayey layers. In this circumstances, the uniform and empirical method based on the N-values is not rational and precise to evaluate the pile bearing capacity. Thus, a method for evaluating the pile vertical bearing capacity should be introduced together with a proper geotechnical engineering map for foundation design, which is derived by geotechnical Table 1. Soil constants of each layer N-value Alluvial clay Pleistocene

Ac

Volucanic Dutf

Sandy Clayey

(Upper)

DUs DUc

Pleistocene (Lower)

* OCR i

t

11.0 30.4 0.0 17.0 40.0 27.0 32.4 47.5

Sch-w Sch

29.7 98.3

M etamorphic rocks l d

d t

Strength parameters

c' ' 'cv OCR* (tf/cm3) (tf/m2) (degs.) (degs.) 0.0 0.53 0.292 0.0 33.0 1

DLs1 DLs2-5 DLc1 Clayey DLc2-6 Gravel DLg Sandy

'

0.66 0.90 0.53 0.90 0.53 0.94 0.97 0.99 0.90 0.90

th d th

0.6 0.0 8.1 2.6 5.5 2.6 4.4 0.0 5.7 5.7

30.0 37.0 24.0 35.4 32.6 34.8 29.6 36.0 22.7 22.7

33.7 34.6

1-6 1-2

36.9 35.7 35.7 35.5 36.4 36.0

1-6 1-2 1-2 2-8 2-8 1

-

1 1


bearing capacity strongly depends on the degree of the blockade effect and thus the precise prediction of the end bearing capacity was considered to be quite difficult. Then, as shown in Figure 6, the skin friction mobilized through the internal face of the pile under the bearing stratum was assumed as the equivalent end bearing capacity in the design. Therefore, the second term qdA is expressed as ULfi.

Steel pile with diameters of 1m Skin friction mobilized Bearing stratum Penetration depth L

4.2 Evaluation of skin friction

Skin Friction mobilized here is assumed as pile-tip resistance Figure 6. Basic idea of pile bearing capacity

considerations based on the results of the large numbers of insitu and laboratory tests. The resultant geotechnical engineering map and the soil constants of each layers as characteristic values are summarized in Figure 5 and Table 1, in which the soil constants are mainly obtained by the standared consolidation and triaxial undrained and drained compression tests. 3) Fullscale pile load tests are conducted to confirm the validity of the predicting method used for foundation design. The possibility of reducing the safety factor for design to 2.5 from 3.0 is considered through the geotechnical point of view based on the field investigations, laboratory test results and the accuracy of the predicting method with full scale pile load tests. 4 EVALUATION OF VERTICAL BEARING CAPACITY OF DRIVEN PILES 4.1 Basic idea Specification for Highway Bridge gives a following equation as an estimating method of the ultimate pile bearing capacity based on the results of the field and laboratory investigations (JRA, 1996): R u  U  Li f i  q d A

(1)

Where Ru: ultimate bearing capacity of pile, A: pile tip area, qd: pile end bearing capacity, U: pile circumference, Li: thickness in each layer, fi: maximum skin friction of pile. The first and second terms are related to the skin friction of pile and pile-tip bearing capacity, respectively. However, the main part of the vertical bearing capacity of a pile is often mobilized from the

20

S3

S4 DLc S5 DLs DLc S6 S7 DLs S8

25 30

Strain gage

15 20

c’and 'are the adhesion and friction parameters between pile and soil, and 'h is the effective lateral stress acting on the pile. 4.2.2 Soil constants as characteristic values An idea that the adhesion between pile and soils is roughly equal to the apparent cohesion of soils c’ is widely used for a practical design. It is mentioned that the applicability of this idea is effective, irrespective of type of soils such as clay and sand (e.g. Tomlinson 1980). Therefore, c’in eq. (4) was assumed to be equal to the apparent cohesion c’ of soils. In practical design, the axial pile capacity is estimated for the settlements of approximately 10% of the pile diameter. The 10% settlements usually exceed those for mobilizing the maximum skin friction of pile. Further, when considering that the mobilized mechanism of skin friction between pile and soils surrounding the pile, it is reasonable to use the friction angles at the critical state corresponding to sufficiently large displacement ’cv as’(Yasufuku et al. 1997). Here ’ is assumed to be conservatively two-third of ’. ’is thus given by 2 (3)  '   ' 3 where, ’: effective friction angle at peak strength state. 4.2.3 Coefficient of lateral effective stress K The mobilization of the skin friction is dependent on the lateral effective stress 'h and thus in turn is dependent on the overburden pressure 'v. When considering 'h is given by K ' v , Eq.(2) is rewritten by

