FHWA NHI-07-071 Earth Retaining Strucutures June 2008 (Without Solutions)

U.S. Department of Transportation Federal Highway Administration

Publication No. FHWA NHI-07-071 June 2008

NHI Course No. 132036__________________________________

Earth Retaining Structures Reference Manual

National Highway Institute

NOTICE The contents of this report reflect the views of the authors, who are responsible for the facts and the accuracy of the data presented herein. The contents do not necessarily reflect policy of the Department of Transportation. This report does not constitute a standard, specification, or regulation. The United States Government does not endorse products or manufacturers. Trade or manufacturer's names appear herein only because they are considered essential to the objective of this document.

Technical Report Documentation Page 1. REPORT NO.

FHWA-NHI-07-071

2. GOVERNMENT ACCESSION NO.

4. TITLE AND SUBTITLE

3. RECIPIENT'S CATALOG NO.

5. REPORT DATE

Earth Retaining Structures

June 2008 6. PERFORMING ORGANIZATION CODE

7. AUTHOR(S)

8. PERFORMING ORGANIZATION REPORT NO.

Burak F. Tanyu, Ph.D. and Paul J. Sabatini, P.E., Ph.D, and Ryan R. Berg, P.E. 9. PERFORMING ORGANIZATION NAME AND ADDRESS

Ryan R. Berg & Associates, Inc. 2190 Leyland Alcove Woodbury, MN 55125 12. SPONSORING AGENCY NAME AND ADDRESS

National Highway Institute Federal Highway Administration U.S. Department of Transportation Washington, D.C.

10. WORK UNIT NO. 11. CONTRACT OR GRANT NO.

DTFH61-06-D-00019 13. TYPE OF REPORT & PERIOD COVERED 14. SPONSORING AGENCY CODE

15. SUPPLEMENTARY NOTES

This manual is the updated version of FHWA NHI-99-025 manual prepared by Parsons Brinckerhoff Quade & Douglas, Inc., authored by G. A. Munfakh, N. C. Samtani, R. J. Castelli, and J. Wang. This updated version presents design procedures for walls, except in-situ reinforced walls, using the load and resistance factor design (LRFD) method. FHWA COTR – Louisa M. Ward FHWA Technical Consultants – J. DiMaggio, B. Siel, and D. Alzamora. 16. ABSTRACT This is the reference manual for FHWA NHI course No. 132036 on Earth Retaining Structures. Detailed information on subsurface investigation, soil and rock property design parameter selection, lateral earth pressures for wall system design, and load and resistance factor design (LRFD) for retaining walls are provided. Wall types discussed include gravity and semi-gravity walls, modular gravity walls, MSE walls, nongravity cantilever walls and anchored walls, and in-situ reinforced walls. Information on wall system feasibility and selection, construction materials and methods, cost information, design and performance information, and required elements of construction inspection are presented for each wall type. Different contracting approaches for the various wall systems are also discussed.

lateral earth pressures, LRFD, gravity and semi-gravity walls, modular gravity walls, mechanically stabilized earth walls, sheet pile walls, soldier pile and lagging walls, slurry walls, tangent/secant walls, anchored walls, soil nailing, micropiles, wall selection, and contracting approaches.

18. DISTRIBUTION STATEMENT

19. SECURITY CLASSIF.

20. SECURITY CLASSIF.

21. NO. OF PAGES

Unclassified

Unclassified

17. KEY WORDS

No restrictions.

764

22. PRICE

---

Symbol

SI CONVERSION FACTORS APPROXIMATE CONVERSIONS FROM SI UNITS When You Multiply By To Find Know

Symbol

LENGTH mm m m km

millimeters meters meters kilometers

0.039 3.28 1.09 0.621 AREA

inches feet yards miles

in ft yd mi

mm2 m2 m2 ha km2

square millimeters square meters square meters hectares square kilometers

0.0016 10.764 1.195 2.47 0.386

square inches square feet square yards acres square miles

in2 ft2 yd2 ac mi2

VOLUME ml l m3 m3

millimeters liters cubic meters cubic meters

0.034 0.264 35.71 1.307 MASS

fluid ounces gallons cubic feet cubic yards

fl oz gal ft3 yd3

g kg tonnes

grams kilograms tonnes

ounces pounds tons

oz lb tons

EC

Celsius

0.035 2.202 1.103 TEMPERATURE 1.8 C + 32 WEIGHT DENSITY

Fahrenheit

EF

kN/m3

kilonewton / cubic meter

poundforce / cubic foot

pcf

6.36

FORCE and PRESSURE or STRESS N kN kPa kPa

newtons kilonewtons kilopascals kilopascals

0.225 225 0.145 20.9

poundforce poundforce poundforce / square inch poundforce / square foot

lbf lbf psi psf

PREFACE This document is the update of the 1999 NHI course manual for NHI Course 13236 – Module 6, Earth Retaining Structures. This document has been written to provide up-todate information on earth retaining systems currently being constructed in the United States for highway applications. Earth retaining systems discussed in this manual include: • • • • •

Rigid gravity and semi-gravity walls; Prefabricated modular gravity walls; Mechanically stabilized earth (MSE) walls; Non-gravity cantilevered and anchored walls including sheet-pile walls, soldier beam and lagging walls, slurry walls, tangent and secant pile walls, soil mixed walls, and jet-grouted walls; and In-situ reinforced walls including soil nail walls and micropile walls.

This document is a synthesis of practical information on selection, construction, design, cost, and construction inspection for these wall systems. In addition, information on subsurface investigation, soil and rock design parameter selection, and lateral earth pressure evaluation applied specifically to wall systems, and LRFD methods for retaining walls is provided. The design procedures presented in this manual are consistent with the AASHTO LRFD Bridge Design Specifications (AASHTO, 2007). Significant portions of this manual are based on the following FHWA references: • • • • • • •

Mechanically Stabilized Earth Walls and Reinforced Soil Slopes – Design and Construction Guidelines, by V. Elias, B.R. Christopher and R.R. Berg, FHWA NHI-00-043. Design and Construction Monitoring of Soil Nail Walls, by R. J. Byrne, D. Cotton, J.Porterfield, C. Wolschlag and G. Ueblacker, FHWA-SA-96-069. Earth Retaining Systems, Geotechnical Engineering Circular No. 2, by P.J. Sabatini, V. Elias, G.R. Schmertmann, and R. Bonaparte, FHWA- SA-96-038. Ground Anchors and Anchored Systems, Geotechnical Engineering Circular No. 4, by P.J. Sabatini, D. Pass, and R.C. Bachus, FHWA- SA-99-015. Evaluation of Soil and Rock Properties, Geotechnical Engineering Circular No. 5, by P.J. Sabatini, R.C. Bachus, P.W. Mayne, J.A. Schneider, and T.E. Zettler, FHWA- IF-02-034. Soil Nail Walls, Geotechnical Engineering Circular No. 7, by C.A. Lazarte, V. Elias, R.D. Espinoza, and P.J. Sabatini, FHWA-IF-03-017. Subsurface Investigations, by. P.W. Mayne, B.R. Christopher, and J. DeJong, FHA NHI-01-031.

The authors recognize the efforts of Barry Siel, Daniel Alzamora, and Jerry A. DiMaggio who served as FHWA Technical Consultants for this work.

The authors further acknowledge the efforts of the following individuals who provided valuable technical input during the development of this manual. • • • • • • • • • • • • • • • • • • • • • •

Craig Actis – FHWA Resource Center Tom Armour – DBM Contractors Donald Bruce – Geosystems, L.P. Barry Christopher – Christopher Consultants James Collin – The Collin Group Dave Davis – Northstar Vinyl Inc. Leo Fontaine – Connecticut DOT Jeff Horsfall – Wisconsin DOT Scott Ludlow – Earth Exploration Inc. Steve Maxwell – Wisconsin DOT Peter Osborn – FHWA Bruce Pfister – Wisconsin DOT Ray Poletto – Mueser Rutledge Consulting Engineers Dan Priest – Contech Construction Products Tom Richards – Nicholson Construction Company Ray Sandiford - Port Authority of New York and New Jersey Brian Sweeney - Haley & Aldrich, Inc. Trevor Wang – Colorado DOT Rob Wendt – North American Steel Sheet Piling Association Joe Wiedemann – Berkel and Company Contractors John Wolosick – Hayward Baker, Inc. David Yang – Ratio, Inc.

TABLE OF CONTENTS CHAPTER 1 1.1 1.2 1.3

1.4 1.5 CHAPTER 2 2.1 2.2

2.3

2.4

2.5

INTRODUCTION....................................................................................................................... 1-1 PURPOSE AND SCOPE .............................................................................................................. 1-1 HISTORY OF EARTH RETAINING SYSTEMS ....................................................................... 1-2 CLASSIFICATION OF EARTH RETAINING STRUCTURES ................................................. 1-4 1.3.1 Classification by Load Support Mechanism ................................................................... 1-5 1.3.2 Classification by Construction Concept.......................................................................... 1-5 1.3.3 Classification by System Rigidity................................................................................... 1-7 1.3.4 Temporary and Permanent Wall Applications................................................................ 1-7 ORGANIZATION OF THE MANUAL ....................................................................................... 1-8 PRIMARY REFERENCES FOR THIS MANUAL ..................................................................... 1-9 EVALUATION OF GEOTECHNICAL PARAMETERS FOR EARTH RETAINING SYSTEMS.................................................................................................................................... 2-1 INTRODUCTION ........................................................................................................................ 2-1 PLANNING A SUBSURFACE INVESTIGATION AND LABORATORY TESTING PROGRAM .................................................................................................................... 2-2 2.2.1 General .......................................................................................................................... 2-2 2.2.2 Identify Data Needs and Review Available Information ................................................ 2-2 2.2.3 Developing a Site Investigation Program ....................................................................... 2-4 2.2.4 Sampling Methods .......................................................................................................... 2-9 2.2.5 Developing a Laboratory Testing Program....................................................................2-11 FUNDAMENTAL CONCEPTS OF SHEAR STRENGTH ........................................................2-13 2.3.1 General .........................................................................................................................2-13 2.3.2 Drained Versus Undrained Loading ..............................................................................2-13 2.3.3 Drained Stress-Strain-Strength Behavior.......................................................................2-14 2.3.4 Undrained Stress-Strain-Strength Behavior...................................................................2-17 2.3.5 Effective Stress Parameters ...........................................................................................2-18 2.3.6 Total Stress Parameters..................................................................................................2-18 LABORATORY INDEX TESTS ................................................................................................2-19 2.4.1 General .........................................................................................................................2-19 2.4.2 Moisture Content ...........................................................................................................2-19 2.4.3 Unit Weight ...................................................................................................................2-20 2.4.4 Atterberg Limits.............................................................................................................2-20 2.4.5 Particle Size Distribution ...............................................................................................2-21 2.4.6 Laboratory Classification...............................................................................................2-22 2.4.7 Specific Gravity .............................................................................................................2-22 2.4.8 Organic Content.............................................................................................................2-22 PARAMETERS FOR BACKFILL SOILS ..................................................................................2-23 2.5.1 General .........................................................................................................................2-23 2.5.2 Classification of Backfill Soils ......................................................................................2-23 2.5.3 Unit Weight ...................................................................................................................2-24 2.5.4 Backfill Shear Strength Parameters ...............................................................................2-25 2.5.5 Electrochemical Parameters of Backfill Soils................................................................2-26 2.5.6 Placement and Compaction of Backfill Soils ................................................................2-26

FHWA NHI-07-071 Earth Retaining Structures

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2.6

2.7

CHAPTER 3 3.1 3.2

3.3

3.4 3.5

3.6

2.5.7 Backfill Soil Permeability..............................................................................................2-30 2.5.8 Filtration Requirements .................................................................................................2-31 2.5.9 Other Backfill Materials ................................................................................................2-34 PARAMETERS FOR IN-SITU SOILS .......................................................................................2-37 2.6.1 General .........................................................................................................................2-37 2.6.2 Stratigraphy ...................................................................................................................2-37 2.6.3 Unit Weight ...................................................................................................................2-38 2.6.4 Shear Strength Parameters .............................................................................................2-39 2.6.5 Consolidation Parameters ..............................................................................................2-44 2.6.6 Electrochemical Parameters...........................................................................................2-53 2.6.7 Permeability...................................................................................................................2-53 ROCK PARAMETERS ...............................................................................................................2-56 2.7.1 General .........................................................................................................................2-56 2.7.2 Rock Mass Classification...............................................................................................2-56 2.7.3 Compressive Strength of Intact Rock ............................................................................2-69 2.7.4 Smooth Discontinuity ....................................................................................................2-71 2.7.5 Rough Discontinuity......................................................................................................2-72 2.7.6 Infillings.........................................................................................................................2-72 2.7.7 Shear Strength of Fractured Rock Masses .....................................................................2-73 2.7.8 Permeability...................................................................................................................2-77 LATERAL EARTH AND WATER PRESSURES................................................................... 3-1 INTRODUCTION ........................................................................................................................ 3-1 EARTH PRESSURE THEORY ................................................................................................... 3-2 3.2.1 Active and Passive Earth Pressures ................................................................................ 3-2 3.2.2 At-Rest Earth Pressure.................................................................................................... 3-5 3.2.3 Rankine Earth Pressure Theory ...................................................................................... 3-6 3.2.4 Coulomb Earth Pressure Theory..................................................................................... 3-8 3.2.5 Effect of Wall Friction on Earth Pressures ..................................................................... 3-8 3.2.6 Equivalent Fluid Pressure ..............................................................................................3-11 TEMPORARY CONDITIONS....................................................................................................3-13 3.3.1 General .........................................................................................................................3-13 3.3.2 Softening........................................................................................................................3-13 3.3.3 Tension Cracks ..............................................................................................................3-14 3.3.4 Soil Fabric......................................................................................................................3-15 PERMANENT CONDITIONS....................................................................................................3-16 INFLUENCE OF MOVEMENT ON EARTH PRESSURES......................................................3-17 3.5.1 General .........................................................................................................................3-17 3.5.2 Magnitude of Movement Needed to Mobilize Limit Pressures .....................................3-17 3.5.3 Flexibility of the Wall....................................................................................................3-18 3.5.4 Influence of Movement on Soil Properties ....................................................................3-18 DESIGN EARTH AND WATER PRESSURES .........................................................................3-21 3.6.1 Wall Friction and Adhesion...........................................................................................3-21 3.6.2 Earth Pressure Coefficients............................................................................................3-22 3.6.3 Theoretical Earth Pressures in Stratified Soils...............................................................3-27 3.6.4 Semi Empirical Earth Pressure Diagrams......................................................................3-28


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3.7

3.8 3.9 3.10

3.11 3.12 3.13

3.6.5 Design Water Pressures .................................................................................................3-32 EARTH PRESSURE FROM SURCHARGE LOADS ................................................................3-36 3.7.1 General .........................................................................................................................3-36 3.7.2 Uniform Surcharge Loads..............................................................................................3-36 EARTH PRESSURES DUE TO COMPACTION.......................................................................3-37 EARTH PRESSURES RESULTING FROM SILO EFFECT .....................................................3-40 EARTH PRESSURES RESULTING FROM SEISMIC FORCES .............................................3-41 3.10.1 General .........................................................................................................................3-41 3.10.2 Seismic Earth Pressures .................................................................................................3-41 3.10.3 Displacement Approach.................................................................................................3-46 LATERAL EARTH PRESSURES IN COHESIVE BACKFILLS..............................................3-47 LOADS FROM JOINTED ROCK MASSES ..............................................................................3-49 EXAMPLE PROBLEMS.............................................................................................................3-49 3.13.1 Example Problem 1........................................................................................................3-50 3.13.2 Example Problem 2........................................................................................................3-51 3.13.3 Example Problem 3........................................................................................................3-52

CHAPTER 4 4.1 4.2 4.3 4.4 4.5

LOAD RESISTANCE FACTOR DESIGN FOR EARTH RETAINING SYSTEMS........... 4-1 INTRODUCTION ........................................................................................................................ 4-1 CONCEPT OF LIMIT STATES................................................................................................... 4-2 COMMON LIMIT STATES IN ERS DESIGN............................................................................ 4-3 LOAD COMBINATIONS IN LIMIT STATES ........................................................................... 4-3 EVALUATION OF RESISTANCE FACTORS........................................................................... 4-4

CHAPTER 5 5.1 5.2

CAST-IN-PLACE (CIP) GRAVITY AND SEMI-GRAVITY WALLS ................................. 5-1 INTRODUCTION ........................................................................................................................ 5-1 TYPES OF SEMI-GRAVITY CIP WALLS................................................................................. 5-3 5.2.1 Cantilever........................................................................................................................ 5-3 5.2.2 Counterfort Walls ........................................................................................................... 5-5 5.2.3 Buttress Walls................................................................................................................. 5-6 5.2.4 Other CIP Semi-Gravity Walls ....................................................................................... 5-6 WALL CONSTRUCTION ........................................................................................................... 5-8 COST ........................................................................................................................................ 5-9 WALL DESIGN ..........................................................................................................................5-10 5.5.1 Steps 1, 2, and 3.............................................................................................................5-10 5.5.2 Step 4 – Select Base Dimension ....................................................................................5-12 5.5.3 Step 5 – Select Lateral Earth Pressure Distribution .......................................................5-12 5.5.4 Step 6 – Evaluate Factored Loads..................................................................................5-13 5.5.5 Step 7 - Evaluate Bearing Resistance ............................................................................5-17 5.5.6 Step 8 – Check Eccentricity...........................................................................................5-19 5.5.8 Step 10 – Evaluate Overall Stability..............................................................................5-21 5.5.9 Step 11 – Estimate Maximum Lateral Wall Movement, Tilt, and Wall Settlement at the Service I Limit State........................................................................................5-21 5.5.10 Step 12 – Design Wall Drainage Systems .....................................................................5-23 CONSTRUCTION INSPECTION...............................................................................................5-29

5.3 5.4 5.5

5.6


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CHAPTER 6 6.1 6.2

6.3

6.4

6.5

6.6

6.7 6.8 6.9

CHAPTER 7 7.1 7.2 7.3 7.4

MODULAR GRAVITY WALLS .............................................................................................. 6-1 INTRODUCTION ........................................................................................................................ 6-1 CRIB WALLS .............................................................................................................................. 6-4 6.2.1 General .......................................................................................................................... 6-4 6.2.2 Wall Construction........................................................................................................... 6-4 6.2.3 Materials ......................................................................................................................... 6-6 6.2.4 Cost .......................................................................................................................... 6-7 CONCRETE MODULE WALLS................................................................................................. 6-8 6.3.1 General .......................................................................................................................... 6-8 6.3.2 Wall Construction........................................................................................................... 6-8 6.3.3 Materials ........................................................................................................................6-11 6.3.4 Cost .........................................................................................................................6-11 GABION WALLS .......................................................................................................................6-12 6.4.1 General .........................................................................................................................6-12 6.4.2 Wall Construction..........................................................................................................6-13 6.4.3 Materials ........................................................................................................................6-14 6.4.4 Cost .........................................................................................................................6-17 BIN WALLS................................................................................................................................6-17 6.5.1 General .........................................................................................................................6-17 6.5.2 Wall Construction..........................................................................................................6-18 6.5.3 Materials ........................................................................................................................6-23 6.5.4 Cost .........................................................................................................................6-23 MODULAR GRAVITY WALL DESIGN ..................................................................................6-23 6.6.1 General .........................................................................................................................6-23 6.6.2 Lateral Earth Pressures ..................................................................................................6-25 6.6.3 Sliding .........................................................................................................................6-25 6.6.4 Limiting Eccentricity .....................................................................................................6-27 6.6.5 Bearing Resistance.........................................................................................................6-28 6.6.6 Structural Capacity of Wall Modules ............................................................................6-29 6.6.7 Lateral and Vertical Displacements at the Service Limit State......................................6-29 6.6.8 Overall Stability at the Service Limit State ...................................................................6-29 6.6.9 Gabion Walls .................................................................................................................6-30 EXAMPLE PROBLEM...............................................................................................................6-30 CONSTRUCTION INSPECTION...............................................................................................6-34 T-WALLS ....................................................................................................................................6-36 6.9.1 Overview .......................................................................................................................6-36 6.9.2 T-Wall Construction ......................................................................................................6-38 6.9.3 T-Wall Cost ...................................................................................................................6-38 6.9.4 T-Wall Design ...............................................................................................................6-38 MECHANICALLY STABILIZED EARTH WALLS ............................................................. 7-1 INTRODUCTION ........................................................................................................................ 7-1 PRINCIPAL COMPONENTS OF MSE WALLS ........................................................................ 7-2 APPLICATIONS OF MSE WALLS ............................................................................................ 7-5 MATERIALS FOR MSE WALLS ............................................................................................... 7-9 7.4.1 Backfill Material............................................................................................................. 7-9


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7.5

7.6 7.7 7.8

7.9

CHAPTER 8 8.1 8.2

8.3

8.4

8.5

8.6

7.4.2 Reinforcing Elements ....................................................................................................7-12 7.4.3 Wall Facing Elements....................................................................................................7-15 MSE WALL CONSTRUCTION .................................................................................................7-19 7.5.1 Construction of MSE Walls with Precast Facings .........................................................7-21 7.5.2 Construction of MSE Walls with Flexible Facings .......................................................7-22 7.5.3 Backfill Placement and Compaction..............................................................................7-25 MSE WALL COST......................................................................................................................7-27 CONSTRUCTION INSPECTION OF MSE WALLS.................................................................7-29 MSE WALL DESIGN .................................................................................................................7-32 7.8.1 Initial Design Steps (Steps 1 through 3) ........................................................................7-35 7.8.2 Strength Limit States – External Stability (Steps 4 through 8) ......................................7-36 7.8.3 Strength Limit States – Internal Stability (Steps 9 through 13) .....................................7-48 7.8.4 Service Limit States (Steps 14 and 15) ..........................................................................7-69 7.8.5 Design of Wall Drainage Systems (Step 16) .................................................................7-72 7.8.6 Additional Design Considerations .................................................................................7-74 MSE WALL EXAMPLE PROBLEM .........................................................................................7-76 7.9.1 Problem Statement.........................................................................................................7-76 7.9.2 Solution of the Problem .................................................................................................7-77 NONGRAVITY CANTILEVERED AND ANCHORED WALLS......................................... 8-1 INTRODUCTION ........................................................................................................................ 8-1 SHEET PILE WALLS .................................................................................................................. 8-4 8.2.1 General .......................................................................................................................... 8-4 8.2.2 Wall Construction........................................................................................................... 8-7 8.2.3 Cost .........................................................................................................................8-12 8.2.4 Construction Inspection .................................................................................................8-12 SOLDIER PILE AND LAGGING WALLS ................................................................................8-13 8.3.1 General .........................................................................................................................8-13 8.3.2 Wall Construction..........................................................................................................8-17 8.3.3 Cost .........................................................................................................................8-19 8.3.4 Construction Inspection .................................................................................................8-19 SLURRY WALLS .......................................................................................................................8-20 8.4.1 General .........................................................................................................................8-20 8.4.2 Slurry Wall Types..........................................................................................................8-23 8.4.3 Wall Construction..........................................................................................................8-26 8.4.4 Cost .........................................................................................................................8-34 8.4.5 Construction Inspection .................................................................................................8-34 TANGENT/SECANT PILE WALLS ..........................................................................................8-37 8.5.1 General .........................................................................................................................8-37 8.5.2 Wall Construction..........................................................................................................8-39 8.5.3 Cost .........................................................................................................................8-40 8.5.4 Construction Inspection .................................................................................................8-40 JET GROUTED WALLS ............................................................................................................8-42 8.6.1 General .........................................................................................................................8-42 8.6.2 Wall Construction..........................................................................................................8-45 8.6.3 Cost .........................................................................................................................8-54


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8.7

8.8

8.9

8.10

8.11

8.12

8.13 CHAPTER 9 9.1 9.2

8.6.4 Construction Inspection .................................................................................................8-54 DEEP MIXING METHOD (DMM) WALLS..............................................................................8-57 8.7.1 General .........................................................................................................................8-57 8.7.2 Wall Construction..........................................................................................................8-60 8.7.3 Cost .........................................................................................................................8-64 8.7.4 Construction Inspection .................................................................................................8-64 GROUND ANCHORS ................................................................................................................8-67 8.8.1 General .........................................................................................................................8-67 8.8.2 Wall Construction..........................................................................................................8-68 8.8.3 Load Testing of Ground Anchors ..................................................................................8-76 8.8.4 Cost .........................................................................................................................8-84 8.8.5 Construction Inspection .................................................................................................8-84 DESIGN OF FLEXIBLE NONGRAVITY CANTILEVERED WALLS....................................8-86 8.9.1 Sheet Pile and Soldier Pile and Lagging Walls..............................................................8-86 8.9.2 Design Steps for Flexible Nongravity Cantilevered Walls ............................................8-87 ANCHORED BULK HEAD (DEADMAN ANCHOR) WALLS ...............................................8-95 8.10.1 Overview .......................................................................................................................8-95 8.10.2 Free Earth Support.........................................................................................................8-96 8.10.3 Fixed Earth Support.......................................................................................................8-97 8.10.4 Deadman Anchor Systems...........................................................................................8-102 ANCHORED WALL DESIGN .................................................................................................8-102 8.11.1 Step 3: Select Corrosion Protection of Ground Anchors .............................................8-106 8.11.2 Step 4: Select Earth Pressure Diagram ........................................................................8-114 8.11.3 Step 5: Evaluate Factored Loads..................................................................................8-118 8.11.4 Step 6: Evaluate Individual Anchor Loads and Subgrade Reaction Force ..................8-118 8.11.5 Step 7: Evaluate Anchor Inclination ............................................................................8-122 8.11.6 Step 8: Select Tendon Type and Check Tensile Resistance.........................................8-122 8.11.7 Step 9: Evaluate Anchor Bond Length ........................................................................8-124 8.11.8 Step 10: Evaluate Factored Bending Moment and Flexural Resistance of Wall .........8-127 8.11.9 Step 11: Evaluate Bearing Resistance of Vertical Wall Element.................................8-131 8.11.10 Step 12: Evaluate Overall Stability of Anchored Wall at Service Limit State.............8-132 8.11.11 Step 13: Estimate Maximum Lateral Wall Movements and Settlements at the Service Limit State .....................................................................................................8-135 DESIGN OF STIFF CANTILEVERED AND ANCHORED WALLS.....................................8-137 8.12.1 General .......................................................................................................................8-137 8.12.2 Soil-Structure Interaction analysis Methods ................................................................8-140 8.12.3 Slurry Walls.................................................................................................................8-151 8.12.4 Tangent and Secant Pile Walls ....................................................................................8-152 8.12.5 Jet Grouted Walls ........................................................................................................8-153 8.12.6 DMM Walls.................................................................................................................8-156 DRAINAGE SYSTEMS FOR CUT WALLS ...........................................................................8-160 IN-SITU REINFORCED WALLS ............................................................................................ 9-1 INTRODUCTION ........................................................................................................................ 9-1 SOIL NAIL WALLS .................................................................................................................... 9-1 9.2.1 General .......................................................................................................................... 9-1


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9.3

CHAPTER 10 10.1 10.2 10.3 10.4 10.5

10.6 10.7 10.8

CHAPTER 11 11.1 11.2

11.3

11.4 11.5

9.2.2 Feasibility of Soil Nails .................................................................................................. 9-3 9.2.3 Construction Materials and Methods .............................................................................9-13 9.2.4 Soil Nail Corrosion Protection.......................................................................................9-30 9.2.5 Cost .........................................................................................................................9-34 9.2.6 Design Concepts for Soil Nails......................................................................................9-35 9.2.7 Design of Soil Nail Walls ..............................................................................................9-72 9.2.8 Soil Nail Load Testing.................................................................................................9-101 9.2.9 Construction Inspection of Soil Nail Walls .................................................................9-106 MICROPILE WALLS ...............................................................................................................9-108 9.3.1 Introduction .................................................................................................................9-108 9.3.2 Construction Materials and Methods ...........................................................................9-109 9.3.3 Micropile Wall Design ................................................................................................9-112 9.3.4 Load Testing ................................................................................................................9-114 9.3.5 Construction Inspection ...............................................................................................9-115 WALL SELECTION .................................................................................................................10-1 INTRODUCTION .......................................................................................................................10-1 STEP 1: IDENTIFY NEED FOR AN EARTH RETAINING SYSTEM ....................................10-1 STEP 2: IDENTIFY SITE CONSTRAINTS AND PROJECT REQUIREMENTS ....................10-2 STEP 3: EVALUATE PROJECT REQUIREMENTS AGAINST FACTORS AFFECTING WALL SELECTION .....................................................................................................10-3 STEP 4: EVALUATE WALL ALTERNATIVES AGAINST WALL SELECTION FACTORS10-5 10.5.1 General .........................................................................................................................10-5 10.5.2 Selection Issues for Cut Wall Systems ..........................................................................10-5 10.5.3 Selection Issues for Fill Wall Systems...........................................................................10-6 10.5.4 Wall System Alternatives Evaluation ............................................................................10-6 STEP 5: SELECT AN ACCEPTABLE WALL TYPE..............................................................10-14 WALL SELECTION EXAMPLE..............................................................................................10-14 HYBRID WALL SYSTEMS.....................................................................................................10-19 10.8.1 General .......................................................................................................................10-19 10.8.2 Compatibility of Deformations....................................................................................10-19 10.8.3 Overall Stability of Hybrid Systems ............................................................................10-21 10.8.4 Example Hybrid Systems.............................................................................................10-23 CONTRACTING APPROACHES...........................................................................................11-1 INTRODUCTION .......................................................................................................................11-1 METHOD CONTRACTING APPROACH.................................................................................11-2 11.2.1 Introduction ...................................................................................................................11-2 11.2.2 Contract Documents for Method Approach...................................................................11-3 PERFORMANCE CONTRACTING APPROACH ....................................................................11-8 11.3.1 Introduction ...................................................................................................................11-8 11.3.2 Implementing Performance Contracting Approach .......................................................11-8 11.3.3 Contract Documents for Performance Approach.........................................................11-10 11.3.4 Review and Approval ..................................................................................................11-12 CONTRACTOR DESIGN/BUILD APPROACH......................................................................11-12 RECOMMENDATIONS ...........................................................................................................11-13


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11.6 CHAPTER 12

SPECIFICATION RESOURCES ..............................................................................................11-14 REFERENCES...........................................................................................................................12-1

APPENIDX – STUDENT EXERCISES AND SOLUTIONS ..............................................................................A-1 Active Earth Pressure Calculation Lateral Load Distribution Due to Strip Load Sliding Resistance MSE Wall Design External Stability MSE Wall Internal Stability Corrosion Calculation Apparent Earth Pressure Diagrams Single-Tier Anchored Soldier Beam and Lagging Wall Design of a Two-Tier Anchored Soldier Pile and Lagging Wall Lateral Wall Movement Soil Nail Wall Design Fill Wall Selection Cut Wall Selection


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LIST OF TABLES Table 2-1 Table 2-2 Table 2-3 Table 2-4 Table 2-5 Table 2-6 Table 2-7 Table 2-8 Table 2-9 Table 2-10 Table 2-11 Table 2-12 Table 2-13 Table 2-14 Table 2-15 Table 2-16 Table 2-17 Table 2-18 Table 2-19

Summary of Information Needs and Testing Considerations for Fill and Cut Wall Applications ....... 2-3 Sources of Historical Site Data ............................................................................................................ 2-5 In-Situ Testing Methods Used in Soil .................................................................................................. 2-8 Sampling Guidelines (after WFLHD, 2004) .......................................................................................2-12 Typical Properties of Compacted Soils (after NAVFAC, 1986).........................................................2-27 Unconfined Compressive Strength of Particles for Rockfill Grades in Figure 2-4 .............................2-28 Range of Dry Densities for Lightweight Fills (after Elias et al., 2006)...............................................2-35 Correlation between N and g of Granular Soils (after Bowles, 1988).................................................2-38 Correlation between N and g of Cohesive Soils (after Bowles, 1988) ................................................2-38 Relationship between SPT N Value and Internal Friction angle of Granular Soils (after AASHTO LRFD, 2007) .......................................................................................................................................2-45 Summary of Correlations for Cc (after Holtz and Kovacs, 1981) .......................................................2-48 Casagrande Method to Evaluate σp’ ....................................................................................................2-50 Description of Geological Mapping Terms.........................................................................................2-59 Rock Material Strengths......................................................................................................................2-60 Weathering Grades..............................................................................................................................2-60 Rock Quality Description Based on Rqd ............................................................................................2-61 Csir Classification of Jointed Rock Mass............................................................................................2-64 Typical Ranges of Friction angles for a Variety of Rock Types .........................................................2-71 Permeability of Typical Rocks and Soils (after Wyllie and Mah, 1998).............................................2-79

Table 3-1 Table 3-2 Table 3-3 Table 3-4

Typical Values for Equivalent Fluid Unit Weight of Soils (after AASHTO, 2007) ...........................3-12 Wall Friction and Adhesion for Dissimilar Materials (after NAVFAC, 1986) ...................................3-22 Lateral Earth and Hydrostatic Pressures at Various Depths for Example 1 ........................................3-50 Computation of Lateral Earth Pressures Due to Line Load.................................................................3-54

Table 4-1 Table 4-2 Table 4-3

Load Combinations and Load Factors (AASHTO, 2007).................................................................... 4-5 Load Factors for Permanent Loads, γP (modified after AASHTO, 2007) ............................................ 4-6 Values of Resistance Factors Corresponding to Different Values of Factor of Safety and Dead to Live Load Ratios for γDC = 1.25 and γLL = 1.75 ................................................................................... 4-8


Design Steps for Gravity and Semi-Gravity Walls .............................................................................5-11 Suggested Gradation for Backfill for Cantilever Semi-Gravity and Gravity Retaining Walls............5-12 Inspector Responsibilities for a Typical CIP Gravity and Semi Gravity Wall Project........................5-30

Table 6-1 Table 6-2

Typical Height-Thickness Relationship for Bin Walls (after Contech) ..............................................6-18 Inspector Responsibilities for a Typical Modular Gravity Wall Project .............................................6-34

Table 7-1

Summary of Reinforcement and Face Panel Details for Selected MSE Wall Systems (after Elias et al., 2004)........................................................................................................................................... 7-3 Recommended Limits of Electrochemical Properties for Select Backfills When Using Steel Reinforcement .....................................................................................................................................7-10 Recommended Limits of Electrochemical Properties for Select Backfills When Using Geosynthetic Reinforcements .............................................................................................................7-11 MSE Wall Field Inspection Checklist .................................................................................................7-30



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Table 7-5 Table 7-6 Table 7-7 Table 7-8 Table 7-9 Table 7-10 Table 7-11 Table 7-12 Table 7-13 Table 7-14 Table 7-15 Table 7-16 Table 7-17 Table 7-18 Table 7-19 Table 7-20 Table 8-1 Table 8-2 Table 8-3 Table 8-4 Table 8-5 Table 8-6 Table 8-7 Table 8-8 Table 8-9 Table 8-10 Table 8-11 Table 8-12 Table 8-13 Table 8-14 Table 8-15 Table 8-16 Table 8-17 Table 8-18 Table 8-19 Table 8-20 Table 8-21 Table 8-22 Table 8-23

Design Steps for MSE Walls...............................................................................................................7-34 Minimum Embedment Requirements for MSE Walls (after AASHTO, 2007)...................................7-36 Typical Values for a (after AASHTO, 2007) ......................................................................................7-58 Resistance Factors for Tensile Resistance (after AASHTO, 2007).....................................................7-62 Installation Damage Reduction Factors (after Elias, 2000).................................................................7-67 Creep Reduction Factors (RFCR) .......................................................................................................7-67 Aging Reduction Factors (RFD) for PET............................................................................................7-68 Relationship between Joint Width and Limiting Differential Settlements for MSE Precast Panels (after Elias et al., 2001) .......................................................................................................................7-71 Equivalent Height of Soil for Vehicular Loading (after AASHTO, 2007) .........................................7-79 Summary of Unfactored Vertical Loads and Moment Arms for Design Example..............................7-80 Summary of Unfactored Horizontal Loads and Moment Arms for Design Example .........................7-80 Load Factors and Load Combinations.................................................................................................7-80 Factored Vertical Loads and Moments ...............................................................................................7-81 Factored Horizontal Loads and Moments ...........................................................................................7-81 Summary for Eccentricity Check ........................................................................................................7-81 Summary for Checking Bearing Resistance........................................................................................7-84 Inspector Responsibilities for a Typical Sheet Pile Wall Project ........................................................8-12 Inspector Responsibilities for a Typical Soldier Pile and Lagging Wall Project ................................8-19 Inspector Responsibilities for a Typical Slurry Wall Project ..............................................................8-35 Inspector Responsibilities for a Typical Tangent/Secant Pile Wall Project ........................................8-41 Inspector Responsibilities for a Typical Jet-Grouted Wall Project .....................................................8-56 Inspector Responsibilities for Typical DMM Walls............................................................................8-65 Example of QA/QC Testing Program for DMM Walls ......................................................................8-66 Load Schedule and Observation Periods for Extended Creep Test for Permanent Anchor ................8-79 Inspector Responsibilities for Ground Anchors and Anchored Walls.................................................8-85 Design Steps for Flexible Nongravity Cantilevered Walls..................................................................8-88 Recommended Thickness of Wood Lagging (after Goldberg et al., 1976).........................................8-94 Bending Moments for Facing Design (AASHTO, 2007)....................................................................8-95 Design Steps for Anchored Walls .....................................................................................................8-105 Criteria for Electrochemical Properties of Soils for Ground Anchor Applications (after Cheney, 1988)..........................................................................................................................8-113 Properties of Prestressing Steel Bars (ASTM A722) ........................................................................8-123 Properties of 0.6 in. Diameter Prestressing Steel Strands (ASTM A416, Grade 270) ......................8-123 Guidance Relationship between Tendon Size and Trumpet Opening Size (after Sabatini et al., 1999) ................................................................................................................8-124 Presumptive (Nominal) Bond Stress for Ground/Grout Interface Along Anchor Bond Zone (after PTI, 1996)................................................................................................................................8-128 Cutoff Functions and Watertightness of Excavation Walls (after ASCE, 1997)...............................8-148 Summary of System Variables and their Impact on Basic Design Elements ....................................8-154 Typical Range of Jet Grouting Parameters and Jet-Grouted Soil Properties (adapted from Kauschinger and Welsh, 1989) ..................................................................................8-155 Range of Typical Soilcrete Strengths (Three Fluid System) (after Elias et al., 2006) ......................8-156 Typical Improved Engineering Characteristics of Soils Treated with DMM (Wet Mix) (after Elias et al., 2006) .....................................................................................................................8-157


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Table 9-1 Table 9-2 Table 9-3 Table 9-4 Table 9-5 Table 9-6 Table 9-7 Table 9-8 Table 9-9 Table 9-10 Table 9-11 Table 9-12 Table 9-13 Table 9-14 Table 9-15

Criteria for Assessing Ground Corrosion Potential.............................................................................9-31 Recommendations for Minimum Levels of Corrosion Protection for Soil Nails ................................9-33 Estimated Ultimate Bond Strength of Soil Nails in Soil and Rock (after Elias and Juran, 1991) .......9-47 Factors CF (after Lazarte, 2003) ..........................................................................................................9-56 Values of (δh/H)i and C As Functions of Soil Conditions ...................................................................9-68 Design Steps for Soil Nail Walls.........................................................................................................9-72 Minimum Recommended Factors of Safety for the Design of Soil Nail Walls Using the ASD Method (after Lazarte, 2003) ..............................................................................................................9-73 Variable Parameters Used in Design Charts (Lazarte et al., 2003) .....................................................9-87 Threaded Bar Properties......................................................................................................................9-90 Welded Wire Mesh Dimensions (after WRI, 2001) ............................................................................9-94 Headed-Stud Dimensions (Metric and English Units) (after Byrne et al., 1996) ................................9-96 Facing Resistance for Various Failure Modes (after Lazarte et al., 2003) ..........................................9-98 Soil Nail Load Test Types.................................................................................................................9-102 Inspector Responsibilities for a Typical Soil Nail Wall ....................................................................9-106 Inspector Responsibilities for a Typical Micropile Wall Project ......................................................9-115

Table 10-1 Table 10-2 Table 10-3 Table 10-4 Table 10-5 Table 10-6 Table 10-7

Wall Selection Factors ........................................................................................................................10-4 Summary of Evaluation Factors for Cut Walls (after Sabatini et al., 1997)........................................10-7 Summary of Evaluation Factors for Fill Walls (after Sabatini et al., 1997) ........................................10-8 Summary of Cut Wall Costs..............................................................................................................10-15 Summary of Fill Wall Costs..............................................................................................................10-15 Initial Rating of Each Wall Selection Factor for Each Wall Alternative...........................................10-18 Wall Selection Matrix and Total Score of Each Wall Alternative ....................................................10-20

Table 11-1

Identified Sources for Wall Specifications........................................................................................11-15


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LIST OF FIGURES Figure 1-1 Figure 1-2 Figure 1-3

Schematic of a Retaining Wall and Common Terminology .............................................................. 1-1 Variety of Retaining Walls (after O’Rourke and Jones, 1990) .......................................................... 1-3 Classification of Earth Retaining Systems (after O’Rourke and Jones, 1990)................................... 1-6

Figure 2-1 Figure 2-2 Figure 2-3 Figure 2-4 Figure 2-5

Boring Layout for Anchored Wall..................................................................................................... 2-6 Drained Stress-Strain Behavior.........................................................................................................2-16 Mohr-Coulomb Failure Criteria........................................................................................................2-16 Typical Ranges of Friction angle for Rockfills, Gravels, and Sands (after Terzaghi et al., 1996)....2-28 Correlation between Drained Friction angle and the Dry Unit Weight, Relative Density, and Soil Classification (after NAVFAC, 1986) ..............................................................................................2-29 Example Shear Strength Property Evaluation for Clayey Backfill ...................................................2-30 Classification Chart for Swelling Potential (after Seed et al., 1962) ................................................2-36 Test Data from Triaxial Compression Test with Pore Pressure Measurements ................................2-40 Plasticity Based VST Correction Factors..........................................................................................2-41 Relationship between φ’ and PI (after Terzaghi et al., 1996) ...........................................................2-43 Residual Friction angles for Clayey Soils (after Stark and Eid, 1994) .............................................2-43 Friction Angle of Cohesionless Soils (a) from Uncorrected SPT N Values (modified after Peck, Hanson, and Thornburn 1974) and (b) As a Function of Normalized Overburden (modified after Schmertmann, 1975).........................................................................................................................2-45 Correlation of φ’ with Normalized CPT qt Data in Clean Sands ......................................................2-46 Definition of Cc, Cr, Cs, and σp’ (Sabatini et al., 2002) ...................................................................2-47 Illustration of Casagrande Method to Evaluate Preconsolidation Stress...........................................2-50 Evaluation of Cα ...............................................................................................................................2-53 Range of Hydraulic Conductivity Values Based on Soil Type (after Holtz and Kovacs, 1981).......2-55 Range of Hydraulic Conductivity Based on Grain Size (after Geosyntec, 1991) .............................2-56 Illustration of Geological Mapping Terms (after Wyllie, 1999, Foundations on Rock, Figure 4.4b, p. 101, E&Fn Spon) ..........................................................................................................................2-57 List of Parameters and Categories Describing Rock Mass Characteristics (after Wyllie, 1999) ......2-58 Calculation of Core Recovery and ROD...........................................................................................2-62 General Chart for GSI Estimates from Geological Observations (after Marinos et al., 2004)..........2-67 GSI Estimates for Heterogeneous Rock Masses (after Marinos et al., 2004) ...................................2-68 Point Load Strength Test Equipment ................................................................................................2-70 Definition of Joint Roughness Coefficient, JRC (after Barton, 1973) ..............................................2-74 Simplified Division of Filled Discontinuities into Displaced, and Normally Consolidated and Overconsolidated Categories (after Wyllie, 1999)............................................................................2-75 RocLab Window to Select Uniaxial Compressive Strength of Intact Rock......................................2-76 RocLab Window to Select GSI.........................................................................................................2-77 RocLab Window to Select mi Value.................................................................................................2-78 Shear Strength Parameters for Weathered Shale Using RocLab Software .......................................2-79 Influence of Joint Openings and Joint Spacing on the Permeability of Rocks (after Hoek and Bray, 1977) .............................................................................................................2-80

Figure 2-6 Figure 2-7 Figure 2-8 Figure 2-9 Figure 2-10 Figure 2-11 Figure 2-12

Figure 2-13 Figure 2-14 Figure 2-15 Figure 2-16 Figure 2-17 Figure 2-18 Figure 2-19 Figure 2-20 Figure 2-21 Figure 2-22 Figure 2-23 Figure 2-24 Figure 2-25 Figure 2-26 Figure 2-27 Figure 2-28 Figure 2-29 Figure 2-30 Figure 2-31

Figure 3-1 Figure 3-2

Magnitudes and Patterns of Movement to Develop Lateral Earth Pressures (after Sabatini et al., 1997)................................................................................................................. 3-3 Mobilization of Rankine Active and Passive Horizontal Pressures for a Smooth Retaining Wall .... 3-4


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Figure 3-29 Figure 3-30 Figure 3-31 Figure 3-32 Figure 3-33

Limiting Active and Passive Horizontal Pressures ............................................................................ 3-4 (a) Wall Pressures for a Cohesionless Soil; and (b) Wall Pressures for Soil with a Cohesion Intercept (after Padfield and Mair, 1984)........................................................................................... 3-7 Coulomb Coefficients Ka and Kp for Sloping Wall with Wall Friction and Sloping Backfill (after NAVFAC, 1986) ...................................................................................................................... 3-9 Wall Friction on Soil Wedges (after Padfield and Mair, 1984) ........................................................3-10 Comparison of Plane and Curved (Log-Spiral) Failure Surfaces (a) Active Case and (b) Passive Case (after Sokolovski, 1954) ........................................................................................3-12 Earth Pressure Immediately after Loading (after Padfield and Mair, 1984) .....................................3-15 Effect of Wall Movement on Wall Pressures (after Canadian Foundation Engineering Manual, 1992)...................................................................................................................................3-19 Simplified Drained Stress-Displacement Relationship for a Stiff Clay (modified after Padfield and Mair, 1984) ..................................................................................................................3-20 Passive Coefficients for Sloping Wall with Wall Friction and Horizontal Backfill (Caquot and Kerisel, 1948; NAVFAC, 1986) .......................................................................................................3-25 Passive Coefficients for Vertical Wall with Wall Friction and Sloping Backfill (Caquot and Kerisel, 1948; NAVFAC, 1986) .......................................................................................................3-26 Pressure Distribution for Stratified Soils ..........................................................................................3-27 Cross Section of Model Wall (modified after Mueller et al., 1998) .................................................3-28 Lateral Wall Movements and Earth Pressures with Excavation at First Anchor Level (Cantilever Stage) (modified after Mueller et al., 1998)......................................................................................3-29 Lateral Wall Movements and Earth Pressures During Anchor Stressing (modified after Mueller et al., 1998)..........................................................................................................................3-30 Lateral Wall Movements and Earth Pressures with Excavation at Lower Anchor Level (modified after Mueller et al., 1998).................................................................................................3-31 Lateral Wall Movements and Earth Pressures with Excavation at Design Grade (modified after Mueller et al., 1998).................................................................................................3-31 Apparent Earth Pressure Diagram for Anchored Wall in Sand ........................................................3-33 Computation of Lateral Pressures for Static Groundwater Case.......................................................3-33 Flow Net for a Retaining Wall (after Padfield and Mair, 1984) .......................................................3-34 Gross and Net Water Pressures Across a Retaining Wall (modified after Padfield and Mair, 1984) ..................................................................................................................3-35 (a) Retaining Wall with Uniform Surcharge Load and (b) Retaining Wall with Line Loads (see Railway Tracks) and Point Loads (see Catenary Structure) ......................................................3-38 Lateral Pressure Due to Surcharge Loadings (after USS Steel Sheet-Pile Manual, 1975)................3-39 Typical Residual Earth Pressure after Compaction of Backfill Behind an Unyielding Wall (after Clough and Duncan, 1991)......................................................................................................3-40 Estimation of “Silo” Pressures..........................................................................................................3-42 Seismic Forces Behind a Gravity Wall .............................................................................................3-43 Effects of Seismic Coefficients and Friction angle on Seismic Active Pressure Coefficient (after Lam and Martin, 1986)............................................................................................................3-45 Example Problem 1 Geometry and Soil Conditions .........................................................................3-50 (a) Lateral Effective Earth Pressure Diagram and (b) Water Pressure Diagram...............................3-51 Pressure Diagrams for Example Problem 2 ......................................................................................3-52 Geometry of Example Problem 3 .....................................................................................................3-52 Lateral Pressure with Depth Due to Line Load.................................................................................3-54

Figure 4-1

Equations Used to Relate LRFD Resistance Factor to ASD FS (after Samtani, 2007) ..................... 4-6

Figure 3-5 Figure 3-6 Figure 3-7 Figure 3-8 Figure 3-9 Figure 3-10 Figure 3-11 Figure 3-12 Figure 3-13 Figure 3-14 Figure 3-15 Figure 3-16 Figure 3-17 Figure 3-18 Figure 3-19 Figure 3-20 Figure 3-21 Figure 3-22 Figure 3-23 Figure 3-24 Figure 3-25 Figure 3-26 Figure 3-27 Figure 3-28


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Figure 5-1 Figure 5-2 Figure 5-3 Figure 5-4 Figure 5-5 Figure 5-6 Figure 5-7 Figure 5-8 Figure 5-9 Figure 5-10 Figure 5-11 Figure 5-12 Figure 5-13 Figure 5-14 Figure 5-15 Figure 5-16 Figure 5-17 Figure 5-18 Figure 5-19 Figure 5-20 Figure 5-21 Figure 5-22

Cast-in-Place Gravity Wall ................................................................................................................ 5-2 Cast-in-Place (CIP) Concrete Retaining Walls and Terminology (a) Cantilever Wall (Bowles, 1988); (b) Counterfort Wall; and (c) Buttress Wall (Teng, 1962) ..................................................... 5-3 Common Proportions of Cantilever Walls (after Teng, 1962)........................................................... 5-4 CIP Cantilever Retaining Wall .......................................................................................................... 5-4 Common Proportions of Counterfort Walls (after Teng, 1962)......................................................... 5-5 CIP Counterfort Wall Construction in a Cut Application .................................................................. 5-6 Other Types of Cast-in-Place (CIP) Walls: (a) Fill Wall with Limited Row, (b) Cut Wall with Limited Row, and (c) U-Wall for Depressed Roadway ..................................................................... 5-7 CIP U-Wall for Depressed Roadway ................................................................................................. 5-8 Strength Limit States for Rigid Gravity and Semi-Gravity Walls ....................................................5-10 CIP Abutment with Integral Wingwalls............................................................................................5-13 Typical Application of Load Factors for Eccentricity and Sliding ...................................................5-14 Typical Application of Load Factors for Bearing Resistance ...........................................................5-14 Example CIP Wall with Traffic Barrier............................................................................................5-16 Computation of Barrier Load at Footing Level.................................................................................5-16 Loading and Eccentricity for Walls Founded on Soil.......................................................................5-18 Loading and Eccentricity for Walls Founded on Rock .....................................................................5-19 Typical Modes of Global Stability (after Bowles, 1988) ..................................................................5-22 Typical Movement of Pile Supported Cast-in-Place (CIP) Wall with Soft Foundation ...................5-23 Potential Sources of Subsurface Water.............................................................................................5-25 Typical Retaining Wall Drainage Alternatives (after Sabatini et al., 1997) .....................................5-25 Inclined Drain for Reducing Water Pressure Behind Wall (after Cedergren, 1989).........................5-26 Drains Behind Backfill in Cantilever Wall in a Cut Situation ..........................................................5-27

Figure 6-1

Modular Gravity Walls (a) Metal Bin Wall; (b) Precast Concrete Crib Wall; (c) Precast Concrete Module Wall; and (d) Gabion Wall (after AASHTO, 2007) ............................................................. 6-1 Figure 6-2 Examples of Modular Gravity Wall Applications (after Contech, 1997) .......................................... 6-3 Figure 6-3 Crib Walls (a) Uniform Cross-Section; (b) Stepped Cross-Section; and (c) Typical Details of a Reinforced Concrete Crib Wall (after HKGEO, 1993)...................................................................... 6-5 Figure 6-4 Setting Precast Elements for an Open Faced Crib Wall..................................................................... 6-6 Figure 6-5 Concrete Module Wall (a) Typical Section; (b) Typical Module; and (c) Precast Parapet (after Doublewal Corporation) .................................................................................................................... 6-9 Figure 6-6 Construction of Concrete Module Wall (a) Placement of Precast Modules; (b) Placing Fill Within the Modules; and (c) Compacting the Infill Material (after Doublewal Corp.) ................................6-10 Figure 6-7 Gabion Baskets (a) Module Without Diaphragms and (b) Module with Diaphragms ......................6-13 Figure 6-8 Gabion Wall Construction (a) Filling Gabion Baskets with Stone and (b) Closing Gabion Lid for Tying ...........................................................................................................................................6-15 Figure 6-9 Typical Geometry of Type 2 Bin Wall (a) Plan, (b) Elevation and (c) Section a-a (after Contech) ..................................................................................................................................6-19 Figure 6-10 Elements of Bin Walls (a) T-Shaped Vertical Connector for Bin Wall Type 2 and (b) Channel Shaped Vertical Connector for Bin Wall Type 1(after Contech) ..................................6-20 Figure 6-11a Construction of a Bin Wall (a) Setting Preassembled Panels and (b) Filling the Completed Bins (after Contech) ..................................................................................................................................6-21 Figure 6-11b Construction of a Bin Wall (a) Setting Preassembled Panels and (b) Filling the Completed Bins (after Contech) ..................................................................................................................................6-21 Figure 6-12 Construction of Bin Walls at Curves (a) Typical Outside Corner; and (b) Typical Inside Corner (after Contech) ..................................................................................................................................6-22


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Figure 6-13 Figure 6-14 Figure 6-15 Figure 6-16 Figure 6-17 Figure 6-18 Figure 6-19 Figure 7-1 Figure 7-2 Figure 7-3 Figure 7-4 Figure 7-5 Figure 7-6 Figure 7-7 Figure 7-8 Figure 7-9

Figure 7-10 Figure 7-11 Figure 7-12 Figure 7-13 Figure 7-14 Figure 7-15 Figure 7-16 Figure 7-17 Figure 7-18 Figure 7-19 Figure 7-20 Figure 7-21 Figure 7-22 Figure 7-23 Figure 7-24 Figure 7-25 Figure 7-26

Bin Wall with (a) Corrugated Steel Face Panels; and (b) Precast Concrete Face Panels (after Contech) ..................................................................................................................................6-24 Earth Pressure Distribution for Modular Walls with Continuous Pressure Surfaces (Figure 3.11.5.9-1 from AASHTO, 2007) ........................................................................................6-26 Earth Pressure Distribution for Modular Walls with Irregular Pressure Surfaces (Figure 3.11.5.9-2 from AASHTO, 2007) ........................................................................................6-27 Load Distribution for Modular Wall with Footings..........................................................................6-28 Geometry and Parameters of Example Problem ...............................................................................6-31 Vertically Installed T-Wall Units (after The Neel Company, 2004).................................................6-36 Battered T-Wall Units (after The Neel Company, 2004)..................................................................6-37 Principal Components of a Mechanically Stabilized Earth Wall (after Christopher, et al., 1990)..... 7-2 Examples of MSE Wall Applications (a) Retaining Wall; (b) Access Ramp; (c) Waterfront Structure; and (d) Bridge Abutment .................................................................................................. 7-6 Examples of MSE Walls.................................................................................................................... 7-7 Metallic Reinforcements (a) Ribbed Metal Strip and (b) Welded Bar Mat ......................................7-13 Geosynthetic Reinforcements (a) Geogrid and (b) Geotextile Facing ..............................................7-14 Various Types of Wall Facing (after Wu, 1994)...............................................................................7-17 MSE Wall Surface Textures .............................................................................................................7-18 Examples of Commercially Available MBW Units (after Simac Et Al, 1996) ................................7-20 Construction of MSE Wall with Inextensible Strip Reinforcements: (a) Concrete Leveling Pad Construction, (b) Erection of Facing Elements, (c) Placement of Reinforcements (d) Placement of Backfill, (e) Spreading of Backfill, and (f) Backfill Compaction.................................................7-23 Construction Procedures of Geotextile Retained Earth Wall with Wrap-Around Facing.................7-24 Geotextile Retained Earth Wall Details ............................................................................................7-24 Lightweight Compaction Adjacent to Wall Facing ..........................................................................7-28 Potential External Failure Mechanisms for MSE Walls ...................................................................7-33 Pressure Diagram for MSE Walls with Horizontal Backslope and Traffic Surcharge (after AASHTO, 2007) .....................................................................................................................7-37 Typical Application of Live Load Surcharge for MSE Walls (after AASHTO, 2007).....................7-38 Distribution of Stress from Concentrated Vertical Load (after AASHTO, 2007).............................7-39 Distribution of Stress from Concentrated Horizontal Loads for External and Internal Stability Calculations (after AASHTO, 2007) ................................................................................................7-40 Pressure Diagram for MSE Walls with Sloping Backslope (after AASHTO, 2007) ........................7-42 Pressure Diagram for MSE Walls with Broken Backslope (after AASHTO, 2007).........................7-42 Calculation of Eccentricity for Horizontal Backslope with Traffic Surcharge Condition (after Elias and Christopher 1996)....................................................................................................7-44 Calculation of Eccentricity for Sloping Backslope Condition (after Elias and Christopher 1996)...7-45 Mechanisms of Internal Failure in MSE Walls (a) Tension Failure and (b) Pullout Failure (after Christopher et al., 1990)..........................................................................................................7-49 Location of Potential Failure Surface for Internal Stability Design of MSE Walls (a) Inextensible Reinforcements and (b) Extensible Reinforcements (after Elias et al., 2001) ..................................7-50 Variation of the Coefficient of Lateral Stress Ratio (K/Ka) with Depth in a Mechanically Stabilized Earth Wall (after AASHTO, 2007) ..................................................................................7-52 Definitions of B, Sh, and Sv for (a) Metal and (b) Geosynthetic Reinforcements (after AASHTO, 2007) .....................................................................................................................7-55 Mechanisms of Pullout Resistance. (a) by Friction, (b) by Passive Resistance (after Christopher et al., 1990 and Lawson, 1992)............................................................................7-56


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Figure 7-27 Figure 7-28 Figure 7-29 Figure 7-30 Figure 7-31 Figure 7-32 Figure 7-33 Figure 7-34

Typical Values for F* (after AASHTO, 2007) .................................................................................7-61 Cross Section Area for Strips............................................................................................................7-64 Cross Section Area for Bars..............................................................................................................7-64 Drain Immediately Behind Concrete Facing in MSE Wall...............................................................7-73 Drain Behind the Backfill in a Wall, in a Backcut, and Fill Situation ..............................................7-73 Impervious Geomembrane Details ...................................................................................................7-74 (a) Geometry of the Problem and (b) External Forces to Be Considered in analysis........................7-77 Wall Face Panels and Spacing between Reinforcements..................................................................7-78

Figure 8-1

Primary Types of Externally Supported Structural Walls: (a) Soldier Pile and Lagging Wall; (b) Soldier Pile and Cast-in-Place Concrete Lagging Wall; (c) Master Pile Wall; (d) Sheet Pile Wall; (e) Slurry (Diaphragm) Wall; (f) Secant Pile Wall; (g) Tangent Pile Wall; and (h) Interlocking H-Pile Wall (after Dismuke, 1991) .......................................................................... 8-2 Wall Support Systems: (a) Cantilever Wall; (b) Earth Berm Support; (c) Raker System; (d) Deadman Anchor; (e) Cross-Lot Braced Wall; and (f) Anchored Wall (after NAVFAC, 1986) ...... 8-3 (a) Sheet Pile Wall for Earth Support Behind a Cast-in-Place Wall; (b) Cofferdam for Construction of Foundations in Water; (c) Cofferdam for Footing Construction on Land; (d) Anchored Bulkhead; and (e) Bridge Abutment............................................................................ 8-5 Steel Sheet Pile Sections Commonly Used for Retaining Walls and Cofferdams. (a) Z -Section, (b) U-Section, (c) Cold Formed Section ............................................................................................ 8-6 Vinyl Sheet Pile Walls....................................................................................................................... 8-7 Sequence of Construction for a Backfilled Sheet Pile Structure (after Das, 1990)............................ 8-8 Sequence of Construction for an Excavated Sheet Pile Structure...................................................... 8-9 Sheet Pile Pitch and Drive Method (after Tespa, 2001)..................................................................... 8-9 Sheet Pile Panel Driving (after Tespa, 2001)....................................................................................8-10 Sheet Pile Staggered Driving (after Tespa, 2001).............................................................................8-11 Soldier Pile and Lagging Wall and Bracing for Temporary Excavation Support .............................8-14 Permanent Soldier Pile and Lagging Walls for (a) Roadway Embankment, and (b) Roadway Cut...............................................................................................................................8-14 Types of Soldier Piles (a) Wide Flange Section, (b) Pipe Section Without Anchor, (c) Double Channel Section, and (d) Pipe Section with Anchor (after Xanthakos, et al. 1994) .........................8-15 Contact Lagging................................................................................................................................8-16 Soldier Pile and Lagging Wall: (a) Drilling for Pile Installation, and (b) Excavation between Soldier Piles in Cohesive Soil for Installation of Lagging................................................................8-18 Typical Construction Sequence for a Slurry Wall: (a) Excavation; (b) Insertion of Steel Tubing (End Stops); (c) Placement of Reinforcement Cage; and (d) Concrete Placement (Xanthakos, 1994).............................................................................................................................8-21 Conventional Reinforced Concrete Wall (Tamaro, 1990) ................................................................8-24 Soldier-Pile-Tremie-Concrete (SPTC) Wall (Tamaro, 1990) ...........................................................8-25 Precast-Concrete-Panel Wall (Tamaro, 1990) ..................................................................................8-25 Post-Tensioned-Concrete Wall (Tamaro, 1990) ...............................................................................8-26 Slurry Trench Excavation Equipment: (a) Hydraulically Operated Clamshell Bucket, and (b) Hydromill ....................................................................................................................................8-29 Cross-Section of a Guide Wall (a) Compact Cohesive Soil, (b) Loose Cohesionless Soil (Goldberg, et al., 1976).....................................................................................................................8-30 Slurry Wall Reinforcement Cage (a) on Fabrication Bed, Showing Styrofoam Knock-Out Panel, and (b) During Lifting for Installation into Slurry Filled Trench......................................................8-32 Reinforced Panel in Cast-in-Place Slurry Wall (Tamaro and Poletto, 1992)....................................8-32 Slurry Filled Trench with Tremie Pipes Just Prior to Concrete Placement.......................................8-33


Figure 8-4 Figure 8-5 Figure 8-6 Figure 8-7 Figure 8-8 Figure 8-9 Figure 8-10 Figure 8-11 Figure 8-12 Figure 8-13 Figure 8-14 Figure 8-15 Figure 8-16

Figure 8-17 Figure 8-18 Figure 8-19 Figure 8-20 Figure 8-21 Figure 8-22 Figure 8-23 Figure 8-24 Figure 8-25


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Figure 8-26

Tangent Pile Wall a) with Structural Steel Section As Reinforcement, and b) Face of Completed Wall ................................................................................................................................8-37 Figure 8-27 Various Configurations of Bored Pile Walls: (a) Tangent Pile Wall; (b) Staggered Tangent Pile Wall; (c) Secant Pile Wall; (d) Intermittent Pile Wall with Grouted Openings; and (e) Intermittent Pile Wall with Lagging............................................................................................8-38 Figure 8-28 Construction Sequence (a) Tangent Pile and (b) Secant Pile Walls (after Xanthakos, 1994)...........8-39 Figure 8-29 Grout Columns Layouts on Jet-Grouted Walls.................................................................................8-43 Figure 8-30 Jet-Grouted Wall Applications for Excavation Support, Underpinning, Settlement Control, and Water Control.............................................................................................................................8-44 Figure 8-31 Range of Soil Types Treatable by Chemical and Jet-Grouting (after Welsh et al., 1986) ................8-45 Figure 8-32 Jet Grouting Procedure (after Pacchiosi, 1985) ................................................................................8-46 Figure 8-33 Details of Jet Grouting Monitors (a) Single Fluid System and (b) Triple Fluid System...................8-47 Figure 8-34 Schematic of Jet Grouting Systems (a) Single Fluid System, (b) Double Fluid System, and (c) Triple Fluid System .....................................................................................................................8-49 Figure 8-35 Jet Grout Set Up................................................................................................................................8-51 Figure 8-36 Grout Mixing and Injection Plant .....................................................................................................8-51 Figure 8-37 Drilling and Grouting Rig.................................................................................................................8-52 Figure 8-38 Forming a Jet-Grouted Column ........................................................................................................8-52 Figure 8-39 Jet Grouting Monitor ........................................................................................................................8-52 Figure 8-40 Rods for a Triple Fluid Grouting System..........................................................................................8-53 Figure 8-41 Plan View of Typical DMM Wall Layouts: (a) Cut-Off Wall, (b) and (c) Excavation-Support Wall, and (d) Lattice Pattern for Liquefaction Control.....................................................................8-58 Figure 8-42 Various DMM Wall Applications: (a) as Containment/Cutoff Wall and (b) as Structural Retaining Wall ..................................................................................................................................8-58 Figure 8-43 DMM Equipment (a) Mixing Shaft for General Use, (b) Mixing Shaft for Soil with Boulders, and (c) Mixing Shaft for Cohesive Soil (after Taki and Yang, 1991)...............................................8-61 Figure 8-44 Drilling and Mixing Unit ..................................................................................................................8-62 Figure 8-45 Mixing Plant (Courtesy of MnDOT) ................................................................................................8-62 Figure 8-46 DMM Installation Procedure (after Taki and Yang, 1991)...............................................................8-63 Figure 8-47 Components of a Ground Anchor (after Sabatini et al., 1999)..........................................................8-67 Figure 8-48 Main Types of Grouted Ground Anchors (after Littlejohn, 1990)....................................................8-70 Figure 8-49 Cut Away View of Bar Tendon (after Sabatini et al., 1999).............................................................8-72 Figure 8-50 Cut Away View of Strand Tendon (after Sabatini et al., 1999) ........................................................8-72 Figure 8-51 Construction Sequence for Permanent Soldier Pile and Lagging Wall (after Sabatini et al., 1999)................................................................................................................8-74 Figure 8-52 Drilling Equipment for Installation of Ground Anchors: (a) Hollow Stem Auger and (b) Down-The-Hole Hammer ...........................................................................................................8-75 Figure 8-53 Installation of Anchor Tendon ..........................................................................................................8-75 Figure 8-54 Typical Equipment for Load Testing of (a) Strand Ground Anchor and (b) Bar Ground Anchor....8-77 Figure 8-55 Typical Plots of Tendon Movement for (a) Performance Test, and (b) Proof Test (after PTI, 1996) ...............................................................................................................................8-81 Figure 8-56 Typical Deformation Conditions and Pressure Distribution for Cantilever Walls: (a) Yielding Pattern of Cantilever Wall Penetrating a Sand Layer; (b) Net Actual Earth Pressure Distribution; (c) Simplified Net Earth Pressure Distribution (after Das, 1990) .....................................................8-88 Figure 8-57 Design analysis for Nongravity Cantilevered (Sheet Pile) Wall.......................................................8-91 Figure 8-58 Variation of Deflection and Bending Moment (a) Free Earth Support Method and (b) Fixed Earth Support Method (after Das, 1990)...........................................................................................8-96 Figure 8-59a analysis by Free Earth Support Method for Sheet Piling in Granular Soils (after Teng, 1962 and USS, 1975) ....................................................................................................8-98 FHWA NHI-07-071 Earth Retaining Structures