 K 1  sin  'OCR sin  '

25 30

(b)

(2)

(4)

K is a coefficient of lateral effective stress and ’v is vertical effective stress. The coefficient of lateral effective stress K was estimated from the previous research findings related to the K0value. K-values in Pleistocene clayey layers were determined by the following equation (Mayne and Kulhawy, 1982).

10

(a) (a)

 f c'   h' tan '

 f c'  K v' tan '

5

Depth (m)

DLs

(L=29m)

(m) Depth z

S2

(MN)

Driven Pile

Acl

5

15

0

S1

0

10

Axial force N-Values 0 20 40 60 0 2 4 6 8 10

1.0m

P12 site

4.2.1 Basic equation The following basic equation is therefore used for calculating the skin friction of piles which is determined as the sum of pile to soil adhesion and friction components:

(c)

(d)

Figure 7. Soil profile, N-values and measured axial force in pile load test at P12 site

skin friction in practical designs within the limits of allowable displacement, because relatively large displacements are needed to mobilize the end bearing capacity. In addition, as a normal open-end pile is used as a type of pile foundation, the end

639

(5)

where, OCR is over-consolidation ratio defined as the ratio of the consolidation yield stress pc to the overburden pressure 'v. Values of OCR, ’ in average and the calculated K-values in Eq.(5) are measured against elevation. We can say that applying this equation into the Pleistocene clayey layers, most of Kvalues became more than 1.0. Based on the experimental evidence, K-value for design was decided as 1.0, irrespective of type of Pleistocene layers. Thus, the presented model for evaluating the vertical bearing capacity is expressed as Ru  U  Li f i  ULf i (6)


Measured Values

0

2 0

(a) P12

N-Values

2

4

4.6MN

Predicted Values

4

6

Measured Values

Total skin friction

6

8.4MN

8

N-Values

6.5MN 6.4MN

Total bearing capacity

9.2MN

10.0MN

Pile-tip resistance

Pile-tip resistance

Predicted Values

Total bearing capacity

8

9.5MN

Total skin friction

9.19 10 16.2MN

10

(b) P22

Figure 8. Comparison of predicted total bearing capacities with those of pile load tests

2  f i c '  1.0 ' z tan   '   3 

(7)

where, z is an arbitrary depth from the surface and L is a penetration depth from the bearing stratum (see Figure 6).

summarized in Table 2, which is a result of trial calculation. Note that the cost is normalized by the cost obtained by the standard manner for evaluating the pile bearing capacity using N-values (JRA, 1993) without any full scale pile load tests. For comparison, the layer of the bearing stratum for each case was assumed to be same, however, the penetration depth L was considered to depend on the calculation manner. Total cost are divided by 2 parts, in which one is the part for the cost related to the geotechnical investigations which include in-situ and laboratory soil tests, and full scale pile load tests, and the other is related to the normalized total pile construction cost in terms of P1 to P24 piers. The presented manner used here is expected to cut the cost more than 15% comparing with the total cost by the standard approach using N-values. Thus even if the cost of the geotechnical investigations became roughly two times higher comparing with the general manner, the appropriate insitu and laboratory investigation with a reasonable considerations can reduce the total cost in the project. This is due to the highly accurate ground profile and the proper evaluation method of pile bearing capacity with the results of the full scale pile load tests which reflected the decrease of safety factor from 3.0 to 2.5. It is believed that the geotechnical considerations and manner treated here can give an important information for the geotechnical investigators, structural designers and construction engineers.