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Figure 8-60a Moment Reduction for Anchored Sheet Pile Wall analyzed by Free Earth Support Method for Granular Soils (after Rowe, 1952; Rowe, 1957; and Das, 1990)....................................................8-100 Figure 8-61 analysis by Equivalent Beam Method (after USS, 1975 and Teng, 1962)......................................8-103 Figure 8-62 Effect of Anchor Location Relative to the Wall (after NAVFAC, 1986) .......................................8-104 Figure 8-63 Examples of Corrosion Protection for Anchorages (a) Strand Tendon and (b) Bar Tendon (after Sabatini et al., 1999)..............................................................................................................8-107 Figure 8-64 Simple Corrosion Protected Tendons (a) Strand Tendon and (b) Bar Tendon (after Sabatini et al., 1999)..............................................................................................................8-108 Figure 8-65 Encapsulated Double Corrosion Protection (a) Strand Tendon and (b) Bar Tendon (after Sabatini et al., 1999)..............................................................................................................8-109 Figure 8-66 Decision Tree for Selection of Corrosion Protection Level (after PTI, 1996) ................................8-112 Figure 8-67 Apparent Earth Pressure Diagram for Sand....................................................................................8-115 Figure 8-68 Apparent Earth Pressure Diagram for Stiff to Hard Clays..............................................................8-116 Figure 8-69 Apparent Earth Pressure Diagram for Soft to Medium Clay ..........................................................8-117 Figure 8-70 Calculation of Nominal Anchor Loads for one-Level Wall............................................................8-119 Figure 8-71 Calculation of Nominal Anchor Loads for Multi-Level Wall.........................................................8-120 Figure 8-72 Pressure Diagram for Temporary Continuous Wall in Soft to Medium Clay (after AASHTO, 2007) ...................................................................................................................8-121 Figure 8-73 Calculation of Wall Bending Moments Using Hinge Method (after Sabatini et al., 1999) ............8-129 Figure 8-74 Calculation of Wall Bending Moments Using Tributary Area Method (after Sabatini et al., 1999)..............................................................................................................8-130 Figure 8-75 analysis of Basal Stability (modified after Terzaghi et al., 1996.)..................................................8-134 Figure 8-76 Failure Surfaces for External Stability Evaluations ........................................................................8-135 Figure 8-77 a and B. Evaluation of Externally Supported Wall Movements (after Clough and O’Rourke, 1990) ................................................................................................8-138 Figure 8-78 Model for Soil-Structure Interaction analysis .................................................................................8-142 Figure 8-79 Significant Features of Excavation Geometry in Relation to Groundwater Control (after ASCE, 1997) .........................................................................................................................8-144 Figure 8-80 Significant Features of Excavation Geometry in Relation to Groundwater Control (after ASCE, 1997) .........................................................................................................................8-145 Figure 8-81 Significant Features of Excavation Geometry in Relation to Groundwater Control (after ASCE, 1997) .........................................................................................................................8-146 Figure 8-82 Significant Features of Excavation Geometry in Relation to Groundwater Control (after ASCE, 1997) .........................................................................................................................8-147 Figure 8-83 Diagram Illustrating a Ground Anchor T-y Curve (after Weatherby, 1998) ..................................8-150 Figure 8-84 DMM Wall Design: (a) analysis for Punch-through Shear; (b) analysis for Compressive Action of Arching Effects; (c) Empirical Guideline for Avoiding Bending Failure (after Taki and Yang, 1991)............................................................................................................8-159 Figure 9-1 Figure 9-2 Figure 9-3 Figure 9-4 Figure 9-5 Figure 9-6 Figure 9-7

Soil Nail Wall Application (a) Temporary Shoring; (b) Roadway Widening Under Existing Bridge; (c) Slope Stabilization; and (d) Roadway Cut (after Porterfield et al., 1994) .................................... 9-2 Typical Nail Wall Construction Sequence (after Porterfield et al., 1994) ......................................... 9-4 Soil Nail Wall Applications (a) Temporary Shoring, (b) Roadway Widening Under Existing Bridge, (c) Roadway Cut, and (d) Slope Stabilization......................................................... 9-5 Main Components of a Typical Soil Nail (after Porterfield et al., 1994)..........................................9-15 Grout Placement (Tremie) through Pipe (after Porterfield et al., 1994) ...........................................9-18 Typical Pvc Centralizers attached to a Nail Bar Prior to Nail Installation (after Porterfield et al., 1994)............................................................................................................9-19 Initial Excavation Lift and Nail Installation (after Porterfield et al., 1994) ......................................9-22


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Figure 9-8 Figure 9-9 Figure 9-10 Figure 9-11 Figure 9-12 Figure 9-13 Figure 9-14 Figure 9-15 Figure 9-16 Figure 9-17 Figure 9-18 Figure 9-19 Figure 9-20 Figure 9-21 Figure 9-22 Figure 9-23 Figure 9-24 Figure 9-25 Figure 9-26 Figure 9-27 Figure 9-28 Figure 9-29 Figure 9-30 Figure 9-31 Figure 9-32 Figure 9-33 Figure 9-34 Figure 9-35 Figure 9-36 Figure 9-37 Figure 9-38 Figure 9-39 Figure 9-40 Figure 9-41 Figure 9-42 Figure 9-43 Figure 9-44 Figure 9-45 Figure 9-46 Figure 9-47 Figure 9-48 Figure 9-49

Examples of Alternative Temporary Excavation Support (a) Stabilizing Berm and (b) Slot Excavation (after Porterfield et al., 1994)............................................................................9-23 Typical Drilling of Soil Nails with Rotary Method (after Porterfield et al., 1994)...........................9-24 Shotcrete Temporary Facing (after Porterfield et al., 1994) .............................................................9-26 Headed-Studs Welded to Bearing Plate (after Porterfield et al., 1994).............................................9-28 Cast-in-Place Facing .........................................................................................................................9-29 Precast Panel Facing (after Elias et al., 2001)...................................................................................9-29 Grouted Epoxy-Coated Nail (after Byrne et al., 1998) .....................................................................9-32 Sheathing-Encapsulation on Grouted Nail (after Byrne et al., 1998) ...............................................9-33 Potential Failure Surfaces and Soil Nail Tensile Forces (after Lazarte et al., 2003).........................9-36 Potential Critical Stability During Construction (after Lazarte et al., 2003).....................................9-37 Principal Modes of Failure of Soil Nail Wall Systems (after Lazarte et al., 2003)...........................9-38 Global Stability analysis of Soil Nail Wall Using a Single-Wedge Failure Mechanism (after Lazarte et al., 2003).................................................................................................................9-39 Sliding Stability of a Soil Nail Wall (after Lazarte et al., 2003).......................................................9-42 Bearing Capacity (Heave) analysis (after Terzaghi et al., 1996) ......................................................9-43 Single Nail Stress-Transfer Mode (after Lazarte, 2003) ...................................................................9-46 Soil Nail Stress-Transfer Mechanism (after Lazarte, 2003) .............................................................9-48 Simplified Distribution of Nail Tensile Force (after Lazarte, 2003).................................................9-49 Facing Connection Failure Modes (after Lazarte, 2003) ..................................................................9-52 Progressive Flexural Failure in Wall Facings (after Lazarte, 2003) .................................................9-54 Geometry Used in Flexural Failure Mode (after Lazarte, 2003).......................................................9-56 Punching Shear Failure Modes (after Lazarte, 2003) .......................................................................9-60 Geometry of a Headed-Stud (after Lazarte, 2003)............................................................................9-61 Drainage of Soil Nail Walls (after Lazarte et al., 2003) ...................................................................9-64 Typical Drain Pipe Details to Provide Groundwater Control in Soil Nail Walls (after Byrne et al., 1998)...................................................................................................................9-66 Deformation of Soil Nail Walls (after Byrne et al., 1996)................................................................9-68 Examples of Frost Protection of Soil Nail Walls (after Lazarte et al., 2003)....................................9-71 Soil Nail Patterns on Wall Face (after Lazarte et al., 2003)..............................................................9-75 Varying Soil Nail Patterns (after Lazarte, 2003) ..............................................................................9-77 Batter 0o – Backslope 0o (after Lazarte et al., 2003)........................................................................9-81 Batter 0o – Backslope 10o (after Lazarte et al., 2003)......................................................................9-82 Batter 10o – Backslope 0o (after Lazarte et al., 2003)......................................................................9-83 Batter 10o – Backslope 10o (after Lazarte et al., 2003)....................................................................9-84 Batter 0o – Backslope 30o (after Lazarte et al., 2003)......................................................................9-85 Batter 10o – Backslope 30o (after Lazarte et al., 2003)....................................................................9-86 Correction Factors for Use in Design Chart Solutions (after Lazarte, 2003)....................................9-89 Details of Typical Soil Nail Test Set-Up (Porterfield et al., 1994) .................................................9-101 Typical Data Sheet for Soil Nail Load Testing (after Porterfield et al., 1994) ...............................9-103 Example of Data Reduction from Soil Nail Load Testing (after Porterfield et al., 1994) ..............9-104 Example of Data Reduction from Soil Nail Creep Testing (after Poterfield et al., 1994) ..............9-105 Micropile Wall Cross for Wall 600, Portland, Oregon ...................................................................9-110 Typical Micropile Construction Sequence Using Casing ...............................................................9-111 Cross Section Showing Steep Canyon Slope and Temporary Micropile Shoring (after Macklin et al., 2004) .............................................................................................................9-113


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Figure 10-1 Figure 10-2 Figure 10-3 Figure 10-4 Figure 10-5 Figure 10-6 Figure 10-7 Figure 10-8 Figure 10-9 Figure 10-10

Wall Selection Flow Chart................................................................................................................10-2 MSE Wall on Top of Cantilever Wall (after CDOT, 2003)............................................................10-22 L-Wall on Top of MSE Wall (after CDOT, 2003)..........................................................................10-22 Gabion Wall Anchored with Geogrid (after CDOT, 2003) ............................................................10-23 anchored L-Wall (after CDOT, 2003).............................................................................................10-24 anchored Cantilever Wall (after CDOT, 2003)...............................................................................10-25 Composite T-Wall (after The Neel Company, 2004)......................................................................10-26 Soil Nail and MSE Wall (after Turner and Jensen, 2005) ..............................................................10-27 Soil Nail and Fascia Wall with Geofoam Backfill (after Samtani, Personal Communication).......10-28 Example SMSE Wall ......................................................................................................................10-29


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CHAPTER 1 INTRODUCTION 1.1

PURPOSE AND SCOPE

Earth retaining systems (or retaining walls) are used to hold back earth and maintain a difference in the elevation of the ground surface as shown in Figure 1-1. The retaining wall is designed to withstand the forces exerted by the retained ground or “backfill”, and to transmit these forces safely to a foundation and/or to the portion of restraining elements located beyond the failure surface.

Figure 1-1. Schematic of a Retaining Wall and Common Terminology. In general, the cost of constructing a retaining wall is usually high compared with the cost of forming a new slope. Therefore, the need for a retaining wall should be assessed carefully during preliminary design and an effort should be made to keep the retained height as low as possible. In highway construction, retaining walls are used along cuts or fills where space is inadequate for construction of cut slopes or embankment slopes. Bridge abutments and foundation walls, which must support earth fills, are also designed as retaining walls. FHWA NHI-07-071 Earth Retaining Structures

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1 - Introduction June 2008

Typical earth retaining structure applications for highway construction include: • • • • • • • • • •

new or widened highways in developed areas; new or widened highways at mountain or steep slopes; grade separation; bridge abutments, wing walls and approach embankments; culvert walls; tunnel portals and approaches; flood walls, bulkheads and waterfront structures; cofferdams for construction of bridge foundations; stabilization of new or existing slopes and protection against rockfalls; and groundwater cut-off barriers for excavations or depressed roadways.

Figure 1-2 provides schematic illustrations of several retaining wall systems used in highway applications. A great number of wall systems have been developed in the past 30 years by specialty contractors who have been promoting either a special product or a specialized method of construction, or both. Due to the rapid development of these diversified systems and their many benefits, the design engineer is now faced with the difficult task of having to select the best possible system; design the structure; and ensure its proper construction. The purpose of this module is to provide the practicing engineer with a thorough understanding of various retaining walls and their application in highway facilities.

1.2

HISTORY OF EARTH RETAINING SYSTEMS

Historical reviews of the developments in retaining walls have been presented by Kerisel (1992), Gould (1990), and O’Rourke and Jones (1990). Following is a brief review of these historical developments. Examples of wall construction have been traced back to approximately 3000 B.C. (Kerisel, 1992). These earlier walls were primarily gravity structures that relied on self-weight to resist earth pressures. They were constructed of stone masonry (with or without mortar) or various types of cribs or bins with different filling materials. The development of “modern” retaining walls gained impetus in the late 1940s, 1950s and 1960s due to innovations in construction technology. New construction methods were developed at that time, largely through adaptations and improvements in specialized excavation and drilling equipment. In that period, North America was introduced to new forms of cast-in-place below-grade walls, tiebacks (or ground anchors) in soil and rock (both FHWA NHI-07-071 Earth Retaining Structures

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(a) Externally Stabilized Systems

(b) Internally Stabilized Systems Figure 1-2. Variety of Retaining Walls (after O’Rourke and Jones, 1990).

temporary and permanent), and the concept of earth reinforcement (Vidal, 1966) which led to a wide variety of earth retaining systems (ERS). Most of the procedures now being utilized for construction of ERS were well developed by 1970 or had appeared in full-scale experiments. FHWA NHI-07-071 Earth Retaining Structures

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An important breakthrough in the design of ERS which occurred in this era was the recognition that the earth pressure acting on a wall is a function of the type of wall and the amount and distribution of wall movement. Classical earth pressure theories, which were developed by Coulomb (1776) and Rankine (1857), were formalized for use by Caquot and Kerisel (1948) and others. Sophisticated analyses of soil-structure interaction and wall/soil movements began in the 1960s using finite difference and finite element analytical procedures. The simultaneous advancement of geotechnical instrumentation equipment and monitoring procedures made the “observational method” of design (Peck, 1969) popular and cost effective. Since 1970 there has been a dramatic growth in the methods and products for retaining soil. O’Rourke and Jones (1990) describe two trends in particular which have emerged since 1970. First, there has been an increasing use of reinforcing elements, either by incremental burial to create reinforced soils (MSE walls), or by systematic in situ installation to reinforce natural soils (soil nailing). Mechanically stabilized earth and soil nailing have changed the ways we construct fill or cut walls, respectively, by providing economically attractive alternatives to traditional construction methods. Second, there has been an increasing use of polymeric products to reinforce the soil and control drainage. Rapid developments in polymer manufacturing have supplied a wide array of geosynthetic materials. The use of these products in construction has encouraged a multitude of different earth retention schemes. The rapid development of these new trends and the increased awareness of the impact of construction on the environment, have led to the emergence of the concept of “earth walls”. In this concept, the soil supports itself or is incorporated into the structure and assumes a major structural or load carrying function. With this concept, structural member requirements of the system are reduced, or eliminated altogether. Examples of recently developed earth walls include the soil-reinforcement systems discussed above, as well as systems involving chemical treatment of the in-situ soil such as jet grouting or deep mixing.

1.3

CLASSIFICATION OF EARTH RETAINING STRUCTURES

In this manual, earth retaining systems may be classified according to: • • •

Load support mechanism, i.e., externally or internally stabilized walls; Construction concept, i.e., fill or cut walls; and System rigidity, i.e., rigid or flexible walls.


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It is noted that each retaining wall can be classified using these three factors. For example, a sheet-pile wall would be classified as an externally-stabilized cut wall which is relatively flexible. A mechanically stabilized earth (MSE) wall is an internally stabilized fill wall which is relatively flexible. Further description of these classifications is provided subsequently. 1.3.1

Classification by Load Support Mechanism

For the purpose of this manual, the walls have been organized according to two principal categories: externally and internally stabilized systems (O’Rourke and Jones, 1990). An externally stabilized system uses an external structural wall, against which stabilizing forces are mobilized. An internally stabilized system involves reinforcements installed within and extending beyond the potential failure mass. Hybrid systems combine elements of both internally and externally supported walls. Figure 1-3 presents the organization of walls used in this manual. Virtually all traditional types of walls may be regarded as externally stabilized systems. Gravity walls, in the form of cantilever structures or gravity elements (e.g., bins, cribs and gabions), support the soil through weight and stiffness to resist sliding, overturning, and shear. Bracing systems, such as cross-lot struts and rakers, provide temporary support for in situ structural and chemically stabilized walls. Ground anchors provide support through the pullout capacity of anchors established in stable soil outside of the zone of potential failure. It is in the area of internally stabilized systems that a relatively new concept has been introduced. Shear transfer to mobilize the tensile capacity of closely spaced reinforcing elements has enabled retaining structures to be constructed without a structural wall element, and has substituted instead a composite system of reinforcing elements and soil as the primary structural entity. A facing is required on an internally stabilized system, however, its purpose is to prevent raveling and deterioration, rather than to provide primary structural support. 1.3.2

Classification by Construction Concept

Earth retaining systems can also be classified according to the construction method, i.e., fill construction or cut construction. Fill wall construction refers to a wall system in which the wall is constructed from the base of the wall to the top (i.e., “bottom-up” construction). Cut wall construction refers to a wall system in which the wall is constructed from the top of the wall to the base (i.e., “top-down” construction) concurrent with excavation operations. The classification of each wall system according to its construction concept is also presented in Figure 1-3. FHWA NHI-07-071 Earth Retaining Structures

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1-6

Figure 1-3. Classification of Earth Retaining Systems (after O’Rourke and Jones, 1990). 1 - Introduction June 2008

It is important to recognize that the “cut” and “fill” designations refer to how the wall is constructed, not necessarily the nature of the earthwork (i.e., cut or fill) associated with the project. For example, a prefabricated modular gravity wall, which may be used to retain earth for a major highway cut, is considered a fill wall as it is constructed “bottom-up” after the excavation for the cut has reached its final grade. 1.3.3

Classification by System Rigidity

The rigidity (or flexibility) of a wall system is fundamental to the understanding of the development of earth pressures (discussed in Chapter 3). In simple terms, a wall is considered to be rigid if it moves as a unit (rigid body rotation and/or translation) and does not experience bending deformations. Most gravity walls can be considered rigid walls. Flexible walls are those walls which undergo bending deformations in addition to rigid body motion. Such deformations result in a redistribution of lateral pressures from the more flexible to the stiffer portions of the system. Virtually all wall systems, except the gravity walls, may be considered to be flexible. 1.3.4

Temporary and Permanent Wall Applications

The focus of this document is on design methods and procedures for permanent ERS. Permanent systems are generally considered to have a service life of 75 to 100 years. However, the ERS listed in Figure 1-3 are technically feasible for both temporary and permanent applications. In most cases, however, certain systems may not be cost-effective for temporary applications. Compared to permanent walls, walls used for temporary applications generally have less restrictive requirements on material durability, design factors of safety (or higher resistance factors), performance, and overall appearance. Also, walls that can be constructed rapidly are often used for temporary applications. For example, MSE walls with segmental, precast facings are not typically used for temporary applications since the cost of the facing components and the select backfill may be more than 50 percent of the total cost of the wall. The service life of temporary earth support systems is based on the time required to support the ground while the permanent systems are installed. This document has adopted the American Association of State Highway and Transportation Officials (AASHTO) guidance which considers temporary systems to be those that are removed or abandoned upon completion of the permanent systems. The time period for temporary systems is commonly stated to be 18 to 36 months but may be shorter or longer based on actual project conditions. Furthermore temporary systems may be divided into “support of excavation” (SOE) temporary systems and “critical” temporary systems. In general the owner will determine FHWA NHI-07-071 Earth Retaining Structures

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which temporary systems are to be designated as critical. Often that decision is based on the owner’s need to restrict lateral movement of the support system to minimize ground movements behind the support system. In general, specific components or design features for temporary systems may be designed to the same or similar criteria used for permanent systems. Conversely, SOE systems are commonly designed to less restrictive criteria than permanent systems. The owner commonly assigns the responsibility for design and performance of SOE systems to the contractor. The design of these SOE systems is often based more on system stability than on minimizing ground movements. Whenever appropriate in this document, discussion is provided concerning the differences in design requirements for SOE systems and permanent systems.

1.4

ORGANIZATION OF THE MANUAL

This manual is intended to be a stand-alone document and is geared towards providing the practicing engineer with a thorough understanding of the various types of retaining walls used for highway applications. Accordingly, the manual starts with a discussion of earth pressure theories and the determination of basic soil parameters, and then proceeds to discuss each of the earth retaining systems identified in Section 1.3. An objective of this manual is to provide sufficient information to enable the engineer to systematically review the feasibility of wall systems for a specific project application and to ultimately select a technically feasible and cost-effective wall system for a project. The organization of the manual is presented below. •

Chapter 2 presents information on fundamental concepts of soil shear strength evaluation for wall systems and on the evaluation and selection of soil and rock parameters required for ERS design. This chapter also describes planning a subsurface investigation program.

•

Chapter 3 describes the basic principles for the evaluation of lateral earth pressures and water pressures used in ERS design, including specific considerations for earth pressures in temporary and permanent wall applications and for earth pressures in cohesive backfills. Other topics discussed include earth pressures from surcharge loads, compaction, seismic forces, and swelling soils.

•

Chapter 4 provides an overview of load and resistance factor design (LRFD) for retaining walls. This methodology represents a change from the more traditional (and well-known to most engineers) allowable stress design (ASD) method. A


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comprehensive coverage of LRFD for highway substructures (including retaining walls) is provided in NHI Course No. 130082 “LRFD for Highway Bridge Substructures.” and the AASHTO LRFD Bridge Design Specifications (2007). •

Chapters 5 through 9 provide information on gravity and semi-gravity walls, modular gravity walls, MSE walls, nongravity cantilever walls and anchored walls, and in-situ reinforced walls. The general coverage of information for each wall type includes: o background and feasibility; o construction materials and techniques; o cost information; o design and performance information; and o required elements of construction inspection. The design procedures presented are developed using the LRFD method and are consistent with the most recent AASHTO LRFD Bridge Design Specifications (AASHTO, 2007). Of the wall types described in this manual, in-situ reinforced walls (i.e., soil nail walls and micropile walls) are not yet covered in the AASHTO (2007) specifications. These wall types are described in Chapter 9 and design information is presented in ASD.

1.5

•

Chapter 10 presents a generalized procedure used to select technically viable and cost-effective wall systems for a specific project. An example wall selection evaluation for a specific project is provided in this chapter. This chapter also provides information on so-called hybrid wall systems. Hybrid wall systems generally comprise a combination of cut and fill wall systems or combinations of material components from more than one wall type.

•

Chapter 11 provides information on contracting methods and bidding documents used for the procurement of earth retaining systems. Typical contracting approaches and key plan and specification items and some of the identified resources for wall specifications are provided.

PRIMARY REFERENCES FOR THIS MANUAL

A detailed list of references is provided in Chapter 12, however; several primary references were used in the preparation of this manual. Following is a listing of these primary references.


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AASHTO. (2002). “Standard Specifications for Highway Bridges.” 17th Edition, American Association of State Highway and Transportation Officials, Washington, D.C. AASHTO (2007). “LRFD Bridge Design Specifications.” 4th Edition, American Association of State Highway and Transportation Officials, Washington, D.C. Byrne, R.J., Cotton, D., Porterfield, J., Wolschlag, C., and Ueblacker, G. (1996). “Manual for Design and Construction Monitoring of Soil Nail Walls.” Federal Highway Administration, FHWA-SA-96-069. Cheney, R. S. (1988). “Permanent Ground Anchors.” Federal Highway Administration, Report, FHWA-DP-68-1R, Washington, D. C. Christopher, B. R., Gill, S. A., Giroud, J. P., Juran, I., Mitchell, J. K., Schlosser, F., and Dunnicliff, J. (1990). “Design and Construction Guidelines for Reinforced Soil Structure - Volume 1.” FHWA-RD-89-043, Federal Highway Administration, Washington, D. C. Clouterre. (1991). “Recommendations Clouterre - Soil Nailing Recommendations.” French National Research Project, FHWA-SA-93-026, Federal Highway Administration, Washington, D.C. (in English). Elias, V. (2000). “Corrosion/Degredation of Soil Reinforcements for Mechanically Stabilized Earth Walls and Reinforced Soil Slopes,” Federal Highway Administration, Report FHWA-NHI-00-044, Washington, D. C. Elias, V., Christopher, B. R., and Berg, R. R. (2001). “Mechanically Stabilized Earth Walls and Reinforced Soil Slopes Design and Construction Guidelines.” Federal Highway Administration, Report FHWA-NHI-00-043, Washington, D. C. Goldberg, D. T., Jaworski, W. E., and Gordon, M. D. (1976). “Lateral Support Systems and Underpinning.” Vol.1 Design and Construction, April, FHWA-RD-75-128, Federal Highway Administration, Washington, D. C. Holtz, R. D., Christopher, B. R., and Berg, R. R. (1998). Geosynthetic Design and Construction Guidelines, May, Federal Highway Administration, Report FHWA-HI98-038, McLean, Va. Kimmerling, R. E. (2002). “Shallow Foundations.” Geotechnical Engineering Circular No. 6, FHWA-IF-02-054, Federal Highway Administration, Washington, D.C. FHWA NHI-07-071 Earth Retaining Structures

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Lazarte, C. A., Elias, V., Espinoza, R. D., and Sabatini, P.J. (2003). “Soil Nail Walls.” Geotechnical Engineering Circular No. 7, Federal Highway Administration, Report FHWA-IF-03-017, Washington, D.C. Mayne, P. W., Christopher, B. R., and DeJong, J. (2001). “Manual on Subsurface Investigations”, National Highway Institute Publication, Federal Highway Administration, Report FHWA-NHI-01-031, Washington, D.C. Porterfield, J. A., Cotton, D. M. and Byrne, R. J. (1994). “Soil Nailing Field Inspectors Manual.” FHWA-SA-93-068, Federal Highway Administration, Washington, D. C. Sabatini, P. J., Bachus, R. C., Mayne, P. W., Schneider, J. A., and Zettler, T. E. (2002). “Evaluation of Soil and Rock Properties”, Geotechnical Engineering Circular, No. 5, U.S. Department of Transportation, Federal Highway Administration, FHWA-IF-02034, Washington, D.C. Sabatini, P. J., Pass, D. G. and Bachus, R. C. (1999). “Ground Anchors and Anchored Systems”, Geotechnical Engineering Circular, No. 4, U.S. Department of Transportation, Federal Highway Administration, FHWA-SA-99-015. Sabatini, P. J., Elias, V., Schmertmann, G. R., and Bonaparte, R. (1997). “Earth Retaining Systems”, Geotechnical Engineering Circular, No. 2, Federal Highway Administration, U.S. Department of Transportation, FHWA-SA-96-038, Washington, D.C. Seismic Design of Highway Bridges. (1986). (NHI Course No. 13048), FHWA-93-040, Federal Highway Administration, Washington, D. C. Weatherby, D.E. (1982). “Tiebacks.” Federal Highway Administration, Report FHWA-RD82-047, McLean, VA.


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CHAPTER 2 EVALUATION OF GEOTECHNICAL PARAMETERS FOR EARTH RETAINING SYSTEMS 2.1

INTRODUCTION

This chapter presents information on the evaluation of soil and rock parameters required to perform earth retaining system (ERS) design and to evaluate ERS constructability. These required parameters include subsurface stratigraphy and index and performance properties of in-situ soils (e.g., foundation soils below a gravity retaining structure or retained ground behind a cut wall system), fill wall backfill soils, and rock within the retained ground or as a foundation material. For typical highway projects, one or more retaining walls may be part of a larger project scope in which a detailed geotechnical investigation is performed. Herein, information is presented on the development of a subsurface investigation and laboratory testing program focused on obtaining information necessary for the design and construction of fill and cut wall systems. As indicated above, this chapter describes the evaluation of design properties for wall systems. Of particular importance for walls (like most geotechnical features) is the appropriate evaluation of soil shear strength. In this chapter, background information on the affects of the rate of loading on soil shear strength and drained versus undrained shear strength parameters are discussed, along with common methods to evaluate soil shear strength. A more detailed coverage of the evaluation of soil and rock parameters for the design of geotechnical features is provided in the following sources: • • •

NHI Course No. 132031, Subsurface investigations and accompanying reference manual, FHWA NHI-01-031 (Mayne et al., 2001); and Geotechnical Engineering Circular No. 5, Evaluation of soil and rock properties, FHWA IF-02-034 (Sabatini et al., 2002). Section 10.4 of AASHTO LRFD Bridge Design Specifications (AASHTO, 2007).


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2 – Geotechnical Parameters June 2008

2.2

PLANNING A SUBSURFACE INVESTIGATION AND LABORATORY TESTING PROGRAM

2.2.1

General

Planning of a subsurface investigation and laboratory testing program for a wall project requires that the engineer be aware of parameters needed for design and construction, as well as having an understanding of the geologic conditions and site access restrictions. Specific planning steps for a subsurface investigation and laboratory testing program include: (1) identify data needs and review available information; (2) develop and conduct a site investigation program including a site visit/reconnaissance and collection of disturbed and undisturbed soil and rock samples; and (3) develop and conduct a laboratory-testing program. Specific planning steps are addressed in the following sections. 2.2.2

Identify Data Needs and Review Available Information

The first step of an investigation and testing program requires that the engineer understand the project requirements and the site conditions and/or restrictions. The extent of the investigation should be consistent with the project scope (i.e., location, size, risk, and budget), the project objectives (i.e., purpose of the wall system), and the project constraints (i.e., geometry, constructability, performance, aesthetics, and environmental impact). The ultimate goal of this phase is to identify geotechnical data needs for the project and potential methods available to assess these needs. During this phase it is necessary to: • • • • • • • •

identify design and constructability requirements; identify performance criteria and schedule constraints; identify areas of concern on site and potential variability of local geology; develop likely sequence and phases of construction; identify engineering analyses to be performed; identify engineering properties and parameters required for these analyses; evaluate methods to obtain parameters and assess the validity of such methods for the material type and construction methods; and evaluate number of tests/samples needed and appropriate locations for them.

As an aid to assist in the planning of site investigation and laboratory testing, Table 2-1 provides a summary of the information needs and testing considerations for cut and fill wall applications. Detailed descriptions of the field and laboratory tests listed in Table 2-1 are provided in Chapter 4 of Geotechnical Engineering Circular (GEC) No. 5 (Sabatini et al., 2002). FHWA NHI-07-071 Earth Retaining Structures

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Cut Walls

Fill Walls

Geotechnical Issues

internal stability external stability global stability excavation stability dewatering chemical compatibility of wall/soil lateral earth pressure down-drag on wall pore pressures behind wall obstructions in retained soil and along wall line

• • • • • •

• • • •

• •

• •

internal stability external stability global stability total settlement differential settlement horizontal deformation lateral earth pressures bearing resistance chemical compatibility with soil and wall materials pore pressures behind wall borrow source evaluation (available quantity and quality of borrow soil) obstructions dewatering/drainage

• • • • • • • • •

Engineering Evaluations

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• subsurface profile (soil, groundwater, rock) • shear strength of soil • horizontal earth pressure coefficients and unit weight • interface shear strength (soil and reinforcement) • chemical composition of soils • hydraulic conductivity of soil • geologic mapping including orientation and characteristics of rock discontinuities

• subsurface profile (soil, groundwater, rock) • horizontal earth pressure coefficients and unit weight • interface shear strengths (soil and reinforcement) • foundation soil/wall fill shear strengths • compressibility parameters (including consolidation, shrink/swell potential, and elastic modulus) • chemical composition of fill/ foundation soils • hydraulic conductivity of soils directly behind wall • time-rate consolidation parameters • geologic mapping including orientation and characteristics of rock discontinuities

Required Information for Analyses SPT (granular soils) CPT dilatometer (DMT) vane shear piezometers test fill nuclear density rock coring geophysical testing

• test cut to evaluate stand-up time (soil nail walls, soldier beam and lagging) • well pumping tests • piezometers • SPT (granular soils) • CPT • vane shear • dilatometer (DMT) • pullout tests (anchors, nails) • rock coring • geophysical testing

• • • • • • • • •

Field Testing

• • • • • • • • •

• • • •

• • • • • • • • •


triaxial tests direct shear grain size distribution Atterberg Limits pH, resistivity tests moisture content organic content hydraulic conductivity unit weight

1-D Oedometer triaxial tests direct shear tests grain size distribution Atterberg Limits pH, resistivity tests moisture content organic content moisture-density relationships (unit weight) swell potential hydraulic conductivity reinforcement pullout tests slake durability (if necessary)

Laboratory Testing

Table 2-1. Summary of Information Needs and Testing Considerations for Fill and Cut Wall Applications.

Before any equipment is mobilized to the site, existing data for the site (both regionally and locally) should be evaluated and the geotechnical and design engineers should conduct a site reconnaissance as logical initial steps in the investigation. Existing data and a site visit will provide information which can reduce the scope of the subsurface investigation, help guide the location of sampling and testing points, and reduce the amount of time in the field due to unexpected problems. Currently, many state DOTs are developing geotechnical management systems (GMS) to store historical drilling, sampling, and laboratory test data for locations in their states. Such data, if available, should be used to facilitate development of a testing program that is correctly focused and not redundant. A list of potential information sources for a project along with the type of information available is presented in Table 2-2. 2.2.3

Developing a Site Investigation Program

The planning phase involves decisions regarding sampling/investigation methods, boring locations, number of samples, number and types of laboratory tests, and the number of confirmatory samples. At this stage, the types of potential sampling/investigation methods should have been identified and assessed (see Mayne et al. (2001) for types and features of field subsurface investigation methods). For major retaining wall projects which may be defined, for example, as a wall with a height greater than 20 ft or for highly variable ground conditions, the investigation might consist of a preliminary investigation phase (which may be performed as part of a larger geotechnical investigation during the preliminary project phase) followed by a final design phase investigation. The preliminary phase includes a limited number of fairly widely-spaced borings to define overall subsurface conditions and to identify problem areas. The standard split-spoon sampler with SPT blowcount data is typically used for the preliminary phase. Based on the results of the preliminary phase, a more detailed site investigation and also a laboratory investigation program may prove necessary for final design. The type, number, location, and depth of investigation points are dictated, to a large extent, by the project stage (i.e., feasibility study, preliminary, or final design), availability of existing geotechnical data, variability of subsurface conditions, type of wall system, and other project constraints. For walls less than 100 ft long, at least one investigation point should be selected for each wall within the project. This point should be selected at a distance behind the wall line. For example, for anchored walls investigation points should be advanced behind the proposed wall line within the anchor bond zone, for soil nail walls investigation points should be selected at a distance behind the wall face of 1.0 to 1.5 times the height of the wall, and for MSE walls, investigation points should be advanced just behind and directly below the reinforced backfill zone. For retaining wall systems more than FHWA NHI-07-071 Earth Retaining Structures

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May provide information on nearby soil/rock type; strength parameters; hydrogeological issues; foundation types previously used; environmental concerns Provides information on nearby soil/rock type and characteristics; hydrogeological issues, environmental concerns Provide stratigraphy of the site and/or regional areas Varied quality from state to state Groundwater levels Identifies 100 to 500 yr. Floodplains near water bodies May prevent construction in a floodplain Provide information for evaluation of scour potential Identifies site soil types Permeability of site soils Climatic and geologic information Useful in urban areas Maps for many cities are continuous for over 100 yrs. Identifies building locations and type Identifies business type at a location (e.g., chemical plant) May highlight potential environmental problems at an urban site

•

•

• • • • • • • • • • • • • •

Water/Brine Well Logs

Flood Insurance Maps

Soil Survey


Sanborn Fire Insurance Maps

Geologic Reports and Maps

Prior Subsurface Investigation Reports

Topographic Maps

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Provides good index map of the site Allows for estimation of site topography Identifies physical features in the site area Can be used to assess access restrictions

• • • • •

Aerial Photographs

Identifies manmade structures Identifies potential borrow source areas Provides geologic and hydrological information which can be used as a basis for site reconnaissance Track site changes over time

Functional Use • • •

Source

State Library/Sanborn Company (www. Sanborncompany.com)


A 1929 Sanborn map of St. Louis shows that a lead smelter was on site for 10 years. This information prevents an investigation in a contaminated area.

The local soil survey provides information on nearsurface soils to facilitate preliminary borrow source evaluation

Prior to investigation, the flood map shows that the site is in a 100 yr floodplain and the proposed structure is moved to a new location

Federal Emergency Management Agency (FEMA), USGS, State/Local Agencies www.fema.gov

Local Soil Conservation Service

A boring log of a water supply well two miles from the site area shows site stratigraphy facilitating evaluations of required depth of exploration

A twenty year old report on regional geology identifies rock types, fracture and orientation and groundwater flow patterns

A five year old report for a nearby roadway widening project provides geologic, hydrogeologic, and geotechnical information for the area, reducing the scope of the investigation

Engineer identifies access areas/restrictions, identifies areas of potential slope instability; and can estimate cut/fill capacity before visiting the site

Evaluating a series of aerial photographs may show an area on site which was filled during the time period reviewed

Examples

State Geological Survey/National Resources

USGS and State Geological Survey www.usgs.gov

State DOTs, USGS, United States Environmental Protection Agency (USEPA)

USGS, State Geological Survey

Local Soil Consevation Office, United States Geological Survey (USGS), Local Library, Local and National aerial survey companies

Location

Table 2-2. Sources of Historical Site Data.

100 ft in length, investigation points should be selected both in front and behind the wall and should be spaced every 100 – 200 ft with locations alternating from in front of the wall to behind the wall. A typical boring layout for an anchored wall is shown in Figure 2-1.

Wall boring 1-2H

Figure 2-1. Boring Layout for Anchored Wall. The guidelines for depth interval selection should also be developed in recognition of specific site/project conditions and design property/parameter requirements. For preliminary screening, the subsurface investigation at the wall location should be carried to a depth of up to 2 times the wall height below the bottom of the wall or a minimum of 10 ft into bedrock. Investigation depth should be deep enough to fully penetrate soft highly compressible soils (e.g., peat, organic silt, soft fine grained soils) into competent material of suitable bearing capacity (e.g., stiff to hard cohesive soil, compact dense cohesionless soil, or bedrock). Such information is required to perform external stability analyses of all wall systems (i.e., investigate the stability of the ground outside and below the wall system). Groundwater table and perched groundwater zones must be evaluated as part of a subsurface investigation program for a wall system. In Chapter 3, information is presented on FHWA NHI-07-071 Earth Retaining Structures

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calculating water pressures against walls for hydrostatic conditions and for cases where seepage pressures (resulting from water flow) are encountered in design. The presence of ground water affects lateral pressures applied to the wall facing, drainage system design, and construction procedures. At a minimum, the following items need to be considered for wall systems that will be constructed within or near the water table: • • • • • •

corrosion potential of metallic components based on the acidity/alkalinity of the ground water reduction in frictional resistance between soil and structural components (e.g., anchors, soil nails, etc.) necessity for excavation dewatering and specialized drilling and grouting techniques (in the presence of water) liquefaction potential of loose, cohesionless soils rock/soil instability resulting from seepage forces seasonal high and low groundwater levels for lateral pressure estimates for design

See Mayne et al. (2001) for methods used to evaluate subsurface water levels. The results of the subsurface investigation program should consider the following for earth retaining systems: 1. Special groundwater conditions such as the presence of artesian pressure or perched water can have a critical effect on the construction of a cut wall system. Experience with soil running up into a borehole during withdrawal of drilling tools, loss of drilling fluid, or sloughing of soil from walls of the borehole should be recorded. 2. In boring operations, any evidence of extremely compact materials, cobbles, boulders, cemented layers or lenses, “hardpan”, rubble fill, debris or other material that would impede the installation of a vertical wall element such as a sheetpile or soldier beam should be recorded. Such information should be recorded on boring logs to be included in the contract information for bidders. 3. At a site where rock or rock-like material are present, the borings must define the character of the materials where there is a transition between residual soil and bedrock, and should locate the surface of materials of rock-like hardness. Many engineers underestimate the importance in selecting an appropriate investigation technique(s) and sampling method(s) since in-house SPT capabilities are a presumptive “one size fits all” investigation/sampling approach. Experience has shown the short-falls of this


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approach (see Mayne et. al, 2001) and the benefit of a focused site-specific strategy which may include in-situ testing methods. Equipment for in-situ testing includes cone penetrometers, vane shear devices, and pressuremeters. These devices can provide index parameters or strength parameters. For example, the vane shear test is often used for evaluating the undrained shear strength of clayey soil deposits, especially those for which undisturbed sampling for laboratory testing cannot be performed in a cost-effective manner (e.g., for weak clays). The test is effective in developing a profile of undrained strength with depth and in evaluating the sensitivity (i.e., strength loss upon loading) of the soil. A summary of available in-situ testing devices is provided in Table 2-3. Table 2-3. In-situ Testing Methods Used in Soil. Method

Procedure

Electric Cone Penetrometer (CPT)

A cylindrical probe is hydraulically pushed vertically through the soil measuring the resistance at the conical tip of the probe and along the steel shaft; measurements typically recorded at 1 to 2 in. intervals Same as CPT; additionally, penetration porewater pressures are measured using a transducer and porous filter element

Piezocone Penetrometer (CPTu)

Seismic CPTu (SCPTu)

Flat Plate Dilatometer (DMT)

Same as CPTu; additionally, shear waves generated at the surface are recorded by a geophone at 3 ft intervals throughout the profile for calculation of shear wave velocity A flat plate is hydraulically pushed or driven through the soil to a desired depth; at approximately 8 to 12 in. intervals, the pressure required to expand a thin membrane is recorded; Two to three measurements are typically recorded at each depth


Applicable Soil Types Silts, sands, clays, and peat

Applicable Soil Properties Estimation of soil type and detailed stratigraphy Sand: φ′, Dr, σho′ Clay: su, σp′

Silts, sands, clays, and peat

Same as CPT, with additionally: Sand: uo / water table elevation Clay: σp′, ch, kh OCR


Same as CPTu, with additionally: Vs, Gmax, Emax, ρtot, eo


Estimation of soil type and stratigraphy Total unit weight Sand: φ′, E, Dr, mv Clays: σp′, Ko, su, mv, E, ch, kh

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Limitations / Remarks No soil sample is obtained; The probe may become damaged if testing in gravelly soils is attempted; Test results not particularly good for estimating deformation characteristics If the filter element and ports are not completely saturated, the pore pressure response may be misleading; Compression and wear of a mid-face (u1) element will effect readings; Test results not particularly good for estimating deformation characteristics First arrival times should be used for calculation of shear wave velocity; If first crossover times are used, the error in shear wave velocity will increase with depth Membranes may become deformed if overinflated; Deformed membranes will not provide accurate readings; Leaks in tubing or connections will lead to high readings; Good test for estimating deformation characteristics at small strains


Method

Procedure

Pre-bored Pressuremeter (PMT)

A borehole is drilled and the bottom is carefully prepared for insertion of the equipment; The pressure required to expand the cylindrical membrane to a certain volume or radial strain is recorded A cylindrical probe with a pressuremeter attached behind a conical tip is hydraulically pushed through the soil and paused at select intervals for testing; The pressure required to expand the cylindrical membrane to a certain volume or radial strain is recorded A 4 blade vane is slowly rotated while the torque required to rotate the vane is recorded for calculation of peak undrained shear strength; The vane is rapidly rotated for 10 turns, and the torque required to fail the soil is recorded for calculation of remolded undrained shear strength

Full Displacement Pressuremeter (PMT)

Vane Shear Test (VST)

Symbols used in Table 2-3. φ′: effective stress friction angle D r: relative density σho′: in-situ horizontal effective stress su: undrained shear strength preconsolidation stress σp′: ch: horizontal coefficient of consolidation kh: horizontal hydraulic conductivity OCR: overconsolidation ratio Vs: shear wave velocity

2.2.4

Applicable Soil Types Clays, silts, and peat; marginal response in some sands and gravels

Applicable Soil Properties E, G, mv, su

Clays, silts, and peat in sands

E, G, mv, su

Disturbance during advancement of the probe will lead to stiffer initial modulus and mask liftoff pressure (po); Good test for calculation of lateral deformation characteristics

Clays, Some silts and peats if undrained conditions can be assumed; not for use in granular soils

su, St, σp′

Disturbance may occur in soft sensitive clays, reducing measured shear strength; Partial drainage may occur in fissured clays and silty materials, leading to errors in calculated strength; Rod friction needs to be accounted for in calculation of strength; Vane diameter and torque wrench capacity need to be properly sized for adequate measurements in various clay deposits

Gmax: G: Emax: E: ρtot: eo: mv: Ko: St:

Limitations / Remarks Preparation of the borehole most important step to obtain good results; Good test for calculation of lateral deformation characteristics

small-strain shear modulus shear modulus small-strain Young’s modulus Young’s modulus total density in-situ void ratio volumetric compressibility coefficient coefficient of at-rest earth pressure sensitivity

Sampling Methods

2.2.4.1 Disturbed Sampling of Soil Disturbed sampling provides a means to evaluate stratigraphy by visual examination and to obtain soil specimens for laboratory index testing. Disturbed samples are usually collected using split-barrel samplers (AASHTO T206, ASTM D 1586), although several other techniques are available for disturbed sample collection in boreholes. Typical split-barrel FHWA NHI-07-071 Earth Retaining Structures

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samplers are limited soil grain size which allows for driving (i.e., typically not appropriate for gravels). Shallow disturbed samples can also be obtained using hand augers and test pits. Samples obtained via disturbed sampling methods can often be used for index property testing in the laboratory but explicitly should not be used to prepare specimens for consolidation and strength (i.e. performance) tests. 2.2.4.2 Undisturbed Sampling of Soil Undisturbed soil samples are required for performing laboratory strength and consolidation testing on generally cohesive soils ranging from soft to stiff consistency. High-quality samples for such testing are particularly important for fill wall systems that may stress compressible strata. In reality, it is impossible to collect truly undisturbed samples since changes in the state of stress in the sample will occur upon sampling. The goal of highquality undisturbed sampling is to minimize the potential for: (1) alteration of the soil structure; (2) changes in moisture content or void ratio; and (3) changes in chemical composition of the soil. Due to cost and ease of use, the thin-walled Shelby tube is the most common equipment for obtaining relatively undisturbed samples of soils. Depending upon cohesive soil type (e.g., stiffness and whether significant granular material is in the soil matrix), alternative sampling equipment may be used to obtain nominally undisturbed soil samples including: (1) stationary piston; (2) hydraulic piston; (3) Denison; and (4) pitcher samplers. Detailed procedures for these sampling techniques are provided in FHWA NHI01-031 (Mayne et al., 2001). 2.2.4.3 Rock Coring When considering equipment for rock coring, the dimensions, type of core barrel, type of coring bit, and drilling fluid are important variables. The minimum depth of rock coring should be determined based on the local geology of the site. Coring should also be performed to a depth that assures that refusal was not encountered on a boulder. Four different types of core barrels are described in ASTM D 2113 including: (1) Single Tube; (2) Rigid Double Tube; (3) Swivel Double Tube; and (4) Triple Tube. A brief description of issues related to rock coring is provided subsequently. Additional information on drilling rigs, methods of circulating drill cuttings (i.e., fluid or air), hole diameters, and casings are provided in ASTM D 2113. Since the double core barrel isolates the rock from the drilling fluid stream to yield better recovery, it is the minimum standard of core barrel that should be used in practice when an intact core is required. The inner tube of a swivel-type core barrel does not rotate during FHWA NHI-07-071 Earth Retaining Structures

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drilling, resulting in less disturbance and better recovery in weak and fractured rock. Rigid type double tube core barrels should not be used where core recovery is a concern. Triple tube swivel-type core barrels will produce better recovery and less core breakage than a double tube barrel. Before using correlations for rock parameters, the engineer should account for the specific rock sampling procedure (e.g., double tube or triple tube) used as the basis for the correlation. The standard size rock core is NX 2 1/8 in. diameter. Generally larger core sizes will lead to less mechanical breakage and yield greater recovery, but the associated cost for drilling will be much higher. Since the size of the core will affect the percent recovery, this should be clearly recorded on the log. Additionally, the core length can increase recovery in fractured and weathered rock zones. In these zones a core length of 5 ft is recommended, and core lengths should not be greater than 10 ft under any conditions because of the potential to damage the long cores. Table 2-4 provides summary sampling guidelines for geotechnical investigations. 2.2.5

Developing a Laboratory Testing Program

The final planning step includes the development of a laboratory-testing program. Table 2-1 lists laboratory tests that are used to evaluate design parameters for cut and fill walls. Index tests such as moisture content, Atterberg limits, grain size distribution, and unit weight tests are recommended for both cut and fill wall applications. Index tests are invaluable in establishing general conditions and assessing inherent material variability. Depending on individual site conditions, additional tests may be required or tests may be eliminated from Table 2-1. Once a list of necessary tests has been developed and the field program has been executed, the engineer should review field notes, borings, and design plans to identify “critical areas”. Critical areas correspond to borings/locations where the results of the laboratory tests could result in a significant change in the proposed design. Samples from these critical areas should be identified for performance testing. In heterogenous areas, many samples may be required to obtain comprehensive parameters; in homogeneous areas, few samples may be required. A laboratory-testing program should be performed on representative and critical specimens from geologic layers across the site. To assess the locations where tests should be performed, it is useful to evaluate sample location maps and geologic cross sections. By evaluating a sample location map, it will be easy to quickly identify the locations of disturbed and undisturbed samples that may be used in the laboratory testing program. The generation


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Table 2-4. Sampling Guidelines (after WFLHD, 2004). Sand-Gravel Soils • SPT1 (split-spoon) samples should be taken at 5 ft intervals and at significant changes in soil strata. • Continuous SPT samples are recommended in the top 15 ft of borings made at locations where spread footings may be placed in natural soils. • Representative SPT jar or bag samples should be laboratory classified for verification of field visual soil identification. Silt-Clay Soils • SPT and “undisturbed” thin wall tube samples2 should be taken at 5 ft intervals and at significant changes in strata. • SPT and tube samples may be alternated in same boring or tube samples may be taken in separate undisturbed boring. • Representative SPT jar or bag samples should be laboratory classified for verification of field visual soil identification. • Tube samples should be tested for consolidation (for settlement analysis) and strength (for slope stability and foundation bearing resistance analysis). • Field vane shear testing also recommended to obtain in-place shear strength of soft clays, silts, and nonfibrous peats. Rock • Continuous cores should be obtained in rock or shales using double or triple tube core barrels. • For foundation investigations, core a minimum of 10 ft into the rock. • Core samples should be evaluated for strength testing (unconfined compression) for foundation investigations, and valued for quality tests for quarry investigations (aggregate or riprap). • Determine percent core recovery and RQD3 value for each core run and record in the bore log. Groundwater • Record water level encountered during drilling, at completion of boring, and at 24 hours after completion of boring in the bore log. • When water is used for drilling fluid, adequate time should be permitted after hole completion for the water level to stabilize (more than one week may be required). In impermeable soils, a plastic pipe water observation well should be installed to allow monitoring of the water level over a period of time. • Artesian pressure and seepage zones, if encountered, should also be noted in the bore log. • The top 12 in or so for the annular space between water observations well pipes and the borehole wall should be backfilled with grout, bentonite, or a sand-cement mixture to prevent surface water inflow which can cause erroneous groundwater level readings. Notes:

1) Standard Penetration Test (AASHTO T206); 2) Thin-Walled Tube Sampling of Soils (AASHTO T207); and 3) Rock Quality Designation.


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of detailed subsurface cross sections along wall alignment, including stratigraphy, in-situ testing results, and laboratory index test results, if available, will be useful when identifying representative samples for laboratory performance testing. Evaluation of soil properties form laboratory testing is described subsequently in this chapter.

2.3

FUNDAMENTAL CONCEPTS OF SHEAR STRENGTH

2.3.1

General

As part of the design analyses for most geotechnical projects, the load-carrying capacity of the supporting soil is evaluated. For earth retaining systems, these capacity evaluations include, for example, bearing resistance of a semi-gravity cantilever wall supported on a spread footing or deep foundation, side and tip resistance of drilled-in soldier beams or driven soldier beams for an anchored wall, stability analyses of slopes being restrained by a wall, and passive soil resistance for the toe of a retaining structure. These analyses are concerned with comparing loads imposed to the system to the limit (or failure) state of the supporting soil. This limit state of stress corresponds to the shear strength of the soil. For a given soil, shear strength is not a unique property. The soil strength to be used in design analyses must be qualified in relation to whether the appropriate strength is: drained or undrained, peak, fully softened or residual, intact or remolded, static or cyclic, compression or extension; and other facets, such as direction of loading, rate of loading, and boundary conditions. As a consequence, soil strength is not a fundamental property, but instead, a specific behavioral response to a certain set of loading conditions. Fundamental information relating to shear strength evaluation including concepts of drained and undrained behavior, and total and effective stress analyses, respectively, are presented in this section as they relate to shear strength evaluation for earth retaining systems. The numerous factors that may affect the selection of shear strength values for design analyses are presented. Additional discussion on specific methods used to evaluate soil strength for wall systems is described in subsequent sections. 2.3.2

Drained versus Undrained Loading

In geotechnical practice, it is important to distinguish between "drained" and "undrained" strengths. These terms refer to the ability of the porewater in the soil to move between soil particles resulting in volume change, and the accompanying generation (or lack) of excess FHWA NHI-07-071 Earth Retaining Structures

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porewater pressures, ∆u. Soils can also exhibit any number of partially drained strengths, however, wall design analyses are typically performed using drained and undrained strengths, as these represent limits to the expected range of behavior. For a saturated soil subjected to undrained loading, no drainage of porewater from the void spaces can occur, and thus the soil undergoes no change in volume. During undrained loading, changes in total stress (∆σ) cause the development of either positive porewater pressures (∆u > 0) that will tend to decrease the effective stress in the soil or negative porewater pressures (∆u < 0) that will tend to increase the effective stress in the soil. The drained loading of a saturated soil means that the water in the void spaces is free to move so that no excess pore water pressures develop (∆u = 0). There is usually a change (i.e., increase or decrease) in void ratio and a corresponding change in volume. Again, water may be present, but is free to move either out of the soil mass (termed contractive soil behavior) or into the soil mass (termed dilatant soil behavior). Contractive behavior results in a decrease in volume (e.g., settlement) and dilative behavior causes an increase in volume (e.g., swelling). Most sands have such a high permeability (e.g., k > 10-3 cm/s (2.83 ft/day)) that, under static loading, they are almost always drained. Sands, however, will behave in an undrained mode when subjected to rapid loading, such as that imposed by an earthquake whereby the entire deposit is engaged and water cannot drain. All clays exhibit drained behavior when the rate of loading is very slow, so slow that it does not interfere with the rate of water migration that is controlled by its low permeability (e.g., k < 10-6 cm/s (0.00283 ft/day)). Drained behavior in clays should be considered in evaluating the long-term stability (and ground movements) of cut wall systems. The same clay that behaves in a drained manner in these applications, however, may initially behave in an undrained mode in the short-term if the rate of loading is too fast to permit water inflow. The short-term stability of excavations constructed in soft to medium clays is represented by undrained loading conditions. 2.3.3

Drained Stress-Strain-Strength Behavior

In this section, the stress-strain-strength behavior of soils is introduced for the simple case of drained loading. To illustrate this, a graph of measured shear stress (τ) versus shear strain (γs) from a direct simple shear test is used (Figure 2-2). It is recognized that the direct simple shear test is not commonly performed, but it is most useful in introducing drained stressstrain strength behavior of soils. The more common direct shear (box) test is very similar to the direct simple shear and thus the same basic principles apply.


2-14


The maximum stress on the τ-γs curve is commonly interpreted as the peak shear strength (τmax), corresponding to point 2 in Figure 2-2. Figure 2-2 shows two stress-strain curves, each one corresponding to different effective consolidation stresses (i.e., σv1′ and σv2′). For each specimen, the measured peak strength (τmax) is plotted versus effective consolidation stresses as shown in Figure 2-2. A linear fit is generally forced (minimum of two data sets) to provide the simplified straight line Mohr-Coulomb failure criterion: τmax = c′ + σ′ tan φ′

(2-1)

where φ′ = effective stress friction angle and c′ = effective cohesion intercept. During drained shearing, the soil specimen will likely undergo a change in total volume. If the soil decreases in volume during shear, the response is termed contractive behavior. This response is indicative of loose sands and soft clays. If the soil increases in volume during shearing, a dilative (or dilatant) behavior is observed. This response is common in hard clays and dense sands. If no change in volume occurs during drained shear loading (∆V/V0 = 0), the corresponding stress state is called the critical state. The complete description of soil behavior by this arrangement is termed critical-state soil mechanics, which encompasses normally- and overconsolidated soils for undrained, semi-drained, and drained loading, for both contractive and dilatant behavior (Schofield and Wroth, 1968; Wood, 1990). However, in current U.S. practice, the concept of critical state soil mechanics is not consistently recognized, although this concept is an excellent representation of soil behavior. The adoption of the simplified Mohr-Coulomb strength criterion is widespread in current U.S. practice and it is normally taken that “drained conditions” are analyzed using effective stress parameters (i.e., φ′, c′) and “undrained conditions” are analyzed by total stress methods (i.e., c=cu=su). For cohesive soils, the drained analysis corresponds to long-term conditions and undrained analyses to short-term conditions. For cohesionless soils (i.e., those with relatively high hydraulic conductivity values), drained analyses alone are performed. The Mohr-Coulomb strength parameters, φ′ and c′, are defined in Figure 2-3. Referring to drained strength characteristics in Figure 2-2, after the peak shear strength (τmax) is reached during drained loading, the shear stress reduces to a stable value termed the fullysoftened strength, depicted as point 3 in Figure 2-2. The fully softened strength is intermediate between the peak strength and the residual strength and there are no specific procedures to identify the fully softened strength. Conceptually, the fully softened strength is close in value to the peak strength of the same soil in a normally-consolidated condition.


2-15


σv' =

σv'

effective overburden stress

τ= Shear Stress

− δv, contractive soil

+ δ v, dilative soil

τ

τmax

peak Post-peak Strain-softening

τ = c' + σ' tan φp' Shear Stress, τ

Shear Stress, τ

τmax

2 peak

τmax Gtangent =

σv2'

∆τ ∆γ s

Fully-softened Strength

3

1

Gsecant =

τ γs

σv1'

Go = small-strain shear modulus

τmax 2 tan φ'peak

τ = σ' tan φ'

residual

3

4 effective consolidation stress

4

residual

τr = σ' tan φr'

tan φ'r (residual strength)

c'

Shear Strain, γs

tan φ'NC (fully-softened)

1

Normal Stress, σv' σv1′

σv2′

Figure 2-2. Drained Stress-Strain Behavior.

Figure 2-3. Mohr-Coulomb Failure Criteria. FHWA NHI-07-071 Earth Retaining Structures

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This can be expressed as: τ = σ′ tan φNC′

(2-2)

where φNC′ is the peak strength (or critical-state strength) of the normally-consolidated soil. Note that for long-term analyses, the effective cohesion intercept is a small value for the normally consolidated case and is therefore assumed to be zero (c′ = 0). For clays, if drained loading continues for very large shearing strains, the shear stress value drops even further to the residual strength, denoted τr and indicated by point 4. The residual strength is related to the mineralogical frictional characteristics of the soil in which the platelike clay particles align themselves in a direction parallel to the shear plane that is developed at these very large strains. The residual strength can be represented by: τr = σ′ tan φr′

(2-3)

In commercial practice, φr′ is obtained using 8 to 10 repeated cycles of shearing on the same specimen in a direct shear box using the same direction of shear and the same normal load. The more elaborate ring shear device is purposely suited for obtaining true residual values of φr′. Information on the ring shear-testing device can be found in Terzaghi et al. (1996). For design of walls constructed on or in clay soils that exhibit peak, fully softened, and residual shear strength, the design engineer must consider the level of deformations that may be expected within the soil mass to appropriately select the strength to be used for design calculations. The relationship between deformations and appropriate shear strength to be used for design is discussed in Section 3.5. 2.3.4

Undrained Stress-Strain-Strength Behavior

The undrained shear stress-shear strain curve is similar to that observed for drained loading, except that excess porewater pressures are also generated (∆u ≠ 0). During undrained shearing, a contractive soil will exhibit positive pore pressures, while a dilative soil will show negative pore pressures. From the undrained τ-γs curve, the peak value of τmax is designated as the undrained shear strength (su or cu). Instability under undrained conditions develops mainly for a contractive soil where the soil attempts to mobilize frictional shearing resistance which also causes the soil to contract under the prevailing confining stresses. This tendency to contract during shear is typical for normally to lightly overconsolidated soft to medium clay soils. Since this tendency cannot be realized, due to the clay soil permeability in relation to the rate of shearing, positive FHWA NHI-07-071 Earth Retaining Structures

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porewater pressures are generated in the soil which reduce the effective stress and hence the mobilized frictional shearing resistance. In such cases the short term undrained shearing resistance of the soil is less than would have been the case if drainage (contraction of the soil volume) could have occurred. The short-term condition is critical for temporary walls constructed in normally to lightly overconsolidated clay soils. In clay soils subjected to unloading conditions that may result from an excavation in front of a wall, the soil attempts to expand as it mobilizes frictional shearing resistance. This is resisted causing negative porewater pressure to be developed that increases the effective stress in the soil and hence increases the mobilized frictional shearing resistance. Thus, in overconsolidated clay subject to excavation, the short-term (undrained) strength and stability potentially exceeds that which would apply once drainage has occurred. For the examples cited, the engineer needs to assess both short-term and long-term strengths in their analyses. For dilative soils, the tendency to dissipate pore pressure will reduce the effective stress and thus the strength. 2.3.5

Effective Stress Parameters

Granular soils such as gravels, sand, and non-plastic silts have effective stress failure envelopes that pass through the origin indicating that c′ = 0 for these materials. In fact, for granular soils, only cemented sands appear to have a true c' value (i.e., c′ ≠ 0). The value of φ′ for sands depends on mineralogy and packing arrangement that is related to relative density and effective confining stress level (Bolton, 1986). Ranges of φ′ for clean quartzitic (silica) sands are typically 30° ≤ φ′ ≤ 50°, whereas calcareous (corraline) sands may exhibit somewhat higher values. 2.3.6

Total Stress Parameters

Total stress analyses for soils that do not drain during the loading period involve the principle that if an element of soil in the laboratory is subjected to the same changes in total stress under undrained conditions as an element of the same soil in the field, the same excess pore pressures will develop. Thus, if the total stress in the laboratory and the field are the same, the effective stresses will also be the same. Because soil strength is governed by effective stresses, the strength measured in laboratory tests should be the same as the strength in the field when the pore pressures and total stresses are the same. Thus, under undrained conditions, strengths can be related to total stresses, making it unnecessary to specify undrained excess pore pressures for design analyses.


2-18


Although the total stress principle is simple, experience has shown that many factors influence the pore pressures that develop under undrained loading. As a result, determining undrained strengths by means of laboratory and in-situ testing requires considerable attention to detail if reliable results are to be achieved. Shear strengths for use in undrained total stress analyses must be measured using test specimens and loading conditions that closely duplicate the conditions in the field. Alternatively, they can be reliably measured using the appropriate in-situ test. Total stress type analyses using the “φ = 0” approach are perhaps the most widely used form of analyses performed by highway engineers for design analyses involving clays. For this condition, the Mohr-Coulomb relationship reduces to the form σ1′ = σ3′ + 2c or c = cu = su = ½ (σ1′-σ3′) = undrained shear strength. Since su is stress dependent, its value is commonly normalized by the vertical effective overburden stress (σvo′) at the depth where su is measured.