5 FULL-SCALE PILE LOAD TESTS AND THE REDUCTION OF FACTOR OF SAFETY

Table 2. Total cost benefit

In order to verify the applicability of the presented model and to confirm the characteristics of the pile bearing capacity of each layer, full scale pile load tests were conducted at two representative sites, which locate at 12P and 22P sites shown in Figure 3. As an important engineering judgment in this project, the reduction of the factor of safety from 3.0 to 2.5 for pile foundation design was discussed through comparing the predicted results with the results of full scale pile load tests. Figure 7 shows the soil profiles and N values with depth for 12P site. N-values can be seen to widely change with depth from nearly zero to more than 20 and also N-values at pile tips are roughly 30. The steel piles with a diameter of 1.0m were carefully driven using vibration and hydraulic hummers. The effective length of each pile was about 30m. Tests were conducted based on the multi-cycles method, which is recommended by the JGS (1993). Four strain gauges were located at each of the cross sections as shown by the dots in Figure 7. Figure 8 shows the comparison of the estimating total vertical bearing capacities with those of full-scale pile load tests at 12P and 22P sites, in which Eqs. (6) and (7) was used to calculate the predicted values. The bearing capacity calculated by the empirical model based on the measured N-values recommended by JRA is also depicted in this figure. The model used here can reasonably estimate both total skin friction and pile tip resistance at both sites, comparing with those from JRA recommendation. As shown in Table 1 and Figure 3, we have a clear grasp of the soil characteristics values for each layer and a practically efficient geotecnical profile. Therefore, the model can apply very well to evaluate the pile bearing capacity according to the ground profile at each site, with the consequence that the accuracy of the prediction clearly increased and these facts became an important evidence to reduce the factor of safety for pile foundation design from 3.0 to 2.5. 6

EFFECT OF A REDUCTION IN TOTAL COSTS

The comparison of the cost performance in terms of the construction of pile foundations driven in P1 to P24 sites is

640

M ethod by Nvalues

M ethod proposed here

1

2.11

1 1

0.82

Cost for geotechnical investigations* Construction cost for piles Total cost

0.84

* the cost includes full scale load tests

7

CONCLUSIONS

An importance of integrating the geotechnical investigations with pile foundation design was clarified through a case study in terms of connecting bridge for New-Kitakyushu airport. The following major conclusions were drawn: 1) A rational method for evaluating the pile bearing capacity was presented which reflected the soil characteristic values and geological environmental history. In addition, the applicability of the presented method was confirmed through full-scale pile load tests, with the consequence that the safety factors for pile foundation design were reduced from 3.0 to 2.5. 2) In-situ and laboratory investigation with reasonable geotechnical considerations can reduce the total cost of the construction of the bridge for New-Kitakyushu airport. 8

REFERENCES

JSSMFE Standards for Vertical Load Tests of Piles. 1993. Japanese Geotechnical Society, 113-121. Japan Road Association. 1996. Specifications for Highway Bridge Part IV, 330-337(in Japanese). Mayne P.W. and Kulhawy F.H. 1982. K0-OCR relationships in soils, J. Geotech. Eng. Div., ASCE, 108(GT6), 851-872. Ochiai H. and Yasufuku N. 2003. Investigation, design and construction of the connecting bridge for New-Kitakyushu airport. 9th Annual Meeting of Chinese Soil Mechanics and Geotechnical Engineering, 1, 214-22

Dynamic CBR as a method of embankment compaction assessment Dynamique CBR comme une méthode d'évaluation de compactage du remblai Zabielska-Adamska K., Sulewska M.J.

Bialystok University of Technology, Bialystok, Poland

ABSTRACT: In engineering practice, earth construction requires suitable soil compaction, usually relating to the Proctor methods. Materials of the built-in embankment and the subgrade have their own specifications, dependant on the kind of earth structure and soil plasticity characteristics. Care should be taken not to use compaction degree (% of maximum compaction) as the only parameter to assess soil compaction. This applies to both cohesive soil and to fly ash whose permeability and mechanical properties dependent on moisture content at compaction. Therefore, for these types of soils California Bearing Ratio could be used as a method of compaction assessment being an indicator of soil bearing capacity. The CBR research was done for both static (classic) and dynamic methods on fly ash samples without soaking them to replicate field conditions. A load of 2.44 kPa was applied to all the samples subjected to penetrations. The dynamic CBR tests were conducted by using Light Weight Deflectometer consisting of a falling weight to produce a defined load pulse of the CBR piston. The CBR test could be used for running compaction control during embankment erection, which specially refers to dynamic CBR test due to the speed of research execution. RÉSUMÉ: Dans la pratique d’ingénierie, la construction en terre nécessite un compactage du sol adapté, concernant en général les méthodes Proctor. Les matériels encastrés du remblai et de la plate-forme ont leurs propres spécifications, dépendant du genre de la construction en terre et de caractéristiques de plasticité du sol. Il faut prendre soin de ne pas utiliser le degré de compactage (% de compactage maximum) comme le seul paramètre pour évaluer la compactage du sol. Cela s’applique aux sols cohésifs et à cendres volantes dont la perméabilité et des propriétés mécaniques dépendent de la teneur en humidité au compactage. Donc, pour ceux types de sol l'indice portant californien pourrait être utilisé comme une méthode d’évaluation du compactage étant un indicateur de la capacité portante. Les recherches CBR ont été effectuées pour les méthodes statiques (classiques) et dynamiques sur les échantillons de cendres voltantes sans les faire tremper à reproduire les conditions de terrain. Une charge de 2,44 kPa a été appliqué à toutes échantillons soumis à des pénétrations. Les tests de dynamique CBR ont été effectués a l’aide de déflectomètre constitué par la masse tombante pour produire une impulsion de charge définie du piston CBR. Le test CBR pourrait être utilisé pour exécuter le contrôle du compactange lors de l’érection de remblai, qui se réfère en particulier à l'essai dynamique de CBR en raison de la rapidité d'exécution de la recherche. KEYWORDS: compaction, California Bearing Ratio (CBR), dynamic CBR (CBRd), fly ash, compaction assessment. 1

were done by using impact generator and guide rod, which are the parts of Light Weight Deflectometer (LWD), and additional equipment in a CBR piston. A falling weight is to produce a defined load pulse of the CBR piston that can be used both in laboratory and field tests. The aim of this study was to prove that CBR tests could be used as the methods of road embankment or subgrade compaction assessment. This refers especially to CBRd test which may be used for running compaction control during embankment erection due to the speed of research execution, as well as Light Weight Deflectometer (Sulewska 2012).

INTRODUCTION

In engineering practice, earth construction requires suitable soil compaction, usually relating to the Standard and Modified Proctor methods. Materials of the built-in road embankment and the subgrade have their own specifications, dependant on the kind of earth structure and soil plasticity characteristics. Care should be taken not to use compaction degree (% of maximum compaction) as the only parameter to assess compaction of material in embankments. This applies to both cohesive soil and fly ash. The permeability and mechanical properties of compacted fly ash are dependent on moisture content present during compaction, as are properties of cohesive mineral soils (Turnbull and Foster 1956, Mitchell et al. 1965, ZabielskaAdamska 2006 and 2011). Consequently different values of geotechnical parameters are obtained for water content on either side of the optimum water content on the compaction curve, for the same dry densities. Thus for these types of soils California Bearing Ratio, CBR, may be used as a method of compaction assessment, since it is an indicator of ground bearing capacity broadly used in the design of civil engineering. The laboratory CBR tests by means of both static (classic) and dynamic methods were carried out to establish relationship between bearing ratio and fly ash compaction. Samples, compacted by the Standard or Modified Proctor methods, were prepared without soaking them to replicate field conditions during earth structure erection. The dynamic CBR, CBRd, tests

1

LITERATURE REVIEW

California Bearing Ratio, CBR, is expressed as the percentage ratio of unit load, p, which has to be applied so that a standardized circular piston may be pressed in a soil sample to a definite depth with a rate of 1.25 mm/min and standard load, corresponding to unit load, ps, necessary to press the piston at the same rate into the same depth of a standard compacted crushed rock.

p  100% (1) ps CBR value is used for evaluation of the subgrade or subbase strength, and may be applied to assess the resistance to failure  CBR