2.4

LABORATORY INDEX TESTS

2.4.1

General

In this section, typical laboratory index tests performed in support of the design of an earth retaining system are presented. Index soil properties used in the analysis and design of earth retaining systems include unit weight, moisture content, gradation, and Atterberg Limits. Unit weight of foundation material, backfill soil, and retained soil are used in evaluating earth pressures and in evaluating the external and internal stability of the wall system. Moisture content and Atterberg Limits are used with engineering correlations to estimate compressibility and shear strength of clayey soils. These data also provide an indication of soil creep potential which is important for the feasibility of soil anchors and nails. The results of grain size distribution testing provide an indication of soil permeability and compaction characteristics. Gradation information is also used to develop appropriate drilling and grouting procedures and in the design of dewatering systems. 2.4.2

Moisture Content

Moisture content is defined as the ratio of the mass of the water in a soil specimen to the dry mass of the specimen. Natural moisture contents (wn) of sands are typically 0 ≤ wn ≤ 20 %, whereas for inorganic and insensitive silts and clays, general ranges are: 10 ≤ wn ≤ 40 %. However, depending upon the mineralogy, formation environment, and structure of clay, it is FHWA NHI-07-071 Earth Retaining Structures

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possible to have more water than solids (i.e., w > 100%). Therefore soft and highly compressible clays, as well as sensitive, quick, or organically rich clays, can exhibit water contents 40 ≤ wn ≤ 300 % or more. Moisture content can be tested in a number of different ways including: (1) a drying oven (ASTM D 2216); (2) a microwave oven (ASTM D 4643); or (3) a field stove or blowtorch (ASTM D 4959). While the microwave or field stove (or blowtorch) methods provide a rapid evaluation of moisture content, potential errors inherent with these methods require confirmation of results using ASTM D 2216. The radiation heating induced by the microwave oven and the excessive temperature induced by the field stove may release water entrapped in the soil structure that would normally not be released at 110o C, yielding higher moisture content values than would occur from ASTM D 2216. 2.4.3

Unit Weight

In the laboratory, soil unit weight and mass density are easily measured on tube samples of natural soils. The moist (total) mass density is ρt = Mt/Vt, whereas the dry mass density is given by ρd = Ms/Vt. The moist (total) unit weight is γt = Wt/Vt, whereas the dry unit weight is defined as γd = Ws/Vt. The interrelationship between the total and dry mass density and unit weight is given by: ρd =

ρt (1 + w n )

(2-4)

and the relationship between total and dry unit weight is given by: γd =

γt (1 + w n )

(2-5)

Unit weight evaluation for backfill soils is described in Section 2.5.3. 2.4.4

Atterberg Limits

The Atterberg limits of a fine grained (i.e., clayey or silty) soil represent the moisture content at which the behavior of the soil changes. Atterberg limits results provide an indication of several physical properties of the soil, including strength, permeability, compressibility, and shrink/swell potential. These limits also provide a relative indication of the plasticity of the soil, where plasticity refers to the ability of a silt or clay to retain water without changing state from a semi-solid to a viscous liquid. In geotechnical engineering practice, the FHWA NHI-07-071 Earth Retaining Structures

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Atterberg limits generally refers to the liquid limit (LL), plastic limit (PL), and shrinkage limit (SL); although the SL is less often used in typical geotechnical practice. A measure of a soils plasticity is the plasticity index (PI) which as calculated as PI = LL – PL. The PI is a useful index since numerous engineering correlations have been developed relating PI to clay soil properties, including undrained and drained strength and compression index. Other indices based on the Atterberg limits include the liquidity index (LI) and the activity (A) of a soil. These are defined as: LI =

(w n − PL) PI

A=

PI CF

(2-6)

(2-7)

where wn is the moisture content of the soil and CF is the clay fraction that corresponds to the percentage of particles exhibiting an equivalent diameter (ds) < 0.078 mil (0.002 mm). The use of the liquidity index and activity can provide very useful information concerning the likely behavior of a soil, even though Atterberg limits are performed on completely remolded materials. For example, a LI less than or equal to zero is generally indicative of a heavily overconsolidated soil that may be desiccated or highly expansive. A soil with a LI equal to 1.0 implies that the soil is at its liquid limit and is likely to be relatively weak and compressible. A LI greater than unity indicates that the soil is sensitive. Soils with a LI greater than approximately 0.7 will likely undergo significant consolidation settlements when loaded. Clayey soil backfill should have a liquidity index close to zero or negative. 2.4.5 Particle Size Distribution

Particle size distribution by mechanical sieve and hydrometer are useful for soil classification purposes. Procedures for grain size analyses are contained in ASTM D 422 and AASHTO T88. Testing is accomplished by placing air-dried material on a series of screens of known opening size. Each successive screen has a smaller opening to capture progressively smaller particles. Testing of the finer grained particles is accomplished by suspending the chemically dispersed particles in water column and measuring the specific gravity of the liquid as the particles fall from suspension. Representative samples with fines (particles with diameter less than U.S. No. 200 sieve) should not be oven dried prior to testing because some particles may cement together leading FHWA NHI-07-071 Earth Retaining Structures

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to a calculated lower fines content from mechanical sieve analyses than is actually present. When fine-grained particles are a concern, a wash sieve (ASTM D 1140) should be performed to assess the fines content. Additionally, if the clay content is an important parameter, hydrometer analyses need to be performed. It should be noted that the hydrometer test provides approximate analysis results due to oversimplified assumptions, but the obtained results can be used as a general index of silt and clay content. 2.4.6 Laboratory Classification

In addition to field identification (ASTM D 2488), soils should be classified in the laboratory using the Unified Soil Classification System (USCS) in accordance with ASTM D 2487 or the AASHTO soil classification system (AASHTO T 145). Classification in the laboratory occurs in a controlled environment and more time can be spent on this classification than the identification exercise performed in the field. Laboratory and/or field identification is important so that defects and features of the soil can be recorded that would not typically be noticed from index testing or standard classification. Some of the features include mica content, joints, and fractures. 2.4.7

Specific Gravity

The specific gravity of solids (Gs) is a measure of solid particle density and is referenced to an equivalent volume of water. Specific gravity of solids is defined as Gs = Ms/(Vs × γw) where Ms is the mass of the soil solids and Vs is the volume of the soil solids. Since many sands are comprised of quartz and/or feldspar minerals and many clays consist of the kaolinite and/or illite clay minerals in composition, and since the specific gravity of these minerals are confined to a relatively narrow range, the typical values of specific gravity of most soils also lie within the narrow range of Gs = 2.7 ± 0.1. Exceptions include soils with appreciable organics (i.e., peat), ores (mine tailings), or calcareous (high calcium carbonate content) constituents. It is common to assume a reasonable Gs value, although laboratory testing by AASHTO T100 or ASTM D 854 or D 5550 can be used to verify and confirm its magnitude, particularly on projects where little previous experience exists and unusually low or high unit weights are measured. 2.4.8

Organic Content

A visual assessment of organic materials may be very misleading in terms of engineering analysis. Laboratory test method AASHTO T194 or ASTM D 2974 should be used to evaluate the percentage of organic material in a specimen where the presence of organic material is suspected based on field information or from previous experience at a site. The test involves heating a sample to temperatures of 440°C and holding this temperature until no FHWA NHI-07-071 Earth Retaining Structures

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further change in mass occurs. At this temperature, the sample turns to ash. Therefore, the percentage of organic matter is (100% - % ash) where the % ash is the ratio of the weight of the ash to the weight of the original dried sample. The sample used for the test is a previously dried sample from a moisture content evaluation. Usually organic soils can be distinguished from inorganic soils by their characteristic odor and their dark gray to black color. In doubtful cases, the liquid limit should be determined for an oven-dried sample (i.e., dry preparation method) and for a sample that is not pre-dried before testing (i.e., wet preparation method). If drying decreases the value of the liquid limit by about 30 percent or more, the soil may usually be classified as organic (Terzaghi et al., 1996). Soils with relatively high organic content have the ability to retain water, resulting in high moisture content, high primary and secondary compressibility, and potentially high corrosion potential. Organic soils may or may not be relatively weak depending on the nature of the organic material. Highly organic fibrous peats can exhibit high strengths despite having a very high compressibility.

2.5

PARAMETERS FOR BACKFILL SOILS

2.5.1

General

In this document, backfill soils refer to those soils which are transported to the project site, spread, and compacted just behind a fill wall. Backfill soils may be obtained from an Ownerprovided on-site borrow source or are transported to the site from an off-site location. In general, relatively free-draining soils are usually required for backfill soils for fill walls, although more fine-grained materials may be used subject to specific design considerations (discussed subsequently). The primary backfill soil parameters required for design analyses including unit weight, shear strength, electrochemical properties, and drainage characteristics are described in this section. Information on alternative backfill materials (i.e., lightweight fills and flowable fill) and evaluation of soil swelling potential and degradation for specific backfill types is also presented in this section. 2.5.2

Classification of Backfill Soils

Cohesionless Soils

Soils classified as GW, GP, SW or SP in accordance with the Unified Soil Classification System (USCS) are excellent backfill materials. The important characteristics of these soils are their high frictional resistance and high permeability. With adequate drainage measures FHWA NHI-07-071 Earth Retaining Structures

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(such as weep holes, collector drains, etc.) the build-up of pore water pressure in the backfill can be prevented. Sandy Clays and Clayey Sands

Soils classified as SC, SM, GC or GM, according to the USCS, may be suitable as wall backfill if kept dry, but are subject to frost action when wet. Such soils have low permeability and do not drain rapidly. Hence, their water content may increase substantially as a result of precipitation or flooding. Also, these soils cannot be properly compacted when wet. Silts and Clays

Soils classified as CL, CH, MH, ML or OL, according to the USCS, are often subject to excessive frost action and swelling when used as wall backfill, and the resulting wall movement is likely to be excessive. Moreover, if used behind relatively rigid walls such as concrete gravity walls, lateral wall pressures can become quite large. Accordingly, the use of these materials generally should be avoided for such a wall system. Earth pressures associated with clayey soil for wall backfill is discussed in Chapter 3. 2.5.3

Unit Weight

The unit weight of backfill is used in estimating the vertical stress in the soil which, in turn, is utilized in calculating the lateral earth pressure applied to the wall. For preliminary analyses, dry unit weight (γd) of the backfill soil can be estimated using Table 2-5. Laboratory compaction test results based on ASTM D 698 for Standard Proctor Compaction Effort and ASTM D 1557 for Modified Proctor Compaction Effort are performed as part of final design to evaluate total and dry unit weight values. These tests are used to establish target compaction requirements for the backfill soil and are used to develop appropriate moisture-density relationships for field control of backfill compaction. Field measurements of soil mass density (unit weight) are generally restricted to shallow surface samples, usually when placing compacted fills, and can be accomplished using drive tubes (ASTM D 2937), sand cone method (ASTM D 1556), or nuclear gauge (ASTM D 2922). These field measurements of dry unit weight represent a quality assurance activity, which is performed to confirm adequate shear strength and stiffness of wall backfill materials.


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2.5.4

Backfill Shear Strength Parameters

2.5.4.1 General Backfill soil shear strength parameters are required for lateral earth pressure evaluations for fill walls. Typically, drained shear strength parameters are used for relatively free-draining granular backfill soils and for evaluation of long-term conditions for clayey backfill soils; and undrained shear strength parameters are used for analyses of short-term (i.e., temporary) loading conditions involving clayey backfills. 2.5.4.2 Drained Strength Parameters Typical (or default) drained shear strength parameters of backfill soils can be obtained from Table 2-5 or alternatively from Figure 2-4 or Figure 2-5. Figure 2-4 shows typical ranges of friction angle for rockfills, gravels, and sands over a wide range of confining stresses and with initial porosities ranging from 0.17 to 0.48. To use this figure, the engineer must select the range of confining stresses that the granular soil will be subject to in the field. For wall backfill, this corresponds to the range of vertical stresses within the backfill (including stresses due to surcharge loadings). If this range is relatively large, the friction angle of the soil will vary over the range of confining stresses. A conservative single value can be selected based on calculating confining stresses at the bottom, middle, and top of the backfill soil and then averaging. The rockfill grades for quarried materials shown on Figure 2-4, are summarized in Table 2-6 (A through E). It is noted, however, that where project-specific testing is not performed, most state agencies will provide conservative backfill strength properties for walls. These conservative properties are suitable for inclusion in standard specifications or special provisions when project specific testing is not feasible. Project-specific drained shear strength parameters can be obtained from direct shear testing (ASTM D 3080). The test is performed on recompacted soils with normal stress range consistent with the anticipated range of stresses for the field application. Tests should be performed at varying compaction and density conditions to be representative of the anticipated compaction conditions that will be achieved in the field. Consideration should be given to the potential for the soils to become saturated after construction and, if saturation is possible, tests should be performed on saturated (or submerged) samples. Tests should always be carried out until a stable large-displacement shear stress is measured. 2.5.4.3 Undrained Strength Parameters Walls designed with clayey backfills need to consider both drained and undrained shear strength parameters for evaluation of design earth pressures. Herein, methods to evaluate FHWA NHI-07-071 Earth Retaining Structures

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undrained strength parameters for recompacted clayey soils used for backfills are discussed. Undrained strength parameters for clayey backfill soils are typically determined using direct shear testing or consolidated undrained triaxial compression testing. For these tests, the soil should be compacted in the laboratory to the maximum moisture content anticipated in the field. This moisture content should at least be to the maximum allowed by the Specifications, for example, up to 2 points wet of optimum moisture or to some worst-case value prescribed by the Engineer. Typically, compacted clayey backfills will be unsaturated under field placement and compaction conditions, however, it is strongly suggested that laboratory testing be performed on saturated recompacted samples to provide “worst-case” undrained strength parameters. An example consolidated undrained triaxial compression test with pore pressure measurements for a lean clay soil used for a welded wire-faced MSE wall is provided in Figure 2-6. 2.5.5

Electrochemical Parameters of Backfill Soils

The design of buried steel elements such as that for metallic reinforcements used for MSE walls is predicated on the backfill soils exhibiting minimum or maximum electrochemical properties and then designing the structure for maximum corrosion rates associated with these properties. For corrosion potential evaluation, backfill soils are tested for: (1) pH (AASHTO T-289); (2) electrical resistivity (AASHTO T-288); (3) sulfate content (AASHTO T-290); and (4) chloride content (AASHTO T-291). Some soils have tendency to be more aggressive then others. For example, clayey and silty soils are generally more aggressive than granular soils because their fine-grained nature results in high water holding capacity, poor aeration, and poor drainage. These characteristics tend to promote corrosion. Additional discussion on soil corrosion potential evaluation is provided in subsequent chapters for individual wall types. 2.5.6

Placement and Compaction of Backfill Soils

Regardless of backfill type, compaction is required to obtain increased backfill shear strength and stiffness. Compaction also minimizes backfill settlement that may occur during and after construction. Compaction, however, may induce large lateral stresses against the wall, particularly near the top of the wall (see Section 3.8). Backfill soils should be placed in loose lift layers of no more than 12 in. thick, and should be properly compacted by rollers and/or tampers. A dry unit weight not less than 95 percent of the maximum dry density achieved in a Standard Proctor test (AASHTO T99) is commonly specified, although higher degrees of compaction may be specified for certain backfill types FHWA NHI-07-071 Earth Retaining Structures

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SC

(1)

65 – 100

75 – 105

70 - 95

80 – 100

95 – 120

100 – 120

95 – 120

105 – 125

(2)

45 - 21

36 - 19

40 - 24

33 - 21

24 - 12

22 - 12

24 - 12

19 - 11

15 - 11

16 - 11

21 - 12

16 - 9

14 - 9

12 - 8

14 - 11

11 - 8

Range of Optimum Moisture Percent(1)

ID

ID

230 ID

2,150 ID

420

ID

ID 1,500

270

460

(3)

ID

19

25

ID

28

32

32

31

230 (4) 190

33

34

420(4) 300 (4)

37

38

>31

>34

>37

>38

φ’ (Effective Stress Envelope Degree)

0

1,800

1,350

1,400

1,550

1,050

1,050

0

0

ID

ID

0

0

0

Cohesion (saturated) psf

0

0

Cohesion (as compacted) psf


2-27


Notes: ID: Insufficient data, Obtained from Standard Compaction test results, Strength characteristics are for effective strength envelopes and are obtained from USBR data, These soil types are generally not used as wall backfill, and (4)Cohesion values are reported in NAVFAC (1986); however use of nonzero cohesion value shall be used with caution as described in this chapter.

OH(3)

Organic clays and silty clays.

Inorganic clays of high plasticity.

CH

Inorganic clayey silts, elastic silts.

(3)

Organic silts and silt-clays, low plasticity.

Inorganic clays of low to medium plasticity.

Mixture of inorganic silt and clay.

MH(3)

OL

(3)

CL(3)

ML-CL

Inorganic silts and clayey silts.

Clayey sands, poorly graded sand-clay mix.

SM-SC

(3)

110 – 130

Sand-silt clay mix with slightly plastic fines.

ML(3)

110 – 125

Silty sands, poorly graded sand-silt mix.

SM

100 – 120

Poorly graded clean sands, sand-gravel mix.

SP

110 – 130

115 – 130

120 – 135

115 – 125

125 – 135

Range of Maximum Dry Unit Weight (pcf)(1)

Well graded clean sands, gravelly sands.

Well graded clean gravels, gravel-sand mixture. Poorly graded clean gravels, gravel-sand mix. Silty gravels, poorly graded gravel-sandsilt. Clayey gravels, poorly graded gravel-sandclay.

Soil Type

SW

GC

GM

GP

GW

Group Symbol

Typical Strength Characteristics(2)

Table 2-5. Typical Properties of Compacted Soils (after NAVFAC, 1986).

ID

0.35

0.47

ID

0.54

0.62

0.62

0.60

0.66

0.67

0.74

0.79

>0.60

>0.67

>0.74

>0.79

Tan φ’

Note: 1 kPa = 0.145 psi Figure 2-4. Typical Ranges of Friction Angle for Rockfills, Gravels, and Sands (after Terzaghi et al., 1996).

Table 2-6. Unconfined Compressive Strength of Particles for Rockfill Grades in Figure 2-4.

Rockfill Grade

Particle Unconfined Compressive Strength (psi)

A B C D E

≥ 32,000 24,000 to 32,000 18,000 to 24,000 13,000 to 18,000 ≤ 13,000


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Figure 2-5. Correlation between Drained Friction Angle and the Dry Unit Weight, Relative Density, and Soil Classification (after NAVFAC, 1986).

and for walls which support foundation loads. To obtain proper compaction, the moisture content of the furnished backfill material should be controlled, typically within optimum and 2 percent wet of the optimum moisture content determined from laboratory compaction tests. Fills which are formed by uncontrolled dumping of the material are generally unsatisfactory, and typically lead to long term settlement problems. During placement, moisture-density tests must be performed to verify that the in-place backfill material meets the specified compaction requirements. The backfill placed immediately behind the structure (i.e., at a distance of up to 3 ft from the FHWA NHI-07-071 Earth Retaining Structures

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5000 winitial = 23%

γd initial = 99.2 pcf LL = 41, PL = 24, and PI = 17 SHEAR STRESS, τ (psf)

4000

φ = 110, cT = 530 psf φ' = 270, c' = 286 psf

EFFECTIVE STRESS τf = 0.5105σf + 285.75

3000

TOTAL STRESS τf = 0.2013σf + 529.91

2000

1000

0 0

1000

2000

3000

4000

5000

6000

7000

8000

NORMAL STRESS, σ (psf)

Figure 2-6. Example Shear Strength Property Evaluation for Clayey Backfill. wall) is typically compacted with light tamping equipment to avoid development of excessive lateral earth pressures or displacement of the wall face. In some cases, as a means to facilitate adequate soil strength and stiffness of the backfill just behind the wall, coarser backfill soils (compared to other sections behind the wall) are used because they can achieve higher densities with less compactive energy. 2.5.7

Backfill Soil Permeability

The permeability of backfill soils is generally not measured since index parameters (e.g., gradation and Atterberg limits) can usually provide sufficient information on soil drainage characteristics. That is, a specific value for the coefficient of permeability (k) of the backfill is generally not required for design analyses. For cases where soil and/or geosynthetic filters are used for chimney and /or footing drains, a value for k may be required. Also, a relatively impermeable soil “cap” may be required to be constructed over the backfill soil to minimize stormwater infiltration into the backfill. In that case, a maximum value for k may be specified. The permeability of the backfill soils can be determined from laboratory tests that are FHWA NHI-07-071 Earth Retaining Structures

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conducted on specimens of recompacted materials. Two types of permeameters are used for laboratory testing, which are the rigid wall and the flexible wall permeameters. Rigid wall permeameter tests (AASHTO T215; ASTM D 2434) are not recommended for low permeability (i.e., k ≤ 10-6 cm/s (0.00283 ft/day)) soils (such as silts and clays) due to the potential for sidewall leakage. With flexible wall permeameters two types of tests can be performed, which are constant head and falling head tests (ASTM D5084). Details of permeameter tests and the interpretation methods of the test results can be found in Chapters 4 and 5 of GEC No. 5 respectively (Sabatini et al., 2002). 2.5.8

Filtration Requirements

2.5.8.1 Soil Filters Design criteria for soil filters are summarized below and are based upon gradations of the two adjacent soils. The particle sizes used in design are the D15, D50, and D85 sizes (subscript denotes the percentage of material, by weight, which has a smaller diameter). These criteria are applicable to adjacent soils with gradation curves that are approximately parallel. The equations are not applicable to gap-graded soils, soil-rock mixtures, non steady-state flow and soils with gradation curves that are not approximately parallel. When criteria are not applicable, filter design should be based upon laboratory filtration tests. See Cedergren (1989) for a comprehensive discussion on soil filtration. The soil filtration criterion to prevent piping (i.e., retention) of the upstream soil into the filter is: D15 filter < 5 D85 soil

(2-8)

To ensure sufficient permeability of the filter material, the ratio of the filter D15 to the upstream soil D15 should be greater than four to five, as shown in Equation 2-9. filter 5 < D15 D15 soil

(2-9)

An additional criterion to prevent movement of soil particles into or through filters is presented in Equation 2-10. For CL and CH soils without sand or silt partings, the D15 size of the filter in Equation 2-9 may be as great as 0.016 in and Equation 2-10 may be disregarded. However, if the upstream soil contains partings of uniform non-plastic fine sand and silt sizes, the filter must be designed to meet these criteria. FHWA NHI-07-071 Earth Retaining Structures

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D50 filter < 25 D50 soil

(2-10)

2.5.8.2 Geotextile Filters A geotextile is often used as a filter between a finer-grained and a more permeable soil. The geotextile must retain the finer-grained soil, while allowing water to readily pass into the more permeable soil, and function throughout the life of the earth retaining structure. Thus, geotextile design must address retention, permeability and clogging. The geotextile must also survive the installation process. The following design criteria are from the FHWA Geosynthetic Design and Construction Guidelines Manual (Holtz et al. 1998). For steady state flow conditions, the retention criterion is: AOS < B D85

(2-11)

where: AOS

=

apparent opening size of the geotextile

B

=

dimensionless coefficient

D85

=

soil particle size for which 85% are smaller

The AOS value of the candidate geotextile is determined from the results of the ASTM D 4751 test method, and is typically the value published by the geotextile manufacturers/suppliers. The B coefficient ranges from 0.5 to 2 and is function of the upstream finer-grained soil, type of geotextile, and/or the flow conditions. For sands, gravelly sands, silty sands and clayey sands (i.e., sands with less than 50% passing the No. 200 sieve), B is a function of the uniformity coefficient, Cu (Cu= D60/D10), of the upstream soil. B values for various Cu values are: Cu < 2

B=1

(2-12a)

2 < Cu < 4

B = 0.5 Cu

(2-12b)

4 < Cu < 8

B = 8 / Cu

(2-12c)

Cu > 8

B=1

(2-12d)


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If the upstream soil contains any fines, use only the portion passing the No. 200 sieve for selecting the geotextile. For silts and clays (more than 50% passing the No. 200 sieve), B is a function of the type of geotextile. for wovens,

B = 1

AOS < D85

(2-13a)

for nonwovens,

B = 1.8

AOS < 1.8 D85

(2-13b)

AOS < 0.012 in

and for both,

(2-14)

The above retention criteria are for internally stable soils. Laboratory performance tests should be conducted for such soils. Again, note that the above criteria are for steady state seepage. For dynamic flow conditions see Holtz et al. (1998). For steady state flow, low hydraulic gradient and well graded or uniform upstream soil, the permeability and permittivity criteria are: for permeability: kgeotextile > ksoil

(2-15)

for permittivity:

Ψ > 0.5 sec-1 for < 15% passing No. 200 sieve

(2-16a)

Ψ > 0.2 sec-1 for 15% to 50% passing No. 200 sieve

(2-16b)

Ψ > 0.5 sec-1 for > 50% passing No. 200 sieve

(2-16c)

where: k = Ψ =

coefficient of permeability (or hydraulic conductivity) and geotextile permittivity, which is equal to kgeotextile/tgeotextile.

For critical or severe applications, a geotextile permeability of 10 times the soil permeability should be used. The geotextile permittivity is determined from the results of the ASTM D 4491 test method.


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For steady state flow, low hydraulic gradient and well graded or uniform upstream soil, the clogging criterion is: AOS > 3 D15 (upstream soil)

(2-17)

This equation applies to soils with Cu > 3. For soils with Cu < 3, a geotextile with the maximum AOS value from the retention criteria should be used. See AASHTO M288 (1997) or see Holtz et al. (1998) for geotextile survivability critieria. For a more thorough treatment of geotextile drains see Holtz et al. (1998) and Koerner (1998). 2.5.9

Other Backfill Materials

2.5.9.1 Lightweight Materials Lightweight materials may be used in place of a select soil backfill to control the settlement of walls constructed over soft compressible soils. According to FHWA NHI-06-019 (Elias et al., 2006), lightweight fills can be grouped into two categories: (1) materials that behave and have similar properties to granular soils; and (2) materials that have an inherent compressive strength and behave similar to cohesive soils. Examples of the first group include wood fiber, blast furnace slag, fly ash, boiler slag, expanded clay or shale, and shredded tires. Examples of the second group include geofoam and foamed concrete. The dry loose unit weight of expanded shale, for example, typically varies from 45 to 65 pcf, while its compacted (Standard Proctor) dry unit weight ranges from 70 to 85 pcf. The internal friction angle for these materials may be 40o or higher, which corresponds to lower active earth pressure coefficients than for typical granular backfill soils. The lower earth pressure coefficient coupled with lower unit weight results in considerably lower earth pressures and corresponding reduction in bending moments in the wall structure and a reduction of the wall section. Table 2-7 provides a summary of typical dry densities for lightweight fills. In addition to the above advantages, the gradation of lightweight fills can be controlled to produce a free-draining backfill, thereby avoiding development of water pressures. Before using these materials in construction, however, laboratory testing should be performed to define their shear strength, unit weight, corrosivity and durability (soundness and resistance to abrasion). Lightweight fills must have adequate hardness and durability to resist degradation during placement and compaction, and to resist long-term deterioration in the underground environment. FHWA NHI-07-071 Earth Retaining Structures

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Table 2-7. Range of Dry Densities for Lightweight Fills (after Elias et al., 2006). Fill Type

Range in Dry Density (pcf)

Geofoam (EPS) Foamed Concrete Wood Fiber Shredded Tires Expanded Shale, Clay, and Slate (ESCS) Fly Ash Boiler Slag

0.75 - 2 20 - 60 35 - 60 38 - 56 38 - 65 70 - 90 70 - 94

The applicability of lightweight material may not be appropriate for all cases. For example, due to their light weight such materials may not be suitable for mechanically stabilized earth walls which rely on the overburden weight to generate friction along the reinforcing elements within the backfill material. Refer to the Lightweight Fill technical summaries in Ground Improvement Methods manual, FHWA NHI-06-019 (Elias et al., 2006) for discussions and details of lightweight fills. Geofoam blocks have been used in construction of several highway wall and embankment projects. 2.5.9.2 Flowable Fill Another backfill option is flowable fill. This backfill material is composed of cement in which air voids are distributed in the form of small, homogeneous, non-interconnected foam cells. High flowability and pumpability permits complete backfilling. In its hardened state, this type of backfill is stable. However, since it is in a fluid state initially, formwork is usually required. 2.5.9.3 Swelling Soils Swelling (or expansive) soils are typically clayey soils that undergo large volume changes in direct response to moisture changes in the soil and generally are not used as wall backfill. Swelling soils tend to increase in volume (i.e., swell) as the moisture content of the soil is increased and decrease in volume (i.e., shrink) as the moisture content of the soil is decreased. Although the expansion potential of a soil can be related to many factors (e.g., soil structure and fabric, environmental conditions, etc.), it is primarily controlled by the clay


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mineralogy. Soils that contain low-plasticity kaolinite will tend to exhibit a lower swell potential than soils containing high-plasticity montmorillonite. To identify expansive soils in the laboratory, several classification methods have been developed. Generally, soils with a plasticity index less than 15 percent will not exhibit expansive behavior. For soils with a plasticity index greater than 15 percent, the clay content of the soil should be evaluated in addition to the Atterberg Limits. Figure 2-7 shows the swelling potential of a remolded soil as related to the soil activity and clay fraction. For the purposes of evaluating expansion potential of a soil, activity can be defined as: Activity (A ) =

Plasticity Index (PI) Clay Fraction (CF)

(2-18)

where CF is the clay fraction that corresponds to the percentage of particles exhibiting an equivalent diameter (ds) < 0.078 mil (0.002 mm) as calculated from a hydrometer test performed in accordance with ASTM D422, Standard Test Method for Particle Size Analysis of Soils.

Note: 0.002 mm = 0.078 mil Figure 2-7. Classification Chart for Swelling Potential (after Seed et al., 1962). FHWA NHI-07-071 Earth Retaining Structures

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2.5.9.4 Degradable Materials On some transportation projects, construction activities involve the use of potentially degradable materials. Although the material may, at first, exhibit rock-like characteristics, it has the potential to degrade to soil-size particles. The gradual but ultimate degradation of the rock to the original parent soil material can occur within minutes or after several years of exposure to air and/or water. Shale, the most common member of this family of materials, can generically be considered to include claystone, siltstone, and mudstone. In many parts of the U.S., high-quality granular material is not locally available for use as borrow material. As a result, degradable materials that, at first, appear to be competent granular materials are used. However, once in contact with water, these materials may degrade causing problems and/or failures during the service life of the structure. Many rock types are prone to degradation when exposed to the cyclic wet/dry and freeze/thaw weathering processes. Rock types that are particularly susceptible to degradation due to these processes are poorly indurated shale and claystone exhibiting high clay content. The degradation can take the form of swelling, weakening, and ultimately disintegration. For wall backfills, the shear strength of the material may decrease with time resulting in greater earth pressures and continued wall lateral displacements. Methods to evaluate degradation potential are provided in Chapter 7 of GEC No. 5 (Sabatini et al., 2002).

2.6

PARAMETERS FOR IN-SITU SOILS

2.6.1

General

In-situ soil parameters used in the design of cut and fill wall systems are described in this section. In-situ soils refer to the retained ground behind the backfill zone for a fill wall and to the retained ground behind a cut wall. In-situ soils also include foundation soils below a fill wall and soils below the final excavated grade for a cut wall. The primary in-situ soil parameters required for retaining system design include stratigraphy, unit weight, shear strength, electrochemical properties, consolidation, and drainage characteristics. Selection of these parameters is provided below. 2.6.2

Stratigraphy

The soil stratigraphy at the site, including the thickness and elevation of individual soil and rock layers is evaluated through the implementation of a project-specific subsurface FHWA NHI-07-071 Earth Retaining Structures

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investigation, as previously described. In addition to drilling, sampling, and laboratory classification, subsurface stratigraphy information can be obtained from CPT and dilatometer testing. Details on stratigraphy evaluation using these methods are provided in Chapter 5 of GEC No. 5 (Sabatini et al., 2002). 2.6.3

Unit Weight

The unit weight of in-situ soils can be measured both in the field and in the laboratory. Measurement of unit weight in the field is generally restricted to shallow surface samples and can be accomplished using drive cylinder method (ASTM D 2937), sand cone method (ASTM D 1556), or nuclear gauge (ASTM D5915). To obtain unit weights with depth, either high-quality thin-walled tube samples must be obtained (ASTM D 1587) or relatively expensive geophysical logging by gamma ray techniques (ASTM D 5195) can be employed. Unit weight of in-situ soils can also be obtained from correlations with standard penetration test (SPT) (ASTM D 1586). The correlation between standard penetration number (N) and unit weight (γ) of granular soils is given in Table 2-8 and unit weight of cohesive soils is given in Table 2-9. These values should only be used for preliminary evaluations. Table 2-8. Correlation between N and γ of Granular Soils (after Bowles, 1988). Description Standard penetration no. N

Very Loose 0

Loose

Medium

4

Approx. range of γ pcf

70 – 100

Dense

10 90 – 115

30 110 – 130

Very Dense 50

110 – 140

130 – 150

Table 2-9. Correlation between N and γ of Cohesive Soils (after Bowles, 1988). Description N, standard penetration resistance

Very Soft 0

γ(sat), pcf


Soft 2

100 – 120

Medium 4

Stiff 8

110 – 130

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Very stiff 16

Hard 32

120 – 140


2.6.4

Shear Strength Parameters

2.6.4.1 General The shear strength parameters of the in-situ soils are used in evaluating the earth pressures on the active and passive side of the retaining wall systems. The shear strength of in-situ soils can be determined from laboratory tests, field tests, and empirical correlations. Laboratory shear strength tests for cohesive soils are performed on undisturbed samples obtained from the field. The shear strength of cohesionless soils can be evaluated from empirical correlations, based on in-situ test parameters (e.g., SPT N value), or from laboratory testing of recompacted (i.e., disturbed) samples. 2.6.4.2 Short-Term Conditions For design analyses of short-term conditions (i.e., temporary walls) in normally to lightly overconsolidated cohesive soils, the undrained shear strength, su, is commonly evaluated. A profile of su with depth should be developed at investigation points across the site. Since undrained strength is not a unique property, profiles of undrained strength developed using different testing methods will be different. Typical practice on retaining system projects is to develop profiles of su based on laboratory triaxial testing (i.e., consolidated-undrained (CU)) and for cases where undisturbed sampling is very difficult due to the soft nature of the soil, based on field vane shear testing. Although not generally recommended for evaluating undrained strength, in the UU test, the total stress undrained shear strength (su) of the soil is calculated based on the measured compressive strength of the soil. Shear strengths calculated from UU tests correspond to the depth at which the sample was taken from in the ground. The CU triaxial test provides data that can be used to interpret total stress strength parameters. With pore pressure measurement, effective stress parameters can also be obtained from CU tests (see Figure 28). Interpretation of UU and CU test results can be found in detail in Chapter 5 of GEC 5 (Sabatini et al., 2002). The vane shear test (VST) involves the use of a simple rotated blade to evaluate the undrained shear strength in soft to stiff clays and silts. Equipment necessary for a VST includes a four-sided vane with a height to diameter (H/D) ratio of 2, rods, and a torquemeasuring device. Vane size selection is a function of the anticipated strength of the soil and accuracy of the torque wrench. Larger vanes are typically used in soft soils and smaller vanes used in stiffer soils. The VST result has to be corrected to account for shear rate and strength anisotropy effects of the test. The correction is made by multiplying the su obtained FHWA NHI-07-071 Earth Retaining Structures

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Figure 2-8. Test Data from Triaxial Compression Test with Pore Pressure Measurements. from VST (su, VST) with a correction factor (µ). This µ factor is shown as a function of PI in Figure 2-9. Procedures of the VST and methods of interpretation are described in ASTM D 2573 in detail.


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Figure 2-9. Plasticity based VST Correction Factors. As mentioned above, a profile of su with depth should be developed to evaluate the su of insitu soils. Specific issues that should be considered when developing a profile of su with depth are described below. •

Strength measurements from hand torvanes, pocket penetrometers, or unconfined compression tests should not be used solely to evaluate undrained shear strength for design analyses. Consolidated undrained triaxial tests with pore pressure measurements and in-situ tests should be used.

•

All available undrained strength data should be plotted with depth. The type of test used to evaluate each undrained shear strength should be clearly identified. Known soil layering should be used so that trends in undrained strength data can be developed for each soil layer.