641


or indicate the load-carrying capacity. It should be noted here that CBR values in pavement design do not reflect the shear stresses that are generated due to repeated traffic loading. The shear stress depends on many factors; none of them is fully controlled or modelled in CBR test (Rico Rodriquez et al. 1988, Brown 1996). In laboratory, CBR penetration test is performed on material compacted in a specified mould and placed in loading machine equipped with a movable base that rises at uniform rate used in forcing the penetration piston into the specimen. Tested specimens are penetrated directly after compaction or are to be previously soaked. CBR test in-situ is carried out with a mechanical screw jack for continuous increase of the applied load to the penetration piston. A reaction forcing the penetration piston into the soil is provided by a lorry equipped with a metal beam and attachments under its rear. The dynamic CBR, CBRd, test can be performed both in laboratory and in situ. The test can be conducted as an alternative to the static CBR test, especially due to the short period of time required. CBRd advantage, compared with the classic CBR, is the elimination of a loading frame necessary in static loading. The CBRd test is carried out with the use of Light Weight Deflectometer, where a falling weight is used to generate a defined load pulse on the CBR piston. CBRd is calculated on the basis empirical formula (Zorn 2002) as: 87.3 (2) (%) s 0 . 59 where 87.3 is the number standing as a value of dynamic loading including empirical coefficient, and s is the settlement in millimetres. CBRd is recommended to specify when it is greater or equalled 20% and is equalled or lower than 150%. Turnbull and Foster (1956) carried out broad studies on CBR for compacted mineral soils. They determined penetration resistance of unsoaked samples of lean clay, compacted by means of four different energy values and at different moisture contents. It was proved that the CBR value for compacted clay is a function for both water content as well as dry density. Compacted samples reached higher CBR values when higher energy values were applied. Moisture increase of compacted samples decreased CBR value and in cases of compacted samples with moisture contents greater than optimum water content, penetration resistance was close to zero. Soaking of samples caused the decrease of CBR value, quite significant in compacted samples – dry of optimum, less significant at optimum water content. The smallest decrease was observed in samples compacted at wet of optimum. Rodriguez et al. (1988) described CBR dependence on compaction parameters– moisture contents and dry densities, as well as on conditions of compaction– energy and methodology of compaction. The authors point to the fact that the CBR value of the soil compacted with higher energy value may be lower than that resulting from the compaction with lower energy value. CBR dependence on moisture in the process of compaction was confirmed in the course of studies conducted by Faure and Viana Da Mata (1994). The authors straightforwardly claim that dry density resulting from the compaction of a sample does not have any impact on CBR value which, on the other hand is influenced by moisture present in the process of compaction. CBR’s relationship with moisture content was also observed in the case of compacted marl from Saudi Arabia (Aiban 1995), where marl was subjected to tests at moisture optimum and moisture on the dry and wet sides of optimum. Moisture– density curves and CBR(w) dependency curves were said to be similar; the highest CBR values were obtained at optimum moisture. The studies of the samples tested immediately after compaction and the soaked samples confirmed that the effect of soaking is decreased when the samples are compacted at moisture greater than optimum. Zabielska-Adamska (2006 and 2011) concluded that the highest CBR values for unsoaked samples of fly ash (class F) CBR 

642

appear in modified compaction – in case of moisture level below optimum, and in standard compaction – in case of moisture level within or slightly below optimum. In saturated samples, the highest values for bearing ratio CBR are present in moisture level equal optimum for both compaction energy levels. Once optimum moisture is exceeded, CBR value drops dramatically, regardless of the compaction energy and method of preparation of samples, soaked or unsoaked. High moisture results in the loss of contact among fly ash grains. Hence CBR value dependence on moisture level of fly ash is quite apparent. CBR of samples compacted by means of modified method for optimum moisture is almost twice as high than in the case of optimum compaction by standard method, which points to a significant influence of compaction energy and dry density. It is interesting how compaction energy influences CBR in samples of the same level of moisture, compacted, however, with the use of different energies. Ash samples with moisture value w, compacted by Proctor modified compaction, where w>wopt1, show far lower CBR than samples of the same moisture level w, but compacted by standard method where w
LABORATORY TESTS

All the tests were conducted on the basis of fly ash from hard coal burning in Bialystok Thermal-Electric Power Plant, stored at a dry storage yard. The fly ash shipment corresponded in graining to sandy silt. Physical parameters are shown in Tab. 1. The laboratory CBR tests were carried out to establish relationship between bearing ratio and fly ash compaction. The tested samples were compacted by two methods: the Standard Proctor and the Modified Proctor at moisture contents within the range of wopt±5% for each compaction method. The fly ash samples were saturated 24 hrs prior to the test so that their moisture content could increase by approx. 2.5%. After that, they were deposited in sealed containers. Each compaction curve point was designated on a separate sample. During the compaction tests, individual samples of fly ash were used only


once, otherwise they could not be regarded as representative (Zabielska-Adamska 2006). The CBR tests were conducted on unsaturated samples. All the samples subjected to penetration, tested both methods – static and dynamic, and were loaded with ASTM 1883-73 recommended load of 2.44 kPa. The static (classic) CBR research was done on fly ash samples directly after compaction. Higher CBR value was accepted as a result