•

Data summaries for each laboratory strength test method should be reviewed. Moisture contents of specimens for strength testing should be compared to moisture contents of other samples at similar depths. Significant changes in moisture content will affect measured undrained strengths. Atterberg limits, grain size, and unit weight measurements should be reviewed to confirm soil layering.

•

A profile of preconsolidation stress, σp′ (or OCR) should be developed and used in evaluating undrained shear strength. Evaluation of σp′ is discussed in Section 2.6.5.3.


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(See Mayne et al. (2001) for guidance on use of empirical or constitutive based on σp′ or OCR to estimate su). In special cases involving excavations for walls (e.g., anchored walls and braced excavations) in deep deposits of soft clay, it is advisable to consider performing a supplemental series of traixial extension and direct simple shear tests. This is because the soil on the passive side of the wall is actually loaded in extension. Experience has shown that strength on the passive side of the wall can be lower than that in the active zone (i.e., behind the wall) due to soil anisotropy, especially for lean clays (i.e., CL soils). These effects lead to a lower stability condition than might be assumed on the basis of compression tests only. 2.6.4.3 Long-Term Conditions Long-term effective stress strength parameters (c′ and φ′) of clays are best evaluated by consolidated drained direct shear test, consolidated drained (CD) triaxial compression test, or CU triaxial tests with pore pressure measurements. In laboratory tests, the rate of shearing should be sufficiently slow to ensure substantially complete dissipation of excess pore pressure in the drained tests or, in undrained tests, complete equalization of pore pressure throughout the specimen. Information on appropriate shearing rates is provided in Chapter 4 of GEC No. 5 (Sabatini et al., 2002). Laboratory tests should always be carried out to displacements sufficiently large to reach a stable, post-peak shear stress. As previously discussed, the selection of peak, fully-softened, or residual strength for design analyses must be made based on a careful review of the expected or tolerable displacements of the soil mass. Where pre-existing weak interfaces are present, direct shear methods should be used, if practical, to set up the sample so that the interface strength can be directly evaluated. Alternatively, soil material from within the shear zone should be sampled and reconstituted to the in-situ moisture content for shear testing. For clays, empirical correlations have been developed to relate drained friction angle (φ′) (Figure 2-10) and drained residual friction angle (φ′r) (Figure 2-11) to the Atterberg Limit and to clay fraction, CF (for residual friction angle) characteristics of the soil. Figure 2-10 shows a slight trend of φ′ decreasing with increasing PI (Mesri and Abdel-Ghaffar, 1993), yet values can be ± 8° in variance. Considering the overall importance of drained friction angle for wall designs involving stiff to hard clays, it is essential to directly assess friction angle by means of the laboratory tests mentioned above. The consequences of estimating friction angle can be economically FHWA NHI-07-071 Earth Retaining Structures

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Fricttion Angle,

’ (degrees)

50 Mexico City Clay Attapulgite

40

30

20

10

0 0

Soft Clays Soft and Stiff Clays Shales Clay Minerials 20

40

60

80

100

200

400 600

1000

Plasticity Index, Ip (%)

Figure 2-10. Relationship between φ′ and PI (after Terzaghi et al., 1996).

psi 15 60 100

Note: CF = % particles < 0.078 mil Figure 2-11. Residual Friction Angles for Clayey Soils (after Stark and Eid, 1994). unwise. As an example, for relatively long, shallow slip surfaces that may be associated with a landslide, the required forces that would need to be resisted by some form of wall system would vary significantly depending on the drained friction angle of the soil. Therefore,


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correlations shown in Figures 2-10 and 2-11 should only be used to estimate drained friction angles of cohesive soils for preliminary analyses. Laboratory testing on undisturbed samples of granular materials is almost impractical. Therefore, it is necessary to rely on correlations to obtain the friction angle of granular soils. The strength of granular soils is typically assessed from penetration test data from field tests particularly from the SPT (Table 2-10 and Figure 2-12) and CPT (Figure 2-13). However, the SPT and CPT correlations should not be used to estimate the drained friction angle of mostly gravel soil, unless the correlations are used conservatively or can be modified based on local experience. For the correlations with SPT and CPT, the vertical effective stress should be calculated at the mid-depth of each soil layer. 2.6.5

Consolidation Parameters

2.6.5.1 General Consolidation parameters are used in settlement analyses for walls founded on compressible soils. Excessive total and/or differential settlement may be detrimental to the long-term performance of many types of wall systems. For instance, large differential settlements between the wall face and the backfill soil may induce significant distress in facing/reinforcement connections. Differential settlements along the wall alignment may affect the overall wall appearance (aesthetics) and performance of the wall facing for both fill and cut wall systems. While this section focuses on evaluation of soil parameters required for a settlement analysis, it is important to recognize that differential settlements (which may control the selection of a particular wall type for a project) are more a function of the thickness of relatively compressible material below the wall than the specific value of a particular compression parameter. For this reason, the necessity to accurately characterize the subsurface stratigraphy cannot be overemphasized. The discussion provided herein pertains primarily to fill wall systems in which the underlying foundation soils may undergo compression (i.e., settlement) in response to the weight of the wall and surcharges or for cut wall systems where the project may involve settlements due to groundwater drawdown. It is noted, however, that ground surface settlements behind a cut wall and vertical wall element settlement are important issues for cut walls. These types of settlements are evaluated semi-empirically and are discussed in Chapter 8.


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Table 2-10. Relationship between SPT N Value and Internal Friction Angle of Granular Soils (after AASHTO LRFD, 2007). State of Packing

Standard Penetration Resistance, N160 (blows/ft)

Friction angle, φ′ (°)

Very loose Loose Compact Dense Very dense

<4 4 10 30 50

25 - 30 27 - 32 30 - 35 35 - 40 38 - 43

Note: N160 = N60/(σ’v/Pa)0.5, where Pa = 1 tsf, σ’v = vertical effective stress, and N60 = ER/60% N (ER = hammer efficiency and N = uncorrected SPT blow count).

50

50

Friction Angle, φ' (degrees)

45

2 φ' = 26.5 + 0.4 N - N 500

σ'vo Pa

= 0.5

1.0

Very Dense 1.5 2.0 2.5 3.0

40

45 Dense 40

35

35

Medium 30

30

25

25 0 10 20 30 40 50 60 SPT N-value (blows/300 mm or bl/ft)

0 10 20 30 40 50 60 SPT N-value (blows/300 mm or bl/ft)

(a)

(b)

Loose Very Loose

Pa: normal atmospheric pressure = 1 atm ~ 100 kN/m2 ~ 1 tsf

Figure 2-12. Friction Angle of Cohesionless Soils (a) from Uncorrected SPT N Values (Modified after Peck, Hanson, and Thornburn 1974) and (b) as a Function of Normalized Overburden (Modified after Schmertmann, 1975).


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Cone Resistance, qt (psi) 0

1500

3000

4500

6000

7500

0

φ'=48°

Vertical Effective Stress (psi)

10

46° 20 44° 30 40° 40 38° 36° 50 28°

30°

32°

34°

60

Figure 2-13. Correlation of φ′ With Normalized CPT qt Data in Clean Sands (after Robertson and Campanella, 1983). The general consolidation characteristics of various major soil types can be used to evaluate whether a settlement problem should be anticipated for a particular wall according to the following: •

• •

Gravels, sands, and non-plastic silts: These soils consolidate rapidly under load and do not typically present settlement problems unless close tolerances are required for the project. Plastic silt-clay mixtures: Soft silts and clays are more compressible than stiff silts and clays. Settlement may continue long after construction is complete. Organic soils: These soils are very compressible as well as biodegradable and can result in large settlements that occur for many years.

Settlement analysis requires compression parameters and preconsolidation stress to be known. Compression parameters and preconsolidation stress can be estimated by correlation or by direct laboratory-measurement. Typically, the laboratory measurement is performed using one-dimensional consolidation test. The most common laboratory test method is the incremental load (IL) oedometer (ASTM D-2435). High-quality undisturbed samples using thin-walled tubes, piston samplers, or other special samplers are required for laboratory FHWA NHI-07-071 Earth Retaining Structures

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consolidation tests. Tests performed on disturbed samples will result in computed consolidation parameters that may result in conservative or unconservative design evaluations. Consolidation test results are traditionally shown in a graph of void ratio (e) versus logarithm of applied vertical effective stress (σv') in the oedometer, as illustrated in Figure 2-14. Each of the compression indices are defined by the change in void ratio per log cycle stress (= ∆e/∆logσv') for the respective ranges of recompression (Cr), virgin compression (Cc), and swelling or rebound (Cs). An alternate version of presenting consolidation test results is using a plot of vertical strain as the ordinate axis, whereby εv = ∆e/(1+eo). In this case, the compression indices are reported as the recompression ratio, Crε = Cr/(1+eo), and compression ratio, Ccε = Cc /(1+eo).

Figure 2-14. Definition of Cc, Cr, Cs, and σp′ (Sabatini et al., 2002).


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The maximum preconsolidation stress delineates the region of semi-elastic behavior (corresponding to overconsolidated states) from the region of primarily plastic behavior (associated with normal consolidation). The degree of preconsolidation is expressed using the overconsolidation ratio, where OCR = σp'/σvo'. An important parameter that is also obtained from the consolidation test that is used for settlement analysis includes a timedependent parameter, which is the coefficient of secondary compression (Cα). 2.6.5.2 Compression and Recompression Indexes The compression index (Cc) and recompression index (Cr), are index parameters required for primary consolidation settlement predictions. The value of Cc is evaluated by drawing a best-fit tangent line to data on an e-log σvc′ representation of consolidation data along the virgin (i.e., part of curve where stresses are greater than σp′) portion of the curve. The modified compression index, Cce, is evaluated similarly on a plot of εv-log σvc′. The line drawn to evaluate Cc (or Ccε) should include the stress range defined by the calculated σp′ and the final effective stresses in the ground. Typically, values of Cr (or Crε) are 10 to 20 percent of the value of Cc (or Ccε). Numerous correlations relating simple soil classification properties (e.g., LL, wn) to Cc and Ccε are available in the literature for silts and clays. These correlations can be used to make first-order predictions of settlements, but should not be relied upon for final design, unless the correlation has been developed using site-specific laboratory consolidation test data. These correlations may be of limited value, however, for highly structured soils that are sensitive. Several correlations that are based on relatively large databases are provided in Table 2-11.

Table 2-11. Summary of Correlations for Cc (after Holtz and Kovacs, 1981). Equation

Cc = 0.009 (LL –10) Cc= 0.007 (LL - 7) Cc = 0.01 wn


Applicable Soils

Undisturbed clays of low to medium sensitivity Remolded clays Chicago clays

2-48


Interpreting a Cc and Cr value for design should be based on a rational assessment of the data. The objective is to assign a value to each behaviorally different subsurface layer or to assign some representative value for the entire subsurface. Assessments to be made in evaluating compression data include: (1) depth ranges where the material is more silty or sandy as compared to other depth ranges; (2) depth of transition from a crust layer to an underlying softer clay layer; and (3) assessment of sampling disturbance. A detailed discussion on the assessment of these compression indices is provided in GEC No. 5 (Sabatini et al., 2002). 2.6.5.3 Preconsolidation Stress The preconsolidation stress (σp′) from oedometer tests is normally interpreted from the elogσv' relationship using the Casagrande graphical technique. In many clays and silts, this approach may be adequate in evaluating a reasonable value of σp'. The steps involved in the Casagrande method are provided in Table 2-12 and Figure 2-15. In very soft, sensitive, or structured materials, particularly those affected by sample disturbance, swelling, and the release of stress associated with removal from the ground, alternative graphical techniques can be utilized to better delineate the magnitude of σp′. Such techniques are described in GEC No. 5 (Sabatini et al., 2002). A profile of σp′ (or OCR = σp′/σvo′) with depth should be developed for the site. This parameter represents the most important value relative to settlement and shear strength evaluations for designs involving in-situ cohesive soils. An upper and lower bound profile should be developed based on laboratory tests and plotted with a profile based on particular in-situ test(s) (if used). There are no specific rules to judge the appropriateness of any particular value; however sample disturbance will typically be the most important factor affecting results. Sample disturbance indices are useful in assessing why, for example, a particular test provides a value of preconsolidation stress that seems too low. These indices are discussed in GEC No. 5 (Sabatini et al., 2002). Specific data quality issues that should be considered in evaluating preconsolidation stress are described below: •

Assess whether a consolidation curve that does not exhibit a sharp break is the result of sample disturbance or that the soil sample may contain relatively significant amounts of non-cohesive material (i.e., sand, non-plastic silt), which will result in a consolidation curve without a distinct break.


2-49


Table 2-12. Casagrande Method to Evaluate σp′. 1.

Construct a line tangent to the steepest portion of the consolidation curve (within the normally consolidated range). Locate the point of maximum curvature in the area of the laboratory curve where the slope transitions from shallow to steep. Construct a horizontal line from this point of maximum curvature. Construct a tangent line to the curve from the point of maximum curvature. Construct a line that bisects the angle between the horizontal line constructed in Step 2 and the tangent line constructed in Step 3.

2.

3. 4.

The point of intersection between the bisector line (Step 4) and the first tangent line (Step 1) is the location of the preconsolidation pressure. A diagram of the Casagrande method is provided in Figure 2-15.

0.85 1 0.75

2 σp' = 33 psi

Void Ratio, e

0.65

4 3

0.55

0.45

0.35

0.25 1

10

100

1000

10000

Effective Consolidation Stress (psi)

Figure 2-15. Illustration of Casagrande Method to Evaluate Preconsolidation Stress.


2-50


•

Make sure that the consolidation curve used to compute the preconsolidation stress corresponds to end of primary consolidation conditions. Most laboratories do not perform analyses to evaluate the end of primary condition, but simply provide consolidation curves that represent the end of the test.

•

The preconsolidation stress can only be reasonably estimated from laboratory oedometer tests if the test has been carried out to sufficiently high pressures. As a general rule of thumb, tests should be carried out to loads equal to approximately 4 to 8 times the estimated σp′. Previous testing and/or geologic information should be used to estimate σp′.

It is particularly important to accurately compute preconsolidation stress values for relatively shallow depths where in-situ effective stresses are low. An underestimate of the preconsolidation stress at shallow depths will result in overly conservative estimates of settlement for shallow soil layers. Preconsolidation stress can also be obtained from in-situ tests. However, it should be kept in mind that in-situ data only provides indirect measures of preconsolidation and therefore should be used to supplement values obtained from laboratory consolidation tests. Some of in-situ tests where σp′ can be estimated from include cone penetration test (CPT), flat dilatometer test (DMT), and field vane shear test (VST). Simple empirical and statistical expressions for estimating σp' from these test results are presented below. These correlations can be used to provide first-order estimates of σp' and to complement preconsolidation stress values evaluated from laboratory oedometer tests. Preliminary estimates of σp' for intact (i.e., not fissured) natural clays may be made using the approximate generalized trends as follows for each of the in-situ tests: Cone Penetration Test (CPT):

σp' = 0.33 (qT - σvo)

(2-19)

Flat (Plate) Dilatometer Test (DMT):

σp' = 0.51 (po - uo)

(2-20)

Field Vane Shear Test (VST):

σp' = 3.54 (su, VST)

(2-21)

where qT is tip resistance from CPT, σvo is the vertical stress, po is the liftoff pressure from DMT, uo is the initial pore water pressure, and su, VST is the undrained shear strength calculated from VST. Refer to GEC No. 5 for additional information regarding the use of insitu testing methods to estimate preconsolidation stress.


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2.6.5.4 Coefficient of Secondary Compression In geotechnical design analyses, it is assumed that secondary settlement occurs after primary consolidation is completed. As noted earlier, secondary compression settlements may be relatively large for organic soils. For example, welded wire mesh-faced MSE walls were used as part of the I-15 expansion project in Utah. Secondary compression settlements on the order of several feet were expected to occur. For this reason, the final permanent wall facing comprising a precast concrete facing was constructed several years after the MSE wall was completed. The analysis of secondary compression settlements includes evaluating time-dependent (creep) settlements for a given period of time. In engineering analyses, it is commonly assumed that the coefficient of secondary compression (Cα) can be represented as a constant value over the period of interest implying that the rate of secondary compression is independent of the thickness of the compressible layer. The accurate interpretation of Cα and Cαε is directly linked to the ability to accurately evaluate the time to the end of primary consolidation in a laboratory consolidation test. Computed values from consolidation tests should be compared to computed values for Cc and Cr and the ratios Cα/Cc and Cα/Cr should be compared to the expected values. For example, for inorganic clays and silts the Cα/Cc value averages 0.04±0.01, for organic clays and silts it averages 0.05±0.01, and for peats it averages 0.06±0.01. The coefficient of secondary compression is calculated using the portion of the deformation versus time plot for a consolidation test that corresponds to a time after primary consolidation is completed. For soils that are expected to demonstrate significant secondary settlements, it is important that load durations extend to times after primary consolidation is completed. Figure 2-16 shows a plot of deformation (for a particular load increment) versus logarithm of time for a soil sample. The Cα is evaluated according to the following equation: Cα =

∆e t log 2

(2-22) t1

where ∆e is the change in void ratio over an elapsed time equal to t2-t1. The times t1 and t2 occur after the time to the end of primary consolidation, tp. Alternatively, the modified coefficient of secondary compression, Cαε, can be evaluated similarly using a plot of volumetric strain versus logarithm of time.


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Stress Range 40 - 80 kPa 6 – 12 psi 2.65 2.6

Void Ratio (e)

2.55 2.5 2.45

One Log Cycle ∆ e=Cα=0.06

2.4 2.35 2.3

tp

2.25 0.1

1

10

100

1000

10000

Elapsed Time (min)

Figure 2-16. Evaluation of Cα.

2.6.6

Electrochemical Parameters

The tests previously described in Section 2.5.5 for soil electrochemical parameters are also used to evaluate the aggressivity of retained ground. These tests are required for the design of corrosion protection systems for ground anchors, soil nails, and micropiles. 2.6.7

Permeability

Groundwater conditions and the potential for groundwater seepage need to be considered in the design of earth retaining systems. As a result, the evaluation of groundwater conditions is a required element of a site investigation program for earth retaining systems. In cut wall applications, problems can be greatly aggravated by the presence of a source of artesian pressures or high seepage flow. For example, if during the excavation for a cut wall in relatively impermeable ground (e.g., fine grained in-situ soils) a water-bearing pervious seam (e.g., sand or gravel) is exposed, uncontrolled groundwater flows and/or ground mass instability may result. Critical ground types include silty fine sands or silts with low


2-53


plasticity because these soils can be source of high seepage flow. A discussion on the measurement of groundwater levels and water pressures is provided in Mayne et al. (2001). Permeability of in-situ soils can be determined from field or laboratory measurements. Field measurements include seepage tests, pumping tests, slug tests, and piezocone dissipation tests. Details of these tests can be found in Mayne et al. (2001). Laboratory tests for in-situ soils are performed the same way as for the backfill soils, except test specimens for in-situ soil tests are typically prepared from tube samples instead of recompacted materials as for backfill soils. The permeability test data, whether obtained from the field or the laboratory, should be scrutinized due to test complexities. These data should be considered in the context of the site geology and the method used to obtain the values. They are only as reliable as the method used to obtain them and it should be kept in mind that even excellent laboratory and field methodology may only provide values within an order of magnitude of actual conditions. The permeability test data (hydraulic conductivity) should always be accompanied by Atterberg limits and grain size distribution information for correlation with field conditions and for textbook correlations and approximations. Almost every soils, geology, and hydrogeology textbooks contain a chart of soil/rock type versus hydraulic conductivity. An example is given in Figure 2-17. After receiving the permeability test data, the hydraulic conductivity values should be checked against the estimated values. If there is a major discrepancy (more than two orders of magnitude) it should be noted that this could be a sign of test error, or possible anomalous soils. Hazen’s equation is the most common correlation equation used to estimate hydraulic conductivity for sands (k > 10-3 cm/s (2.83 ft/day)). This equation is written as: k = C(d10)2

(2-23)

where: k = hydraulic conductivity from permeability test (cm/s); C = coefficient ranging from 0.4 to 1.2 depending on sand size/sorting; and d10 = effective grain size in mm at 10% passing by weight.

Hazen’s equation is based solely on grain size and it requires input (d10) from particle size distribution curves and the use of a coefficient estimated based on sand type (e.g., fine sand, poorly sorted, etc.). Hazen’s equation should be used with caution since it only provides very approximate k estimates applicable only to clean sands (with less than 5% passing the No. 200 sieve) with d10 sizes between 0.004 and 0.12 in. (Holtz and Kovacs, 1981).


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COEFFICIE NT OF PERMEABILITY CM/S (LOG S CALE)

10

2

10

1

1.0

10-

1

10

-2

10

-3

10

-4

Good drainage

Drainage property

Application in earth dams and dikes

10

-6

10

-7

10

-8

10

-9

Practically impervious

Impervious sections of earth dams and dikes

Very fine sands, organic and inorganic silts, mixtures of sand, silt, and clay glacial till, stratif ied clay deposits, etc.

Clean sands, Clean sand and gravel mixtures

Type of soil

-5

Poor drainage

Pervious sections of dams and dikes

Clean gravel

10

“Impervious” soils e.g., homogeneous clays below zone of weathering

“Impervious” soils which are modified by the effect of vegetation and weathering; fissured, weathered clays; fractured OC clays Direct testing of soil in its original position (e.g., well points). If properly conducted, reliable; considerable experience required.

Direct determination of coefficient of permeability

(Note: Considerable experience also required in this range.)

Constant head test in triaxial cell; reliable w ith experience and no leaks.

Constant Head Permeameter; little experience required.

Reliable; Little experience required

Falling Head Per meameter; reliable; Range of unstable permeability;* Fairly considerable experience necessary much experience necessary to (do in triaxial cell) correct interpretation

Computation: From the grain size distribution (e.g., Hazen’s formula). Only applicable to clean, cohesionless sands and gravels

Indirect determination of coefficient of permeability

Computations: from consolidation tests; expensive laboratory equipment and considerable experience required.

Horizontal Capillarity Test: Very little experience necessary; especially useful for rapid testing of a large number of samples in the field w ithout laboratory facilities.

102

101

1.0

10-1

10-2

10-3

10-4

10-5

10-6

10-7

10-8

10-9

*Due to migration of fines, channels , and air in voids.

Figure 2-17. Range of Hydraulic Conductivity Values Based on Soil Type (after Holtz and Kovacs, 1981).

Another method of estimating hydraulic conductivity empirically through grain size was developed by GeoSyntec (1991) and is presented as Figure 2-18 As with Hazen’s equation, grain size distribution information is necessary to develop the input parameter (i.e., d15, grain size at 15% passing by weight) of the material. Using this value and the band of values corresponding to gradient and confining stress, a k value can be estimated. This method, as with other empirical methods, can only be used to provide an estimated value within 1 to 2 orders of magnitude of the in-situ condition. The lower the permeability and the higher the variability of grain size in the soil, the higher the error in using empirical relationships based on uniform particle distribution.


2-55


HYDRAULIC CONDUCTIVITY, ks (cm/s)

1.E+01

Low Gradient ~ 1 to 2

1.E+00 1.E-01

High Gradient ~ 10

1.E-02 Confining Stress ~ 5 35psi kPa

1.E-03 1.E-04 1.E-05 1.E-06

Confining Stress ~ 30 210psi kPa

1.E-07 1.E-08 1.E-09 1000

100

10

1

0.1

0.01

0.001

CHARACTERISTIC PARTICLE SIZE, d15, OF THE SOIL (mm)

Figure 2-18. Range of Hydraulic Conductivity Based on Grain Size (after GeoSyntec, 1991).

2.7

ROCK PARAMETERS

2.7.1

General

The purpose of this section is to provide an overview of rock parameters required for earth retaining wall system design, specifically rock mass classification, rock compressive strength, and rock shear strength. For walls retaining rock or for excavations performed in rock, it is important to know the possible failure modes of the rock. The actual rupture or sliding plane behind a retaining wall depends on the spatial orientation, frequency and distribution of the discontinuities, and the involved shear and interlock resistance to shear among them (Jumikis, 1983). In general, shear strength for rock is evaluated for smooth discontinuity, rough discontinuity, infilled discontinuity, and fractured rock mass. Detailed information on rock parameter for rock slope stability and rock slope design is provided in the NHI Manual “Rock Slopes”, FHWA HI-99-007 (Wyllie and Mah, 1998). 2.7.2

Rock Mass Classification

2.7.2.1 Description of Rock Masses Standardized geologic mapping and logging procedures should be used for describing rock masses. The types of information collected will depend on site access, extent of rock FHWA NHI-07-071 Earth Retaining Structures

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outcrops, and the criticality of the proposed structure to be constructed on or in the rock mass. A method proposed by the International Society of Rock Mechanics (ISRM) (1981) provides standardized quantitative and qualitative information on rock masses. This and other rock mass classification systems are described in ASTM D5878. To introduce and explain the ISRM method, several figures and tables were prepared, primarily to provide a standardized definition of terms. Figure 2-19 provides an illustration of a rock mass and the 13 parameters that are included in a detailed rock description. Figure 2-20 shows how these parameters are divided into five categories. Table 2-13 provides a brief description of each of the terms. In addition to ISRM (1981), details on the use of this method to characterize rock masses are provided in Wyllie (1999) and Wyllie and Mah (1998).

Figure 2-19. Illustration of geological mapping terms (after Wyllie, 1999, Foundations on Rock, Figure 4.4b, p. 101, E&FN Spon).

Using the terms described in Table 2-13 and the first category in Figure 2-20, a typical rock material description would be as follows (Wyllie, 1999): slightly weathered, crystalline, gray, fine grained, medium strong basalt FHWA NHI-07-071 Earth Retaining Structures

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Figure 2-20. List of parameters and categories describing rock mass characteristics (after Wyllie, 1999). FHWA NHI-07-071 Earth Retaining Structures

2-58


Table 2-13. Description of geological mapping terms. Term

Description

Rock Type

The rock type is defined by the origin of the rock (i.e., sedimentary, metamorphic, igneous), color (including whether light or dark minerals predominate), texture or fabric ranging from crystalline, granular, or glassy, and grain size ranging from boulders to silt/clay size particles.

Wall Strength

The compressive strength of the rock forming the walls of discontinuities will influence shear strength and deformability. Rock compressive strength categories and grade vary from extremely strong (> 250 MPa grade R6) to extremely weak (0.25 to 1 MPa grade R0) (see Table 2-14).

Weathering

Reduction of rock strength due to weathering will reduce the shear strength of discontinuities as well as reduce the shear strength of the rock mass due to the reduced strength of the intact rock. Weathering categories and grades are summarized in Table 2-15.

Discontinuity Type

The discontinuity type range from smooth tension joints of limited length to faults containing several centimeters of clay gouge and lengths of many kilometers. Discontinuity types include faults, bedding, foliation, joints, cleavage, and schistosity.

Discontinuity Orientation

The orientation of discontinuities is expressed as the dip and dip direction of the surface. Alternatively, the discontinuity can be represented by strike and dip. The dip of the discontinuity is the maximum angle of the plane to the horizontal (angle ψ in Figure 2-19) and the dip direction is the direction of the horizontal trace of the line of dip, measured clockwise from north (angle α in Figure 2-19).

Roughness

Roughness should be measured in the field on exposed surfaces with lengths of at least 2 m. The degree of roughness can be quantified in terms of the Joint Roughness Coefficient (JRC) as described in Section 2.7.5. Wall roughness is an important component of shear strength, especially in the case of undisplaced and interlocked features (e.g., unfilled joints).

Aperture

Aperture is the perpendicular distance separating the adjacent rock walls of an open discontinuity (thereby distinguishing it from the width of a filled discontinuity), in which the space is air or is water filled. Categories of aperture range from cavernous (> 1 m) to very tight (< 0.1 mm).

Infilling Type and Width

Infilling is the term for material separating the adjacent walls of discontinuities such as fault gouge; the perpendicular distance between adjacent rock walls is termed the width of the filled discontinuity. Filled discontinuities can demonstrate a wide range of behavior and thus their affect on shear strength and deformability can vary widely.

Spacing

Discontinuity spacing can be mapped in rock faces and in drill core; spacing categories range from extremely wide (> 6000 mm) to very narrow (< 6 mm). The spacing of individual discontinuities has a strong influence on the mass permeability and seepage characteristics of the rock mass.

Persistence

Persistence is the measure of the continuous length or area of the discontinuity; persistence categories range from very high (> 20 m) to very low (< 1 m). This parameter is used to define the size of blocks and the length of potential sliding surfaces. Persistence is important in the evaluation of tension crack development behind the crest of a slope.

Number of Sets

The number of sets of discontinuities that intersect one another will influence the extent to which the rock mass can deform without failure of the intact rock. As the number of sets increases and the block sizes reduce, the greater the likelihood for blocks to rotate, translate, and crush under applied loads.

Block Size and Shape

The block size and shape are determined from the discontinuity spacing, persistence, and number of sets. Block shapes include blocky, tabular, shattered and columnar, while block size ranges from very large (> 8 m3) to very small (< 0.0002 m3).

Seepage

Observations of the seepage from discontinuities should be provided. Seepage quantities in unfilled discontinuities range from very tight and dry to continuous flow. Seepage quantities in filled discontinuities range from dry in heavily consolidated infillings to filling materials that are washed out completely and very high water pressures are experienced.


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Table 2-14. Rock Material Strengths. Range of Uniaxial Compressive Strength (MPa)

Grade

Description

Field Identification

R0

Extremely weak rock

Indented by thumbnail

0.25 – 1.0

R1

Very weak rock

Crumbles under firm blows with point of geological hammer; can be peeled by a pocket knife

1.0 – 5.0

R2

Weak rock

Can be peeled by a pocket knife with difficulty; shallow indentations made by firm blow with point of geological hammer

5.0 – 25

R3

Medium strong rock

Cannot be scraped or peeled with a pocket knife; specimen can be fractured with single firm blow of geological hammer

25 – 50

R4

Strong rock

Specimen requires more than one blow of geological hammer to cause fracture

50 – 100

R5

Very strong rock

Specimen requires many blows of geological hammer to cause fracture

100 – 250

R6

Extremely strong rock

Specimen can only be chipped with geological hammer

> 250

1 MPa = 145 psi Table 2-15. Weathering Grades. Term Fresh

Description

Grade

No visible sign of rock material weathering; slight discoloration on major discontinuity surfaces is possible

I

Slightly weathered

Discoloration indicates weathering of rock material and discontinuity surfaces. All the rock material may be discolored by weathering and the external surface may be somewhat weaker than in its fresh condition.

II

Moderately weathered

Less than half of the rock material is decomposed and/or disintegrated to a soil. Fresh or discolored rock is present either as a discontinuous framework or as corestones.

III

Highly weathered

More than half of the rock material is decomposed and/or disintegrated to a soil. Fresh or discolored rock is present either as a discontinuous framework or as corestones.

IV

Completely weathered

All rock material is decomposed and/or disintegrated to soil. The original mass structure is still largely intact.

V

Residual soil

All rock material is converted to soil. The mass structure and material fabric are destroyed but the apparent structure remains intact. There may be a large change in volume, but the soil has not been significantly transported.

VI


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An example of a rock mass description using Table 2-13 and the four remaining categories in Figure 2-20 would be as follows: Columnar jointed basalt with vertical columns and one set of horizontal joints, spacing of vertical joints is very wide, spacing of horizontal joints wide, joint lengths are 3 to 5 m vertically, and 0.5 to 1 m horizontally; the discontinuity infilling is very soft clay with widths of 2 to 5 mm. The vertical columnar joints are smooth, while the horizontal joints are rough. No seepage observed.

2.7.2.2 Core Recovery and Rock Quality Designation The easiest way to characterize the amount of material recovered during rock coring is to calculate “core recovery” as the amount (i.e. length) of recovered material divided by the total length of the core run (presented as a percentage). Rock Quality Designation (RQD) (ASTM D6032) is a modified core recovery percentage in which the lengths of all sound rock core pieces over 4 in. in length are summed and divided by the length of the core run. Pieces of core that are not hard and sound should not be included in the RQD evaluation even if they are at least 4 in. in length. The purpose of the soundness requirement is to downgrade rock quality where the rock has been altered and/or weakened by weathering. For the RQD evaluation, lengths must be measured along the centerline of the core. Figure 2-21 illustrates the correct procedure for calculating core recovery and RQD. Table 2-16 presents the correlation of RQD to rock quality. Table 2-16. Rock Quality Description based on RQD. RQD Value

Description of Rock Quality

0-25 %

Very poor

25-50 %

Poor

50-75 %

Fair

75-90 %

Good

90-100 %

Excellent

The RQD is appropriate for use with all core sizes except for BQ and BX core with NX and NQ core size being optimal. Core breaks caused by the drilling process should be fitted together and counted as a single piece of sound core. Drilling breaks are usually evidenced by rough fresh surfaces, however for laminated rocks (i.e., rocks containing horizontally oriented fracture surfaces), it may be difficult to identify core breaks caused by drilling. FHWA NHI-07-071 Earth Retaining Structures

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L = 0 in.

L = 4.7 in.

L = 8 in.

L=0 < 4 in.

L = 2.4 in.

L = 3.2 in.

L = 7.5 in.

4 in.

L = 10 in.

L = 7.5 in.

L = 10 in.

L = 8 in.

L = 10 in.

Total length of rock re cov ered Total core run length

(10 + 8 + 10 + 7.5 + 2.4 + 3.2 + 4.7) in. 48 in.

(10 + 7.5 + 8) in. *100% 48 in.

Total core run length

∑ Length of sound pieces > 4 in.

RQD = 53%

RQD =

RQD =

CR = 95%

CR =

Core Re cov ery, CR =

2-62

Figure 2-21. Calculation of Core Recovery and RQD.

Core Run Total 48 in.