D50 (mm)

 s (g/cm3)

CU= D60/D10

CC=D302/D60·D10

0.055–0.065

2.11±0.01

3.89–4.25

0.94–1.03

Modified Proctor method

Standard Proctor method

wopt1 (%)

 d max1 (g/cm )

wopt2 (%)

 d max2 (g/cm3)

37.00

1.068

45.50

1.009

3

Table 1. Geotechnical parameters of tested fly ash shipment. Tested fly ash parameters

Figure 1. CBR research (from the left): static test; changed mould basis and prepared mould extension for dynamic CBR; specimen ready for dynamic test (photo by Zabielska-Adamska).

calculated on the basis of pressing piston resistance, represented in a given depth – 2.5 or 5.0 mm. Next, after levelling off the surface of the same specimen and replacement of basis of the mould, dynamic CBR was carried out. The CBRd tests were conducted using Light Weight Deflectometer (LWD) consisting of a falling weight (7.07 kN) vertically movable along the guide rod to produce a defined load pulse (3.6 MN/m2) of the CBR piston. Electronic measurement system gauged the depth of the piston’s penetration in the tested soil after a single impact. CBR tests are shown in Figure 1. Figure 2 represents the results of standard and dynamic CBR testing, depending on moisture content at compaction, in relation to compaction curves of fly ash, obtained by means of two Proctor methods. Static CBR results confirm earlier results obtained by the author. CBR of unsaturated samples of fly ash reaches the highest values in the case of samples compacted at the moisture content lower than optimum. The samples compacted above optimum water content have still lower CBR values simultaneously with an increase of moisture content. These relationships can be observed in both methods of compaction – standard method and modified method. However, samples compacted with the use of modified Proctor method, the curve CBR(w) definitely reaches maximum. The shape of the curves CBRd(w) is similar to that obtained according to the standard method – CBR(w). In the case of modified compaction, curves CBRd(w) and CBR(w) are characterised by a similar scope of moisture content; from wopt1–5% to optimum moisture content, wopt1 (difference in relation to CBR – up to about 2.5%). Once curve CBRd(w) exceeds wopt1, it also exceeds standard curve, passing CBR by 16% at wopt1+5%. In the case of standard compaction, at moisture level wopt2–5%, CBRd value equals CBR value. After this, as the moisture content increases the difference also increases and when the moisture level is equal to wopt2, the CBR difference is exceeded by 5%.

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CBR MP = –1026.64 + 59.09w – 0.82w2 CBRd MP = – 467.44 + 27.46w – 0.37w2 CBR SP = – 297.17 + 16.96w – 0.22w2 CBRd SP = – 519.42 + 25.88w – 0.30w2

2

(R =0.8751) 2 (R =0.7768) 2 (R =0.8047) 2 (R =0.8468)

Figure 2. CBR test results versus moisture content at compaction in comparison with compaction curves: MP – Modified Proctor method, SP – Standard Proctor method, CBR – static test results, CBRd – dynamic test results.

With further increase of moisture content, the difference may be as high as 13%. Significant differences in the results of the studies carried out by means of static and dynamic methods, at moisture level exceeding wopt originate from the differences in speed of loading and lack of possibility of pore pressure


embankment. The results of studies of CBRd, and CBR, are closely connected with the characteristics of compaction. 2. The current compaction quality control of fine grained anthropogenic ground conducted through CBRd tests with the use of Light Weight Deflectometer producing a defined load pulse of the CBR piston is recommended in the cases of embedded material at moisture contents equal optimum or lower. CBRd studies of anthropogenic ground compacted at moisture levels exceeding optimum water content may lead to overstating of the test results due to lack of pore pressure dissipation after impact ground loading. 3. Dynamic CBR test, using Light Weight Deflectometer, should be widely used due to its speed and ease of research as an alternative method to classic method of quality control in compaction process or assessment of subgrade bearing capacity.