In this case, the RQD should be estimated conservatively; for shear strength characterization it is conservative to not count the length near horizontal breaks whereas for estimates of rock blasting requirements, it is conservative to count the length near horizontal breaks. 2.7.2.3 CSIR Classification The ISRM (1981) procedures, coupled with core recovery and RQD designations, provide a means for describing or characterizing rock and rock masses. A consistent method for rock mass classification is essential for evaluating rock properties used in design. The two most widely used rock classification systems were developed by the Council for Scientific and Industrial Research (CSIR) in South Africa (referred to as the “Q-rating system”) and the Norwegian Geotechnical Institute (NGI) (referred to as the Geomechanics system). Both of these methods are described in detail in Wyllie and Mah (1998). The CSIR classification system is the most widely used procedure in the US. Therefore, in this section an overview of the CSIR system is presented. The CSIR classification system considers the specific properties or conditions of the rock/rock mass, as well as an adjustment for the orientations of the joints. The following properties and conditions of the rock or rock mass are explicitly considered: (1) compressive strength of the intact rock (discussed in Section 2.7.3); (2) RQD value; (3) joint spacing; (4) condition of the joints; and (5) groundwater conditions. As shown in Table 2-17, each of these parameters is given a numerical rating based on the relative importance of the specific parameter on the behavior of the rock mass. This rating is adjusted to account for joint orientation depending on the favorability of the joint orientation for the specific project. The overall rating of the rock mass, termed the rock mass rating (RMR), is calculated as the sum of the individual ratings for each of the five parameters minus the adjustment for joint orientation (if applicable). Based on the final RMR, the rock mass is classified as: (1) Class I-very good rock; (2) Class II-good rock; (3) Class III-fair rock; (4) Class IV-poor rock; and (5) Class V-very poor rock. The RMR can be used to estimate rock mass deformation modulus and shear strength of the rock mass, however, an alternative (but similar) method (i.e., the Geological Strength Index described in the next section) is recommended herein. 2.7.2.4 Geological Strength Index The RMR system previously discussed has been shown to be useful for reasonable quality rock masses (i.e., 30 < RMR < 70), however, the RMR system is less useful for poorer quality rock masses primarily because of the importance of RQD in calculating the RMR. That is, for very poor quality rock masses, the RQD values may be zero (or close to zero), or more importantly, are just meaningless.


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Ground water

Relative Rating

Ratio= joint water pressure/major principal stress General Conditions

Inflow per 10 m tunnel length

Relative Rating

Condition of joints


5

4

OR

OR

25

0

None

20

Hard joint wall rock

Separation <1mm

Completely Dry 10

Hard joint wall rock

No separation

Not continuous

Slightly rough

Very rough surfaces surfaces

1 to 3m 25

>3 m 30

Spacing of joints Relative Rating

3

75% to 90% 17

12

15 90% to 100% 20

100 to 200 MPa

>200 MPa

4 to 8 MPa

Drill core quality RQD Relative Rating

Uniaxial compressive strength Relative Rating

Strength of intact rock material

>8 MPa

2

1

Point load strength index

PARAMETER

A. CLASSIFICATION PARAMETERS AND THEIR RATINGS

2-64

OR

OR

(interstitial water) 7

Moist only

0.0 to 0.2

<25 liters/min

12

Soft joint wall rock

Separation <1mm

Slightly rough surfaces

0.3 to 1 m 20

50% to 75% 13

7

50 to 100 MPa

2 to 4 MPa

50 to 300 mm 10

25% to 50% 8

4

25 to 50 MPa

1 to 2 MPa

OR

OR

4

Water under moderate pressure

0.2 to 0.5

25 to 125 liters/min

Continuous joints 6

Joints open 1 to 5 mm

or

Gouge <5 mm thick

Slickensided surfaces or

RANGES OF VALUES

Table 2-17. CSIR Classification of Jointed Rock Mass.

<50mm 5

<25% 3

1

0

OR

OR


0

Severe water problems

>0.5

>125 liters/min

0

Continuous joints

Joints open >5 mm

or

Soft gouge >5 mm thick

2

For this low range – uniaxial compressive test is preferred 10 to 25 3 to 10 1 to 3 MPa MPa MPa

Very favorable 0 0 0


RMR Rating Class No. Description

100 to 81 I Very good rock

C. ROCK MASS CLASSES DETERMINED FROM TOTAL RATINGS

Strike and dip orientations of joints Tunnels Ratings Foundations Slopes

B. RATING ADJUSTMENT FOR JOINT ORIENTATIONS

2-65

80 to 61 II Good rock

Favorable -2 -2 -5

60 to 41 III Fair rock

Fair -5 -7 -25

40 to 21 IV Poor rock

Unfavorable -10 -15 -50


<20 V Very poor rock

Very Unfavorable -12 -25 -60

An alternative system, referred to as the Geological Strength Index (GSI), was developed. This system does not require RQD, places greater emphasis on basic geological observations of rock mass characteristics, and was developed specifically for estimation of rock mass properties. This index is used with the Hoek-Brown failure criterion for rock masses to assess rock mass shear strength parameters (i.e., Mohr-Coulomb friction angle and cohesion intercept) and rock mass modulus. The present form of the GSI was developed to include poor quality rock masses (Marinos and Hoek, 2000, 2001) (see Figure 2-22) and heterogeneous rock masses (Marinos and Hoek, 2001) (see Figure 2-23). In using Figures 2-22 and 2-23, the description of the surface conditions correlates with descriptions provided in Table 2-15. For example, “Very Good” on Figure 2-22 corresponds to fresh, unweathered surfaces (i.e., Grade I from Table 2-15) and “Poor” on Figure 2-22 corresponds to highly weathered (i.e., Grade IV from Table 2-15). For better quality rock masses (GSI > 25), the value of GSI can be estimated directly using RMR by setting the groundwater rating to 10 (“dry”) and the adjustment for orientation set to 0 (“very favorable”). For very poor quality rock masses, the value of RMR is very difficult to estimate as rock cores will likely not provide intact core pieces longer than 4 in.; therefore RMR should not be used for estimating the GSI for poor quality rock masses, i.e., the GSI needs to be estimated directly using Figures 2-22 and 2-23 (Marinos et al.. 2004). The GSI system should not be applied to those rock masses in which there is a clearly defined dominant structural orientation (e.g., undisturbed slate). If the weathering of the rock mass has penetrated the rock to the extent that the discontinuities and the structure have been lost, then the rock mass should be treated as a soil and the GSI system no longer applies (Marinos et al., 2004). The use of the GSI system offers a quantitative approach to evaluating rock mass parameters for geotechnical design. It is noted, however, that while the method provides a way to capture detailed geological information obtained from a careful and thorough geologic characterization of the rock mass, only engineers, geologists, or engineering geologists experienced in using the system and whom have used the system in practice should be called on to assess rock mass parameters for design, particularly for large or critical foundation projects are being considered.


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Figure 2-22. General Chart for GSI Estimates from Geological Observations (after Marinos et al., 2004).


2-67



2-68

Figure 2-23. GSI Estimates for Heterogeneous Rock Masses (after Marinos et al., 2004).


2.7.3

Compressive Strength of Intact Rock

2.7.3.1 Point-Load Strength Test The point load strength test is an appropriate method used to estimate the unconfined compressive strength of rock in which both core samples and fractured rock samples can be tested. The test is conducted by compressing a piece of the rock between two points on coneshaped platens (see Figure 2-24) until the rock specimen breaks in tension between these two points. Each of the cone points has a 5-mm radius of curvature and the cone bodies themselves include a 60° apex angle. The equipment is portable, and tests can be carried out quickly and inexpensively in the field. Because the point load test provides an index value for the strength, usual practice is to calibrate the results with a limited number of uniaxial compressive tests on prepared core samples. The point load test is also used with other index values to assess degradation potential of shales. If the distance between the platens is D and the breaking load is P, then the point load strength, Is is calculated as: Is =

P 2 De

(2-24)

where De is the equivalent core diameter which is given by: (1) De = D2 for diametral tests; or (2) De = 4×A for axial, block, or lump tests where A = W×D. The area A is the minimum cross-sectional area of a lump sample for a plane through the platen contact points where W is the specimen width. The size-corrected point load strength index, Is(50) of a rock specimen is defined as the value of Is that would have been measured by a diametral test with D = 50 mm. For tests performed on specimens other than 50 mm in diameter, the results can be standardized to the size-corrected point load strength index according to: I s ( 50 ) = k PLT I s

(2-25)

The value of the size correction factor, kPLT, is given by: k PLT = ( D / 50) 0.45 ( D in mm)


2-69

(2-26)


Figure 2-24. Point load strength test equipment. It has been found, on average, that the uniaxial compressive strength, σc, is about 20 to 25 times the point load strength index, with a value of 24 commonly used, i.e.,

σ c = 24 I s (50)

(2-27)

However, tests on many different types of rock show that the ratio can vary between 15 and 50, especially for anisotropic rocks. Consequently, the most reliable results are obtained if a series of uniaxial calibration tests are carried out. Point load test results are not acceptable if the failure plane lies partially along a pre-existing fracture in the rock, or is not coincident with the line between the platens. For tests in weak rock where the platens indent the rock, the test results should be adjusted by measuring the amount of indentation and correcting the distance D (Wyllie, 1999). The point load strength test is not recommended for very weak FHWA NHI-07-071 Earth Retaining Structures

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rocks where the uniaxial compressive strength is less than approximately 3,600 psi. When performing the point load strength test, information on the degree of weathering should be noted since the compressive strength of weathered material along the joint may be less than the uniaxial compressive strength. 2.7.3.2 Unconfined Compressive Strength of Intact Rock Core The unconfined compressive strength of intact rock core can be evaluated reasonably accurately using ASTM D 2938. In this test, rock specimens or regular geometry, generally rock cores, are used. The rock core specimen is cut to length so that the length to diameter ratio is 2.5 to 3.0 and the ends of the specimen are machined flat. The ASTM test standard provides tolerance requirements related to the flatness of the ends of the specimen, the perpendicularity of the ends of the specimens, and the smoothness of the length of the specimen. The specimen is placed in a loading frame. Axial load is then continuously applied to the specimen until peak load and failure are obtained. The unconfined (or uniaxial) compressive strength of the specimen is calculated by dividing the maximum load carried by the specimen during the test by the initial cross-sectional area of the specimen. This test is more expensive than the point load strength test, but is also more accurate. Careful consideration of the design requirements should be made in deciding whether to perform this test or the simpler point load strength test. 2.7.4

Smooth Discontinuity

The shear strength for a smooth discontinuity is represented by a friction angle of the parent rock material. To evaluate the friction angle of this type of discontinuity surface for design, direct shear tests on samples can be performed. The range of typical friction angles provided in Table 2-18 should be used in evaluating measured values. Table 2-18. Typical Ranges of Friction Angles For A Variety Of Rock Types (after Barton, 1973 and Jaeger and Cook, 1976). Rock Class

Range of Friction Angles(1)

Typical Rock Types

Low Friction

20 – 27

Schist, shale, and marl

Medium Friction

27 – 34

Sandstone, siltstone, chalk, gneiss, slate

High Friction

34 – 40

Basalt, granite, limestone, conglomerate

Note: (1) Values assume no infilling and little relative movement between joint faces.


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2.7.5

Rough Discontinuity

The shear strength of a discontinuity surface is related to the roughness of the discontinuity (i.e., φ+i) where φ is the rock friction angle and i is the roughness of the discontinuity. The magnitude of additional strength resulting from roughness for a given normal load will depend on the roughness profile and the compressive strength of the joint wall. Equation 228 can be used to obtain an estimate of the joint roughness angle, i, for cases where the normal stresses for design are within the range of 1 to 30 percent of the joint compressive strength.  JCS  i = JRC log10  '   σ 

(2-28)

where JRC is the joint roughness coefficient, JCS is the compressive strength of the rock adjacent to the fracture surface, and σ’ is the effective normal stress. The roughness of the fracture surface is defined by the joint roughness coefficient. By comparing a fracture surface in the field to the standard profiles (see Figure 2-25), the JRC value can be evaluated. Normal stress ranges between 1 to 30 percent are common for many rock slopes. For values less than 1 percent or greater than 30 percent, this procedure is not valid. The maximum value for (φ+i) should be 50° and the minimum value is simply the friction angle of the intact rock corresponding to stress and deformation levels sufficiently large to have sheared through all asperities (i.e., zero dilation). A typical value for the basic rock friction (i.e., for a smooth discontinuity) is 30°. 2.7.6

Infillings

When a major discontinuity with a significant thickness of filling is encountered in a rock mass to be excavated or where the discontinuity is adversely dipping, the designer should assume that shear failure would occur through this material. Figure 2-26 presents a general method of characterizing joint fillings according to whether the filling material is undisplaced or displaced and whether the filling material is normally consolidated or overconsolidated. The effects of the type of infilling and parent rock can have a significant effect on strength parameters. For fillings that are displaced, it should be assumed that the material is at or near residual strength conditions in the field. The residual strength of the material can be evaluated in the laboratory using a direct shear-testing device in which samples of the filling material are FHWA NHI-07-071 Earth Retaining Structures

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reconstituted into the direct shear-testing device at the in-situ moisture content. If the filling is judged to be undisplaced, then the designer needs to evaluate whether the proposed structure can be expected to undergo minimal deformations such that the peak strength of the infilling can be used in design. If it becomes critical to evaluate the peak strength of the infilling (including both friction angle and cohesion intercept), it will likely be necessary to perform laboratory or in-situ direct shear tests of the infilling material. If laboratory testing is selected, great care must be exercised in obtaining samples for testing. The sample halves should not be displaced relative to each other and the samples should be sealed after collection to minimize moisture losses. 2.7.7

Shear Strength of Fractured Rock Masses

The shear strength of fractured rock masses can be evaluated using the method reported in Hoek et al. (2002). This method is based on the well-known Hoek-Brown failure criterion originally introduced in 1980. The method uses the GSI and is applicable over a range of very poor quality rock masses to stronger rocks. The three parameters defining the curved shear stress-normal stress envelope of the fractured rock mass are the uniaxial compressive strength of the intact rock pieces, σci, and three dimensionless constants mb, s, and a. The generalized Hoek-Brown criterion is expressed as:   σ' σ = σ + σ ci  mb 3 + s    σ ci ' 1

a

' 3

(2-29)

where σ1′ and σ3′ are the major and minor principal stresses in the rock mass and mb is a reduced value of the material constant mi and is given as:  GSI −100   mb = mi exp   28 − 14 D 

(2-30)

The constants s and a for the rock mass are given by:  GSI −100   s = exp   9 − 3D 

(

1 1 a = + e −GSI / 15 − e − 20 / 3 2 6


2-73

(2-31)

)

(2-32)


Figure 2-25. Definition of Joint Roughness Coefficient, JRC (after Barton, 1973). FHWA NHI-07-071 Earth Retaining Structures

2-74


Figure 2-26. Simplified Division of Filled Discontinuities into Displaced, and Normally Consolidated and Overconsolidated Categories (after Wyllie, 1999).

The parameter D is a factor which depends on the degree of disturbance to which the rock mass has been subjected by blast damage and stress relaxation. It varies from 0 for undisturbed in situ rock masses to 1 for very disturbed rock masses. The equations described herein are incorporated into the Windows program RocLab. RocLab can be downloaded (free of charge) from the internet site www.rocscience.com. Figures 2-27, 2-28, and 2-29 show typical windows in RocLab to estimate σci, GSI, and mi. With these values, the program calculates the Mohr-Coulomb shear strength parameters and FHWA NHI-07-071 Earth Retaining Structures

2-75


Figure 2-27. RocLab Window to Select Uniaxial Compressive Strength of Intact Rock.

provides the nonlinear Hoek-Brown failure envelope (see Figure 2-30). compressive strength of the rock mass is also provided.

The uniaxial

As an example, the RocLab software was used to develop rock mass shear strength parameters for a moderately to highly weathered clay shale in southeast Ohio. Laboratory uniaxial compression tests yielded an average σci of 0.44 kips per square inch (ksi). A GSI of 30 was selected with an mi value of 6 for shale. The best-fit shear strength parameters for this rock mass are a friction angle of 20.67º and a cohesion intercept of 0.013 ksi. In using these parameters, the user should take note the “best-fit” Mohr-Coulomb parameters tend to overestimate the Hoek-Brown shear strengths at low normal stresses. The user should select a shear strength consistent with the normal stress level being considered in the design.


2-76


Figure 2-28. RocLab Window to Select GSI.

2.7.8

Permeability

The in-situ permeability of rock should be investigated because the presence of groundwater in a rock mass can have detrimental effects on stability if the rock face is exposed as part of wall construction. Water pressures within the discontinuities (e.g., tension cracks or vertical fissures) of the rock can reduce the stability of the rock mass. The permeability of intact rock is generally very low and therefore poor drainage and low discharge would normally be expected in such material; however, if the rock is discontinuous, the permeability can be considerably higher. Table 2-19 summarizes the permeability of typical rocks. Typically, discontinuities in the rock act as channels for the water flow and virtually all groundwater storage and flow occurs in these discontinuities. The discontinuities can be empty (relatively higher permeability) or filled with clay (relatively lower permeability) (see Table 2-19). The influence of joint openings and joint spacing on permeability of rock masses is presented in Figure 2-31. FHWA NHI-07-071 Earth Retaining Structures

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Figure 2-29. RocLab Window to Select mi Value. The permeability of a rock mass can be determined from laboratory and field tests. However, it should be kept in mind that laboratory testing of intact rock is not a reliable method for measuring permeability of a rock mass. This is because the majority of the groundwater in rock mass is carried in the discontinuities and very little through the intact rock. The permeability of the rock mass in the field can be measured from pumping or Lugeon tests. These tests are performed by first pumping water into or out of a borehole section between two packers and then measuring the changes in groundwater levels in the rock due to this pumping. Details of these tests can be found in Wyllie and Mah (1998) and Mayne et al. (2001). It should be noted that just because no seepage appears on the rock face does not mean no groundwater is present in the rock. In many cases, the seepage rate may be lower than the evaporation rate and hence the rock face may appear dry. Yet there may be water at significant pressure within the rock mass. It is the water pressure and not the rate of flow that is responsible for instability in slopes and it is essential that such water pressures be considered in design. FHWA NHI-07-071 Earth Retaining Structures

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Figure 2-30. Shear Strength Parameters for Weathered Shale using RocLab Software. Table 2-19. Permeability of Typical Rocks and Soils (after Wyllie and Mah, 1998). k-cm/s

Practicallly Impermeable

10-10

Slate

10-9

Dolomite

10

-8

10-7 10-6 Low Discharge Poor Discharge

Intact Rock

Fractured Rock

Soil Homogeneous clay below zone of weathering

Granite Clay-filled joints

10-5

Very fine sands, organic and inorganic silts, mixtures of sand and clay, glacial till, stratified clay deposits.

10-4 10-3

High Discharge Free Drainage

10-2

Jointed Rock

10-1

Open-jointed rock

Clean sand, clean sand and gravel mixtures.

Heavily fractured rock

Clean gravel

1 101 102


2-79


Permeability coefficient, k (ft/sec)

1.E-01 1.E-02 1.E-03 1.E-04 1.E-05 1.E-06 1.E-07 1.E-08 0.0001

:0 ,b

.4

s he c n i

s he g c in cin :4 pa b s S g, he int c n i o n i J ac 40 Sp : t ,b in Jo ing c a Sp t in Jo

0.001

0.01

0.1

Joint Opening, e (in)

Figure 2-31. Influence of Joint Openings and Joint Spacing on the Permeability of Rocks (after Hoek and Bray, 1977).


2-80


CHAPTER 3 LATERAL EARTH AND WATER PRESSURES 3.1

INTRODUCTION

This chapter describes the principles for the evaluation of lateral earth pressures and water pressures used in the design of earth retaining structures. These pressures are used to compute a total lateral pressure diagram to be used in assessing the forces acting on the wall from the backfill or retained ground. This chapter focuses primarily on theoretical earth pressure diagrams which are most commonly used in the design of rigid gravity structures, nongravity cantilevered walls, MSE walls, and anchored walls with stiff structural facings (e.g., diaphragm walls). In LRFD, earth pressures and water pressures (and the resultant forces) are multiplied by a load factor and are combined into various load groups (see Section 3 of AASHTO (2007)). This load factor provides a quantitative way to account for the uncertainties associated with a given load. This chapter provides information on “unfactored” earth and water pressures which are equivalent to earth and water pressures that would be used in ASD. The use and application of load factors is described in subsequent chapters for each wall type. A wall system is designed to resist lateral earth pressures and water pressures that develop behind the wall. Earth pressures develop primarily as a result of loads induced by the weight of the backfill and/or retained (in-situ) soil, earthquake ground motions, and various surcharge loads. For purposes of earth retaining system design, three different lateral earth pressures are usually considered: (1) at-rest earth pressure; (2) active earth pressure; and (3) passive earth pressure. •

At-rest earth pressure is defined as the lateral pressure that exists in level ground for a condition of no lateral deformation.

•

Active earth pressure is developed as the wall moves away from the backfill or the retained soil. This movement results in a decrease in lateral pressure relative to the at-rest condition. A relatively small amount of lateral movement is necessary to reach the active condition.

•

Passive earth pressure is developed as the wall moves towards the backfill or the retained soil. This movement results in an increase in lateral pressure relative to the at-rest condition. The movements required to reach the passive condition are approximately ten times greater than those required to develop active earth pressure.


3-1

3 - Lateral Earth and Water Pressures June 2008

Each of these earth pressure conditions can be expressed by an equation with the general form:

σ h = Kσ v

(3-1)

where σh is the lateral earth pressure at a given depth behind the wall, σv, is the vertical stress at the same depth, and K is the earth pressure coefficient that can relate to at-rest earth pressure (Ko), active earth pressure (Ka), or passive earth pressure (Kp). The magnitude of the active and passive earth pressure coefficients are functions of the soil shear strength, the backfill geometry (i.e., horizontal backfill surface or sloping ground surface above the wall), the orientation of the surface where the wall contacts the backfill or retained soil (i.e., vertical or battered), and the friction and cohesive forces that develops on this surface as the wall moves relative to the retained ground. The magnitude of these earth pressure coefficients follow the relationship of Kp > Ko > Ka. The relationship between the earth pressure conditions and their associated lateral movements for a rigid gravity wall and for a nongravity cantilevered wall are illustrated in Figure 3-1. Additional discussion on the relationship between wall movement and development of earth pressures is provided in Section 3.5.

3.2

EARTH PRESSURE THEORY

3.2.1

Active and Passive Earth Pressures

Active and passive horizontal earth pressures may be considered in terms of limiting horizontal stresses within the soil mass, and, for purposes of this initial discussion, a smooth (i.e., zero wall friction) wall retaining ground with a horizontal backslope is considered (Figure 3-2); this case defines Rankine conditions. Consider an element of soil in the ground under a vertical effective stress, σv′ (Figure 3-3). In considering the potential movements of a retaining wall, the element may be brought to failure in two distinct ways. The horizontal soil stress may be increased until the soil element fails at Point B, when the stress reaches its maximum value σ′h (max). This scenario will occur when significant outward movement of the wall increases the lateral passive earth pressure in the soil at the base of the wall (see Figure 3-2). Similarly, the horizontal stress may be reduced until failure at Point A, when the stress reaches its minimum value σ′h (min). This scenario models the outward movement which reduces the lateral active earth pressures behind the wall (see Figure 3-2).


3-2


LATERAL PRESSURE

WALL MOVEMENT / WALL HEIGHT -

/H

ACTIVE CONDITION

PASSIVE CONDITION AT-REST CONDITION

H H Active pressure zone Active pressure zone

H

Passive zone Passive zone

Passive pressure zone

NON-GRAVITY CANTILEVERED WALL (CONTINUOUS VERTICAL ELEMENTS)

RIGID GRAVITY WALL

Figure 3-1. Magnitudes and Patterns of Movement to Develop Lateral Earth Pressures (after Sabatini et al., 1997). FHWA NHI-07-071 Earth Retaining Structures

3-3


Figure 3-2. Mobilization of Rankine Active and Passive Horizontal Pressures for a Smooth Retaining Wall.

Figure 3-3. Limiting Active and Passive Horizontal Pressures. FHWA NHI-07-071 Earth Retaining Structures

3-4


The geometry of Figure 3-3 gives the following two relationships:

σ ' h (min) 1 − sin φ ' = Ka = = tan 2 (45 − φ ' 2) ' σ v′ 1 + sin φ

(3-2)

σ'h (max) 1 + sin φ' = KP = = tan 2 (45 + φ' 2) σ′v 1 − sin φ'

(3-3)

where Ka is the active earth pressure coefficient and Kp is the passive earth pressure coefficient. The definitions of Ka and Kp, based on equations 3-2 and 3-3, are consistent with a Rankine analysis for a cohesionless (i.e., c=0) retained soil. For a cohesive soil defined by effective stress strength parameters φ′ and c′, the active and passive earth pressure coefficients are: K a = tan 2 (45 − φ ' / 2) −

K p = tan 2 (45 + φ ' / 2) +

2c '

σ v' 2c '

σ v'

tan 2 (45 − φ ' / 2)

tan 2 (45 + φ ' / 2)

(3-4)

(3-5)

In addition to the Rankine theory discussed above, values of active and passive earth pressure coefficients (Ka and Kp respectively) can also be calculated using the Coulomb theory. Rankine’s theory is discussed further in Section 3.2.3 and Coulomb’s theory is described in Section 3.2.4. 3.2.2

At-Rest Earth Pressure

The at-rest earth pressure represents the lateral effective stress that exists in a natural soil in its undisturbed state. For cut walls constructed in near normally consolidated soils, the atrest earth pressure coefficient, Ko, can be approximated by the equation (Jaky, 1944):

K o = 1 − sin φ '

(3-6)

where φ′ is the effective (drained) friction angle of the soil. The magnitude of the at-rest earth pressure coefficient is primarily a function of soil shear strength and degree of overconsolidation where this overconsolidation may result from natural geologic processes for retained ground or from compaction effects for backfill soils.


3-5


In overconsolidated soils, Ko can be estimated as (Schmidt, 1966): Ko =

σ h' = (1 − sin φ ' )(OCR) Ω ' σv

(3-7)

where Ω is a dimensionless coefficient which, for most soils, can be taken as sin φ′ (Mayne and Kulhawy, 1982) and OCR is the soil overconsolidation ratio (previously defined in Section 2.6.5). Usually, these correlations for the at-rest earth pressure coefficient are sufficiently accurate for normally to lightly overconsolidated soils provided the overconsolidation ratio has been evaluated from laboratory consolidation testing. For moderately to heavily overconsolidated clays, or where a more accurate assessment is required, laboratory triaxial tests on undisturbed samples and in-situ testing such as pressuremeter testing may be used (see Sabatini, et al., 2002). For normally consolidated clay, Ko is typically in the range of 0.55 to 0.65; for sands, the typical range is 0.4 to 0.5. For lightly overconsolidated clays (OCR ≤ 4), Ko may reach a value up to 1; for heavily overconsolidated clays (OCR > 4), Ko values may range up to or greater than 2. For heavily overconsolidated soils, values for Ko (and hence at-rest earth pressure) can be very large and would thus result in the need for a relatively stiff wall to resist these large forces. For walls constructed in such soils, consideration should be given to performing pressuremeter tests which provide a direct measure of lateral pressures in the ground. In the context of wall design using, for example, steel soldier beams or sheet-pile wall elements, design earth pressures based on at-rest conditions are not typically used. Using atrest earth pressures implicitly assumes that the wall system undergoes no lateral deformation. This condition may be appropriate for use in designing heavily preloaded, stiff wall systems, but designing to this stringent (i.e., zero wall movement) requirement for flexible wall systems is not practical. The relationship between earth pressures and movement for flexible anchored walls is discussed subsequently. 3.2.3

Rankine Earth Pressure Theory

The pressure distributions which result from an assumption of active or passive failure throughout the failing blocks of soil are shown in Figure 3-4(a) for the case of soil possessing no cohesion intercept and in Figure 3-4(b) for soil with a cohesion intercept. These assume a planar failure surface, a smooth wall and a horizontal surface to the retained soil. The FHWA NHI-07-071 Earth Retaining Structures

3-6


effective pressure acting at any point at depth, z, below the surface is, for a cohesionless soil and a groundwater table: pa′ = Ka (γz - u)

(3-8)

pp′ = Kp (γz - u)

(3-9)

and for a soil with an effective stress cohesion intercept pa′ = Ka (γz - u) - 2c′ K a

(3-10)

pp′ = Kp (γz - u) + 2c′ K p

(3-11)

In Figures 3-4(a) and (b), a constant soil unit weight, γ, is assumed and z is measured from the ground surface for active earth pressure calculations and from the excavated surface for passive earth pressure calculations. A triangular distribution of porewater pressure is also assumed. This corresponds in Figures 3-4(a) and (b) to a water table at the ground surfaces behind the wall and at the excavation level.

2c′√Ka

z

z

2c′√Kp z

z

u

p′p

u

p′a u (a)

p′p

p′a

u

(b)

Figure 3-4. (a) Wall Pressures for a Cohesionless Soil; and (b) Wall Pressures for Soil with a Cohesion Intercept (after Padfield and Mair, 1984).

The horizontal total stress acting on the wall is always the sum of the horizontal effective stress and the water pressure. Thus, at depth, z, the total stresses, pa and pp are given by FHWA NHI-07-071 Earth Retaining Structures

3-7


pa = pa′ + u

(3-12)

pp = pp′ + u

(3-13)

The pressures on the wall can also be considered solely in terms of total stress, but this is an approach which is only strictly valid for conditions immediately after wall construction. The relevant soil strength parameters are then c=su and φ=0. If su is substituted for c′ in Equations 3-12 and 3-13, with Ka = K = 1 for φ=0, the total stresses are given by: pa = Ka (γz - u) – 2su K a + u = γz – 2su

(3-15)

pp = Kp (γz - u) + 2 su = γz + 2su 3.2.4

(3-14)

Kp + u

(3-16) (3-17)

Coulomb Earth Pressure Theory

Coulomb theory is based on plane wedge theory and considers the effect of wall friction (via the angle δ in Figure 3-5), sloping backfill and sloping wall face. Calculation of active and passive earth pressure coefficients based on Coulomb theory is given in Figure 3-5. The active and passive pressures calculated with this method are commonly known as the Coulomb earth pressures. Since Coulomb’s method is based on limit equilibrium of a wedge of soil, only the magnitude and direction of the earth pressure is found with this method. Pressure distributions and the location of the resultant are assumed to be triangular. Specific limitations on the use of Coulomb theory for wall design are discussed in Section 3.2.5. Coulomb’s method will result in the same lateral pressures as for the Rankine analysis for the case where δ = θ = β = 0. 3.2.5

Effect of Wall Friction on Earth Pressures

In practice, walls are not smooth. Both wall friction and wall adhesion modify the stress distribution near a wall, so wall friction, δ, and wall adhesion, cw, should both be considered as proportions of φ′, and c′ or su, respectively. For the type of wall considered here, the frictional force exerted by the wall on the soil is in the sense shown in Figure 3-6. The active wedge moves down with respect to the wall, while the passive wedge moves upwards.


3-8


Figure 3-5. Coulomb Coefficients Ka and Kp for Sloping Wall with Wall Friction and Sloping Backfill (after NAVFAC, 1986). An important exception to this is when the wall acts as a significant load-bearing element, when large vertical loads are applied to the top of the wall, or when an inclined ground anchor is stressed to an appreciable load. In such cases, the wall has to move down relative to the soil on both sides of the wall in order to mobilize the required skin friction to support the load. Therefore, the friction acts to increase the pressures on both the active and passive FHWA NHI-07-071 Earth Retaining Structures

3-9


sides, because it acts on the soil wedges in a downward direction. This effect, however, is ignored because limiting (or failure) conditions are considered in calculation of overall stability and the directions in which the frictional forces act should be taken as shown in Figure 3-6.

Active wedge1

Passive wedge2

Note: (1) (2)

Active wedge moving downward relative to wall Passive wedge moving upward relative to wall.

Figure 3-6. Wall Friction on Soil Wedges (after Padfield and Mair, 1984). Wall friction has an important effect on soil pressures, and Equations 3-10 and 3-11 have to be made more general: pa′ = Ka(γz - u) - Kacc′

(3-18)

pp′ = Kp(γz - u) + Kpcc′

(3-19)

where Ka and Kp depend on δ and Kac and Kpc depend on δ and cw, and pa′ and pp′ are the components of effective pressure normal to the wall. Guidance on the selection of appropriate values of δ and cw for design are provided in Section 3.6.1 and are noted specifically, as necessary, for individual wall types in the remainder of this manual. The effect of wall friction on the Rankine and Coulomb methods of earth pressure computation is as follows: 1.

The Rankine method cannot take account of wall friction, and accordingly Ka is overestimated slightly, and Kp is under estimated, thereby making the Rankine method conservative for most applications.


3-10


2.

The Coulomb theory can take account of wall friction, but the results are unreliable for passive earth pressures for wall friction angle values greater than φ′/3 because of the assumed planar failure surface. The failure wedges assumed in the Coulomb analysis take the form shown as straight lines in Figure 3-7. This may be contrasted with the shapes observed in model tests, which are curved. The curvature results from the disturbing influence of wall friction on the stress field near the wall. The error in the Coulomb solutions is that Ka is underestimated slightly, while Kp is overestimated (very significantly for higher values of φ′).