dissipation in the case of impact loading. Similar observations can be made during studies on the influence of penetration ratio on the resistance of saturated clayey soils in cone penetration tests (Kim et al. 2008). Figure 3 presents dependence of static and dynamic CBR on dry density. It can be seen in Figure 3 that there are points standing out, with the coordinates (ρd, CBR) obtained in the case of standard method at moisture content higher than optimum by at least 2.5%, and in modified method higher by at least 5%. This is the result of dependence of mechanical parameters of fly ash on moisture content in the process of compaction. Once these points are excluded, statistically valid relationships CBR(ρd) can appear, especially in the case of CBRd values, where for value CBRd(ρd) coefficient of determination R2=0.8675 was obtained (Fig. 4). CBRd dependence on CBR is also statistically valid. Equation CBRd=17.28+0.52CBR explains 84.9% of variance in the value of statistic CBR. 4

ACKNOWLEDGEMENTS

This work, carried out in 2012 at the Bialystok University of Technology, was supported by Polish financial resources on science. The authors gratefully acknowledge the assistance and cooperation of M. Piasecki and D. Tymosiak who performed the laboratory tests. 5

dynamic

static

Figure 3. Relationship between CBR value and dry density with an indication the points obtained at moisture contents at compaction w=wopt+(2.5–5%): MP – Modified Proctor method, SP – Standard Proctor method, CBR – static test results, CBRd – dynamic test results.

Figure 4. CBR value versus dry density excluding the points obtained at moisture contents at compaction w=wopt+(2.5–5%), along with 95% confidence interval.

3

CONCLUSIONS 1. The dynamic CBR method, as well as static (classic) method can be used to assess compaction of fly ash and cohesive soils embedded in subgrade or layers of

644

REFERENCES

Turnbull W.J. and Foster Ch.R. 1956. Stabilization of materials by compaction. Journal of Soil Mechanics and Foundations Division 82 (SM2), 1-23. Mitchell J.K., Hooper D.R., Campanella R.G. 1965. Permeability of compacted clay. Journal of Soil Mechanics and Foundation Division 91 (SM4), 41-63. Zabielska-Adamska K. 2006. Fly ash as a material for constructing sealing layers. Publishing House of BTU, Bialystok. (in Polish) Zabielska-Adamska K. 2011. Fly ash as a barrier material. In: Geo‐Frontiers 2011 “Advances in Geotechnical Engineering”, ASTM STP 211, ASTM Int., PA, 947-956. Sulewska M.J. 2012. The control of soil compaction degree by means of LFWD. Baltic Journal of Road and Bridge Eng. 7(1), 36-41. Brown S.F. 1996. Soil mechanics in pavement engineering. Géotechnique 46 (3), 383-426. Zorn. 2002. Operating Manual. Light Drop-Weight Tester ZFG 05 for the dynamic CBR test and the dynamic plate loading test. Gerhard Zorn Mechanische Werkstatten, Stendal. Rico Rodrigues A., del Castillo H. and Sowers G.F. 1988. Soil mechanics in highway engineering. Trans Tech Publication, Clausthal-Zellerfeld. Faure A.G. and Viana Da Mata J.D. 1994. Penetration resistance value along compaction curve. Journal of Geotechnical Engineering 120 (1), 46-59. Aiban S.A. 1995. Strength and compressibility of Abqaiq marl, Saudi Arabia. Engineering Geology 39 (3-4), 203-215. Zabielska-Adamska K. and Sulewska M.J. 2009. Neural modelling of CBR values for compacted fly ash. In: Proc. 17th Intern. Conf. on Soil Mechanics and Geotechnical Engineering, Eds. M. Hamza, M. Shahien, Y. El-Mossallamy, IOS Press - Millpress, Vol. I, 781-784. Weingart W., Hanebutt J. and Rummert W. 1990. Dynamic laboratory and field testing device for determination of the CBR value of mineral concrete. Die Strasse 26 (2), 48-51. (in German) Schmidt H.-H. and Volm J. 2000. Dynamic CBR test – new method of embankment quality control. Geotechnik 23 (4), 271-274. (in German) Kim K., Prezzi M., Salgado R. and Lee W. 2008. Effect of penetration rate on cone penetration resistance in saturated clayey soils. Journal of Geotechnical and Geoenvironmental Eng. 134 (8), 1142-1153.

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