If the angle of wall friction (δ) is low, the failure surface is almost plane. For high values of δ, the failure surface is curved and can be approximated by a log-spiral curve. The deviation of the curved surface from a plane surface is minor for the active case and significant for the passive case as shown in Figure 3-7. For most wall applications, the effect of wall friction on active earth pressures is relatively small and is often ignored. For the passive case, a large angle of wall friction causes downward tangential shear forces to act on the passive wedge of soil adjacent to the wall, increasing its resistance to upward movement. This, in turn, causes a curved failure surface to occur in the soil, as shown in Figure 3-7b. The soil fails on this curved surface of least resistance and not on the Coulomb plane, which would require greater lateral driving force. Hence, passive pressures computed on the basis of the plane wedge theory are always higher than those calculated on the basis of a log-spiral failure surface and may be on the unsafe side. 3.2.6

Equivalent Fluid Pressure

Equivalent fluid pressures provide a means of estimating design (unfactored) earth pressures. Although Clough and Duncan (1991) suggest that the use of equivalent fluid pressures is convenient for use with clayey backfills, AASHTO (2007) Section 3.11.5.5 states, “the equivalent- fluid method shall only be used where the backfill is free draining.” The basic earth pressure, p, is: p = γeqz

(3-20)

where: γeq = equivalent fluid density of soil, not less than 30 pcf and z = depth below surface of soil. The use of equivalent fluid pressure requires information on backfill soil material properties and geometry of the wall. For these reasons, the equivalent fluid density of soil can change if the wall cross section geometry changes.


3-11


“Actual” Failure Surface Failure Surface by Coulomb Analysis (a) Active Case (φ′ = 30°, δ = 30°)

(b) Passive Case (φ′ = 30°, δ = 30°) Figure 3-7. Comparison of Plane and Curved (Log-Spiral) Failure Surfaces (a) Active Case and (b) Passive Case (after Sokolovski, 1954).

The resultant horizontal earth load acts at a height of H/3. Typical values of equivalent fluid unit weights for wall heights not exceeding 20 ft for loose sand or gravel, medium dense sand or gravel, and dense sand or gravel are provided in Table 3-1. Values are presented for atrest conditions and for walls that can tolerate movements of 1 in. in 20 ft (i.e., ∆/H = 1/240) and for level backfill and sloped backfills with β equal to 25 degrees (approximately a 2H:1V backfill slope). Table 3-1. Typical Values for Equivalent Fluid Unit Weight of Soils (after AASHTO, 2007). Level Backfill Type of Soil

At-rest γeq (pcf)

Loose sand or gravel Medium dense sand or gravel Dense sand or gravel


55 50 45

Active ∆/H = 1/240 γeq (pcf) 40 35 30

3-12

Backfill with β = 25º At-rest γeq (pcf) 65 60 55

Active ∆/H = 1/240 γeq (pcf) 50 45 40


The vertical component of the earth pressure resultant for the case of sloping backfill surface can be calculated from:

3.3

TEMPORARY CONDITIONS

3.3.1

General

Pv = Phtanβ

(3-21)

Ph = 0.5γeqH2

(3-22)

The critical case for a wall constructed in soft to medium clays is the short-term (or temporary) case and earth pressures are evaluated using total stress evaluations (with undrained strength parameters) presented previously, or, for some cut walls, using apparent earth pressure diagrams (discussed subsequently). For a fill wall which uses clayey backfill, the condition of the clay soil will more likely be closer to a stiff to hard clay (due to typical field placement moisture content and use of compaction) thus making the long-term or drained condition more critical. A discussion on the appropriateness of using total stress analyses (with undrained strength parameters) for stiff to hard clays is discussed below. Additional discussion on the use of undrained versus drained strength parameters is provided in Section 2.3.2 3.3.2

Softening

In some circumstances it may be appropriate to apply total stress calculations using the soil undrained shear strength. The risk of employing such an analysis is directly related to the rate and degree of softening that the clay soil may experience during the service life of the wall. Construction of a retaining wall significantly changes the state of stress in the soil. Soil behind the wall is unloaded horizontally, and in service the long-term stresses are likely to be given by:


σv′ = γz –u

(3-23)

σh′ = Ka σv′

(3-24)

3-13


In this case, because Ka is less than unity, the average effective normal stress of the soil is now very much reduced (typically by a factor of between 2 and 3) compared to the stresses in the soil before the wall was constructed. The soil responds by drawing in water, thereby causing the soil to soften with time. Soil in front of the wall, at the surface of the excavation, is unloaded vertically. However, depending on the amount of this vertical unloading, the horizontal stresses may have increased, as a result of the passive pressure exerted by the wall. Nevertheless, the average normal stress generally decreases as a result of these two influences, because the reduction of vertical stress (caused by excavation) usually outweighs any increase in horizontal stress resulting from the pressure exerted by the wall. As is the case for soil behind the wall, porewater pressures are temporarily lowered in the short-term. As these return to equilibrium, the soil softens. Two features combine to make the process of softening proceed faster than would be predicted from consideration of just the permeability of the clay soil. These are tension cracks and soil fabric, which are discussed below. 3.3.3

Tension Cracks

The theoretical active pressure distribution for a soil possessing a cohesion intercept is shown in Figure 3-4(b). Immediately upon loading, when the original strength of the soil can be characterized by cu (also referred to as su), the diagram takes the form shown in Figure 3-8. The total active pressure in the short term acting on a smooth wall is given by Equation 3-15. Over a depth of about 2cu/γ, the theoretical lateral pressure is negative. Soil cannot support tension for long, so tension cracks may penetrate to this depth. The cracks may fill with any available water, and this then causes the surrounding soil to soften rapidly. Unless specific drainage measures are in place to prevent surface water infiltration, it should be assumed that the crack becomes filled with water and that hydrostatic water pressures act to this depth. Based on this discussion, one may conclude that if the cohesive strength component of the soil is large enough and if protection from surface water infiltration is provided, that there may be occasion where the wall would be supporting only negligible loads (if any). This may be the case for short-term conditions, but for most design analyses, the formation of tension cracks should be assumed and the depth of tension cracks can be estimated as: zo =


2 cu

γ

3-14

(3-25)


2cu 2cu/γ 2cu d

H

z

z

γd + 2cu

γ(d+H) - 2cu

Figure 3-8. Earth Pressure Immediately After Loading (after Padfield and Mair, 1984). However, for such cohesive materials, the critical shear strength condition corresponds to long-term conditions wherein the negative pore pressures in the soil (resulting from movement of the wall away from the backfill or retained ground) dissipate and the cohesive strength component may reduce to near zero. In general, the use of the cohesive component of strength in retaining wall designs should be based on previous experience in the same soil deposit. The depth of tension cracks for the long-term, drained condition of a cohesive soil can be estimated as: zo =

3.3.4

2c'

γ Ka

(3-26)

Soil Fabric

Stiff overconsolidated clays are generally layered as a result of variations in the depositional environment. Extremely small sandy or silty partings can have a large effect on the horizontal permeability of the soil. This influences the rate at which water may migrate from one part of the soil mass to another, and the rate at which water may be drawn into the soil. If the laminations dip, they may form preferred slip planes, with ready access to water, and these are therefore particularly prone to softening.


3-15


Superimposed on the laminated fabric of a typical stiff overconsolidated clay deposit is a network of joints and fissures. When stresses are released, the fissures may form interconnections between the coarser-grained partings, and they ensure that the permeability of the soil is much greater than is the case for an intact mass. Based on the discussion provided herein, it is apparent that the selection of engineering parameters for temporary wall designs involving stiff overconsolidated clays is difficult. This is further complicated by the difficulty in obtaining samples for laboratory testing since such soils may be highly fissured. Unless local experience in the same soil type and/or successful retaining wall design experience is available, it is prudent to evaluate earth pressures (and total pressures) for wall system designs in moderately to heavily overconsolidated materials using both undrained and drained strength parameters, and then to perform the design for the largest wall pressures.

3.4

PERMANENT CONDITIONS

For design of permanent earth retaining systems, long-term conditions need to be considered. The appropriate calculations should always be carried out in terms of effective stresses and water pressures. Under long-term conditions, in clays where the water table is close to the ground surface (as is often the case), the depth of any permanent tension crack is associated with the cohesion intercept, c′ according to Equation 3-26. Permanent tension cracks could theoretically only exist to relatively shallow depths below the surface, and, in this region, c′ values are usually very small because of weathering and the low stress level. Therefore, tension cracks under these long-term conditions are generally not assumed for design purposes. If the water table is well below ground level, tension cracks may develop as a result of negative porewater pressures in the clay above the water table (because of capillary rise), or as a result of desiccation of the upper layers. The depth of such cracks, however, depends on factors which are difficult to quantify (e.g. the degree of access of surface water to the clay, and the seasonal effects of droughts and vegetation). Under these conditions, tension cracks filled with water should be considered for design.


3-16


3.5

INFLUENCE OF MOVEMENT ON EARTH PRESSURES

3.5.1

General

The simple linear assumptions about active and passive pressures described previously are a considerable simplification of some very complex processes, which depend on the following factors: • mode of wall movement (e.g., rotation, translation) • wall flexibility • soil stress/strain properties • horizontal prestress in the ground (high Ko) • distribution of soil stiffness with depth. 3.5.2

Magnitude of Movement Needed to Mobilize Limit Pressures

The form of the relationship between the magnitude of movement of a retaining wall (in this case rotated into or away from the retained material about its toe) and the horizontal pressure exerted by the soil on the wall is usually presented as in Figure 3-9, with angular movement along the x axis and the mobilized coefficient of lateral earth pressure on the y axis. The curve shown relates to tests on dense sand, and was first derived by Terzaghi. Two observations should be made from Figure 3-9: 1. The magnitude of the passive coefficient is much greater than the active. 2. The deformation needed to mobilize the full passive limit pressure is much greater than that required to mobilize active resistance. For fill walls, however, zero passive resistance is usually assumed in the design so the required displacement to mobilize passive resistance is really a non-issue. For cut wall systems, however, the lateral resistance of the toe of the wall is a principal source of wall stability. While Figure 3-9 provides some quantitative information regarding movements, it does not provide any information on what the effect of the installation of the wall itself may be on future deformations. For example, the process of installation of a slurry (diaphragm) wall or tangent/secant pile wall prior to excavation in front of the wall will inevitably cause some ground movement, and therefore may reduce the horizontal stress in the soil from its original in-situ value. On the other hand, driving a sheet-pile wall tends to cause an increase of horizontal stress (1
3-17


3.5.3

Flexibility of the Wall

A relatively stiff wall attracts load and imposes a deformation condition on the soil which, in turn, affects the soil pressure distribution. The typical triangular lateral pressure diagrams assume a stiff wall structure. In contrast, in a flexible wall (e.g. an anchored or braced sheetpile wall), the mid-span is able to flex outward. This lowers the pressure at approximately the mid-height of the wall while simultaneously causing higher pressures to develop at the location of the brace (strut) or ground anchor. Peculiarities in the pattern of deformation can result in pressures lower than the Ka limit over parts of the wall, which are offset by corresponding areas at higher levels over which the pressures are significantly higher than the limit value. This phenomenon is often referred to as ‘arching’. 3.5.4

Influence of Movement on Soil Properties

The two main parameters characterizing clayey soil (i.e., cohesive soil) strength are both affected when an overconsolidated clay is sheared, but, for the purposes of retaining wall design, φ′ is affected much less than c′. For cohesionless soils, the drained friction angle is relatively unaffected by soil movement. The behavior of an overconsolidated stiff clay can be illustrated as shown in Figure 3-10. As the sample is sheared under drained conditions, the displacement of the soil sample is relatively uniform until the peak stress, τp, is reached. After the peak, displacements begin to concentrate on the newly formed failure plane or discontinuity, and the shear stress reduces to τd. The shear strength, τd, of the newly formed discontinuity is approximately equal to the shear strength of the same clay constituents in a normally consolidated state (i.e., the fully softened strength), such as that produced by laboratory consolidation from a slurry. For relatively high plasticity clays, further displacement beyond that corresponding to the fully softened strength results in a continued reduction in shear stress, and, eventually, at very large displacements along a major discontinuity, the residual strength of the clay soil, τr, is reached. For wall system design in stiff to hard overconsolidated clays, the design engineer must decide as to which strength, i.e., peak, fully softened, or residual, should be used to evaluate earth pressures for design. Since the additional strength at peak resulting from cohesion (c′ in Figure 3-10) tends to reduce relatively rapidly with increasing strain beyond peak, soil deformations associated with flexible walls may be sufficient to appreciably reduce this cohesion. Therefore, unless local experience indicates that a particular value of cohesion can FHWA NHI-07-071 Earth Retaining Structures

3-18


Figure 3-9. Effect of Wall Movement on Wall Pressures (after Canadian Foundation Engineering Manual, 1992)


3-19


Figure 3-10. Simplified Drained Stress-Displacement Relationship For a Stiff Clay (Modified After Padfield And Mair, 1984). be reliably accounted for, zero cohesion should be used for cut walls in stiff to hard fissured clays for long-term (drained) conditions. Conservative drained shear strength for analysis is therefore the fully softened strength. This strength may be evaluated using triaxial compression testing with pore pressure measurements. Most earth retaining walls in service will not experience sufficient movement to mobilize the residual strength and, therefore, as mentioned above, the fully softened strength is appropriate for design. Residual strengths should be used, however, for wall systems that are designed for a location in which there is evidence of an existing failure surface within the clay (e.g., an anchored system used to stabilize an active landslide). For these conditions, assume that sufficiently large deformations have occurred to reduce the strength to a residual value.


3-20


3.6

DESIGN EARTH AND WATER PRESSURES

3.6.1

Wall Friction and Adhesion

3.6.1.1 Total Stress An assumption has to be made about wall adhesion. For example, the driving of sheet piles into the soil increases horizontal stress prior to excavation, but deforms and smears the clay at the interface, installation of diaphragm and tangent/secant pile walls reduces horizontal stress prior to excavation, but probably deforms the soil less. As an analogy, it is reasonable to consider the observation that the ultimate skin friction of a drilled shaft in stiff clay can be approximated as αsu, where the value of α is approximately 0.5. The following maximum values in the active and passive zones are recommended for wall adhesion used for cut wall designs: Active:

cw = 0.5 su but ≤ 1,000 psf

Passive:

cw = 0.5 su but ≤ 500 psf)

3.6.1.2 Effective Stress Different values of δ are given by several sources, the values depending on the wall material (e.g. timber, steel, precast or in-situ concrete) (see Table 3-2). The maximum wall friction suggested for design is: • • •

Active: Passive: Passive:

δ = 2/3 φ′ δ = 1/2 φ′ (using Figure 3-12) δ = 1/3 φ′ (using Coulomb theory)

Where a cohesion intercept has been used as part of the characterization of strength in terms of effective stress, a maximum wall adhesion of cw = 0.5c′ could be used, but in view of the inevitable remolding of the clay close to the wall by any construction process, it is recommended that no wall adhesion be used in design. The values of wall friction provided above are maximum values for design. These can be adopted in most cases, but the design engineer should consider any circumstances in which they might be affected by the relative movement of the soil and the wall. On the active side, reduced values should be used if there is a tendency for the wall to move downwards (e.g., for load-bearing walls or walls supported by prestressed ground anchors). For walls retaining soft cohesive soils or granular soils that will be subjected to significant vibration (e.g., walls near railway tracks or machine foundations), δ should be assumed to be zero in the design. FHWA NHI-07-071 Earth Retaining Structures

3-21


Table 3-2. Wall Friction and Adhesion for Dissimilar Materials (after NAVFAC, 1986). Interface Materials Mass concrete on the following foundation materials: Clean sound rock Clean gravel, gravel sand mixtures, coarse sand Clean fine to medium sand, silty medium to coarse sand, silty or clayey gravel Clean fine sand, silty or clayey fine to medium sand Fine sandy silt, nonplastic silt Very stiff and hard residual or preconsolidated clay Medium stiff and stiff clay and silty clay (Masonry on foundation materials has same friction factor) Steel sheet piles against the following soils: Clean gravel, gravel-sand mixtures, well-graded rock fill with spalls Clean sand, silty sand-gravel mixtures, single size hard rock fill Silty sand, gravel or sand mixed with silt or clay Fine sandy silt, non plastic silt Formed concrete or concrete sheet piling against the following soils: Clean gravel, gravel-sand mixtures, well-graded rock fill with spalls Clean sand, silty sand-gravel mixtures, single size hard rock fill Silty sand, gravel or sand mixed with silt or clay Fine sandy silt, non plastic silt Various structural materials: Masonry on masonry, igneous and metamorphic rocks: Dressed soft rock on dressed soft rock Dressed hard rock on dressed soft rock Dressed hard rock on dressed hard rock Masonry on wood (cross grain) Steel on steel at sheet pile interlocks

Interface Materials (Cohesion)

Friction angle, δ degrees

0.70 0.55 to 0.60 0.45 to 0.55

25 29 to 31 24 to 29

0.35 to 0.45 0.30 to 0.35 0.40 to 0.50 0.30 to 0.35

19 to 24 17 to 19 22 to 26 17 to 19

0.40 0.30 0.25 0.20

22 17 14 11

0.40 to 0.50 0.30 to 0.40 0.30 0.25

22 to 26 17 to 22 17 14

0.70 0.65 0.55 0.50 0.30

35 33 29 26 17

Adhesion ca (psf)

Very soft cohesive soil (0 – 250 psf) Soft cohesive soil (250 – 500 psf) Medium stiff cohesive soil (500 – 1000 psf) Stiff cohesive soil (1000 – 2000 psf) Very stiff cohesive soil (2000 – 4000 psf)

3.6.2

Friction Factor, tan δ

0 – 250 250 – 500 500 – 750 750 – 950 950 - 1300

Earth Pressure Coefficients

3.6.2.1 Total Stress Design Total stress theory is appropriate for temporary conditions in normally to lightly overconsolidated clays and is appropriate for stiff to hard clays subject to the limitations discussed in Section 3.3. FHWA NHI-07-071 Earth Retaining Structures

3-22


The total active and passive total pressures acting on a wall are given by: pa = Ka σv - Kac su

(3-27)

pp = Kp σv + Kpc su

(3-28)

where σv = γz; Ka = Kp = 1; and Kac = Kpc = 2 1+ c w / su If negative active pressure is predicted by Equation 3-27, the value of cw should be taken as zero, because it cannot be relevant if there is a tension crack. Considerable care should therefore be exercised in use of cw in a total stress analysis. In determining the possible depth of tension crack, cw should be ignored. To take account of softening and the possibility of water filling the tension cracks, the total active pressure on the wall at any depth below the ground surface should be assumed to not be less than 30z psf (z = depth in ft). This is sometimes referred to as the minimum equivalent fluid pressure (MEFP), the density of the equivalent fluid being taken as 30 lb/ft3. 3.6.2.2 Effective Stress Design The calculation of horizontal effective earth pressures resulting from soil weight, pa′ and pp′, involves the calculation of porewater pressures, u, for use in the following equations. pa′ = Ka (γz –u) - Kac c′

(3-29)

pp′ = Kp (γz –u) + Kpc c′

(3-30)

The porewater pressure, u, is added to the horizontal effective stress, pa′ or pp′, to give the total horizontal stress, pa or pp, acting on the wall. Exact calculation of u is sometimes difficult. Simple assumptions concerning water pressures for design are discussed in Section 3.6.5. Where c′ is incorporated into the soil strength characterization, values of Kac and Kpc should be calculated from the following approximate expressions:


3-23


K ac = 2 K a (1 + c w / c ' )

(3-31)

K pc = 2 K p (1 + c w / c ' )

(3-32)

Use of high values of c′ (greater than say, 100 psf) in the retained soil results in a significant depth of theoretical negative active earth pressure. It is important either to: • •

reduce c′ towards the surface to avoid this, which may be realistic for many clays in view of weathering; or assume that the effective pressure on the wall at any depth should not be less than 30z psf (z = depth in ft).

The MEFP of 30 lb/ft3 should only be applied to depths where the calculated total horizontal pressure (effective pressure plus water pressure) would otherwise be less than the MEFP. In many cases, therefore, when water pressure is assumed to act, the MEFP need not be applied, even though a theoretical negative active earth pressure may result from an assumption of c′ acting (see Example Problem 2 in Section 3.13). It is recommended that the log-spiral theory be used for determination of the passive earth pressure coefficients. Charts for two common wall configurations (sloping wall with level backfill and vertical wall with sloping backfill) based on the log-spiral theory (Caquot and Kerisel, 1948; NAVFAC, 1986) are presented in Figures 3-11 and 3-12. For walls which have a sloping backface and a sloping backfill, the passive pressure coefficient can be calculated as indicated in Figure 3-5 using a maximum of δ= φ′/3. For the active case, the resultant load predicted using coefficients based on the plane wedge theory is within 10 percent of that obtained with the more exact log-spiral theory. Hence, for the active case, Coulomb’s theory (Figure 3-5) can be used to calculate the earth pressure coefficient. For some wall types, such as cantilever retaining wall and an MSE wall, the “interface” is within the retained soils where the earth pressures are computed along a vertical soil plane passing through the heel of the base slab. In such cases, there is soil-to-soil contact and the resultant may be oriented at the angle of mobilized friction. The angle of mobilized friction depends on the factor of safety used for the angle of internal friction. For these cases, it is generally conservatively assumed that the earth pressure is parallel to the slope of the backfill.


3-24


Figure 3-11. Passive Coefficients for Sloping Wall with Wall Friction and Horizontal Backfill (Caquot and Kerisel, 1948; NAVFAC, 1986).


3-25


Figure 3-12. Passive Coefficients for Vertical Wall with Wall Friction and Sloping Backfill (Caquot and Kerisel, 1948; NAVFAC, 1986). FHWA NHI-07-071 Earth Retaining Structures

3-26


3.6.3 Theoretical Earth Pressures in Stratified Soils

For stratified (or nonhomogeneous) soils, the theoretical earth pressures are assumed to be distributed as shown in Figure 3-13. Unless the computed earth pressures vary widely with depth, the total applied lateral force determined from the computed pressure diagram may be redistributed to a corresponding simplified equivalent triangular pressure diagram as indicated in Figure 3-13.

• • •

Use buoyant unit weight for soils below water table. Add water pressure as appropriate to obtain total lateral pressure. The simplified distribution may not be justified for all soil conditions. Use judgment to determine validity of such simplified distributions.

Figure 3-13. Pressure Distribution for Stratified Soils. For complex cases such as layered soils, irregular backfill, irregular surcharges, wall friction, and sloping groundwater level, pressures can be determined by graphical solutions. Among the many graphical solutions are Culmann’s (ca. 1866) and the trial wedge method (ca. 1877). These procedures can be found in Bowles (1988) or NAVFAC (1986).


3-27


3.6.4

Semi Empirical Earth Pressure Diagrams

The earth pressures discussed in the above sections are strictly applicable to rigid wall systems; i.e., walls which move as a unit (rigid body rotation and/or translation) and do not experience bending deformations. Most gravity walls can be considered rigid walls. If a wall system undergoes bending deformations in addition to rigid body motion then such a wall system is considered flexible. Virtually all wall systems, except gravity walls, may be considered to be flexible. The bending deformations result in a redistribution of the lateral pressures from the more flexible to the stiffer portions of the system. Thus, in these walls the final distribution and magnitude of the lateral earth pressure may be considerably different than those used for rigid walls. For example, the soldier-pile and lagging walls with multiple levels of support are usually designed using empirical earth pressure distributions based on observed data. These empirical earth pressure distributions may vary from rectangular to trapezoidal shapes and may have varying magnitudes depending on the soil type (see Chapter 8). The results from a study on the performance of a two-level model anchored wall may be used to illustrate the relationship between lateral earth pressure and wall deformation for relevant anchored wall construction stages. The model wall was 6.25 ft high with a final toe embedment of 1.25 ft (Figure 3-14). The results of the model wall study are described in FHWA-RD-98-067 (1998).

1.5 ft 100 pcf 2.5 ft 6.25 ft

1.25 ft

Figure 3-14. Cross Section of Model Wall (modified after Mueller et al., 1998).


3-28


•

Cantilever Stage: During the cantilever stage of construction, soil is excavated down to a level just below the elevation of the first ground anchor. For the portion of the wall above the first excavation level, the earth pressure and deformation pattern are generally consistent with that of active conditions (i.e., a triangular pressure distribution) (Figure 3-15). The wall is in a condition of “fixed-earth support.”

Figure 3-15. Lateral Wall Movements and Earth Pressures with Excavation at First Anchor Level (Cantilever Stage) (modified after Mueller et al., 1998).

•

Stressing of Upper Ground Anchor: Significant changes in lateral earth pressure occur as a result of anchor stressing (Figure 3-16). During stressing, the soldier beam is pushed against the retained ground, resulting in a large increase in lateral pressures that may approach full passive pressures in the vicinity of the load. When the load is reduced to the lock-off load, typically 75 to 100 percent of the unfactored load (referred to as “DL” in Figures 3-16, 3-17, and 3-18), the pressure decreases leaving a pressure bulb around the anchor. Note that this pressure is in excess of active pressures.


3-29


Figure 3-16. Lateral Wall Movements and Earth Pressures During Anchor Stressing (modified after Mueller et al., 1998).

•

Excavation to Lower Anchor: Excavation below the upper anchor results in lateral bulging of the wall and a redistribution of earth pressure (Figure 3-17). Earth pressure between the upper anchor and the excavation subgrade is reduced and load is redistributed to the stiffer upper anchor and excavation subgrade resulting in earth pressure increases in these areas.

•

End of Construction: Stressing of the lower ground anchor results in a local wall deformation pattern similar to that resulting from stressing the upper anchor (Figure 318). A pressure bulb also develops at the location of the lower anchor. As a result of excavation to the final design grade, lateral bulging occurs between the lower anchor and the excavation subgrade.


3-30


Figure 3-17. Lateral Wall Movements and Earth Pressures with Excavation at Lower Anchor Level (modified after Mueller et al., 1998). Lateral wall displacement, y/H (%) 0.3

0.2

0.1

0.0

-0.1

Upper ground anchor

Earth pressure after excavation to final depth

H

Design earth pressure envelope Lower ground anchor

Lateral wall displacement after excavation to final depth

Lateral wall displacement after lower ground anchor lock-off at 0.75 DL

-0.1 0.0 0.1 0.2 0.3 0.4 Lateral earth pressure

H

Figure 3-18. Lateral Wall Movements and Earth Pressures with Excavation at Design Grade (modified after Mueller et al., 1998). FHWA NHI-07-071 Earth Retaining Structures

3-31


A trapezoidal-shaped envelope of earth pressure is also shown on Figure 3-18. This envelope is referred to as an “apparent earth pressure envelope” and closely approximates the shape and magnitude of the earth pressures after completion of the wall. Apparent earth pressure envelopes are described in detail in Chapter 8 and an example apparent earth pressure envelope for anchored walls in sands in shown in Figure 3-19. It is noted that earth pressure envelopes which assume fully active conditions (i.e., triangular pressure distribution) would overestimate the earth pressures near the excavation subgrade, resulting in overly conservative estimates of wall bending moments and required wall embedment depth, and underestimate anchor loads and wall bending moments at upper tier supports. Other factors which may influence the development of earth pressures are the type of construction (e.g., “bottom-up” or “top-down”), the wall support mechanism (e.g., tiebacks, struts, rakers, soil nails, reinforcing elements, single or multiple levels of support, etc.), the geometry of the retained soil (e.g., silo pressure), the superimposed or surcharge loads (e.g, strip, line, concentrated or equipment loads) and the type of analysis (static or seismic). In addition, for cases of soil reinforced by inclusions such as MSE walls or soilnailed walls, different types of earth pressure distributions are used to evaluate the internal and external stability of the wall system. 3.6.5

Design Water Pressures

In retaining wall design, it is general practice to provide drainage paths (commonly known as “weep holes”) through the earth retaining structure, or use other methods to drain groundwater that may otherwise collect behind the structure, and thereby avoid the development of water pressure on the structure. Occasionally, however, it may not be feasible or desirable to drain the water from behind the structure to safeguard against potential settlement of adjacent structures, or prevent contaminated groundwater from entering the excavation. In such instances, the earth retaining structure must be designed for both lateral earth pressure and water pressure. Computation of lateral earth pressures for the case of a uniform backfill and static groundwater is illustrated in Figure 3-20. In this case, the water pressure represents a hydrostatic condition since there is no seepage or flow of water through the soil. The lateral earth pressure below the water level is based on the effective vertical stress, σ′v.


3-32


Figure 3-19. Apparent Earth Pressure Diagram for Anchored Wall in Sand. Pa=Kz[γ hw+ γ’zw]

Pa = Ka γz z

Pw= γ wzw

hw zw

H Pa = ½ Ka γz2 (No water level)

Pa

Hw Pw

Impervious

Figure 3-20. Computation of Lateral Pressures for Static Groundwater Case. FHWA NHI-07-071 Earth Retaining Structures

3-33


The lateral pressure due to water is added to the lateral earth pressure to obtain the total lateral pressure on the wall. The lateral pressure computations should consider the greatest unbalanced water head anticipated to act on the wall, since this generally results in the largest total lateral load. For cases where seepage may occur through or beneath the earth retaining structure, the resulting seepage gradients will result in a reduction in the water pressure. For such cases, flow net procedures can be used to compute the lateral pressure distribution due to water. A typical flow net for a retaining wall in homogeneous soil is shown in Figure 3-21. The calculation of pore water pressure may be simplified by assuming that the head difference (H+i-j) is dissipated uniformly along the flow path (2d+H-i-j) which runs down the back of the wall and up the front. The porewater pressure calculated in this manner results in pressures greater than hydrostatic in front of the wall and less than hydrostatic behind the wall (Figure 3-22).

Figure 3-21. Flow Net for a Retaining Wall (after Padfield and Mair, 1984). In Figure 3-22, the pore water pressure at the bottom of the wall, Uf, is equal on either side of the wall. The value for Uf is given by the following: Uf = FHWA NHI-07-071 Earth Retaining Structures

2 (d + h − j) (d − i) γw 2d + h − i − j 3-34

(3-33) 3 - Lateral Earth and Water Pressures June 2008

Figure 3-22. Gross And Net Water Pressures Across A Retaining Wall (modified after Padfield and Mair, 1984).

The net water pressure acting on the wall is shown on Figure 3-22b. The largest net water pressure occurs at the level of the water table within the excavation: U c = (h + i − j)

2 (d − i ) γw 2d + h − i − j

(3-34)

For comparison, the net water pressure for the condition in which there is no seepage is also shown on Figure 3-22b. In that case, the net pressure is given by: U n = (h + i − j) γ w

(3-35)

The effect of special drainage conditions on pore water pressures can only be assessed by using appropriate seepage flow nets. If, for example, the wall acts as a drain, the pore water pressures will vary significantly with distance behind the wall. The simplified methods shown in Figure 3-22 cannot be used for this case to calculate the pressures on the back of the wall. However, for normal designs, it is usually sufficient to use the simple flow net shown in Figure 3-21 and procedures to calculate pore water pressures shown in Figure 3-22. FHWA NHI-07-071 Earth Retaining Structures

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3.7

EARTH PRESSURE FROM SURCHARGE LOADS

3.7.1

General

Surcharge loads on the backfill surface near an earth retaining structure cause increased lateral earth pressures on the structure. Typical surcharge loadings may result from railroads, highways, sign/light structures, electric/telecommunications towers, buildings, construction equipment, and material stockpiles. The loading cases of particular interest in the determination of lateral earth pressures are: • • • •

uniform surcharge; point loads; line loads parallel to the wall; and strip loads parallel to the wall.

Figure 3-23 shows examples of retaining walls with surcharge loads. The surcharge computations provided herein are used to obtain unfactored surcharge loadings. Load factors used for various types of surcharge loads are provided in Chapter 4 and AASHTO (2007) Section 3.4.1. 3.7.2

Uniform Surcharge Loads

Surcharge loads are vertical loads applied at the ground surface, which are assumed to result in an assumed uniform increase in lateral stress over the entire height of the wall. The increase in lateral stress for uniform surcharge loading can be written as: ∆σ h = Kq s

(3-36)

where: ∆σh is the increase in lateral earth pressure due to the vertical surcharge load, qs is the vertical surcharge stress applied at the ground surface, and K is an appropriate earth pressure coefficient. Examples of surcharge loads for highway wall system applications include: (1) dead load surcharges such as that resulting from the weight of a bridge approach slab or concrete pavement; (2) live load surcharges such as that due to traffic loadings; and (3) surcharges due to equipment or material storage during construction of the wall system. When traffic is expected to come to within a distance from the wall face equivalent to onehalf the wall height, the wall should be designed for a live load surcharge. For temporary walls that are not considered critical, actual surcharge loads may be evaluated and considered FHWA NHI-07-071 Earth Retaining Structures

3-36


in the design as compared to using this prescriptive value. Both temporary and permanent wall designs should account for unusual surcharges such as large material stockpiles and heavy cranes. Calculated lateral pressures resulting from these surcharges should be added explicitly to the design earth pressure envelope. Loads from existing buildings need to be considered if they are within a horizontal distance from the wall equal to the wall height. 3.7.3 Point, Line, and Strip Loads

Point loads, line loads, and strip loads are vertical surface loadings which are applied over limited areas as compared to surcharge loads. As a result, the increase in lateral earth pressure used for wall system design is not constant with depth as is the case for uniform surcharge loadings. These loadings are typically calculated using equations based on elasticity theory for lateral stress distribution with depth (Figure 3-24 and AASHTO (2007) Section 3.11.6.1 and 3.11.6.2). For flexible walls, the pressure distributions provided in AASHTO (2007) Section 3.11.6.3 can be used.

3.8

EARTH PRESSURES DUE TO COMPACTION

In walls supporting fill, compaction of the backfill can generate relatively high horizontal pressures on the upper portion of the wall. It is customary, therefore, to reduce the compaction criteria for the zone immediately behind the wall. However, if a relatively high level of compaction is needed in that zone to reduce future ground deformations, the horizontal stresses generated by that compaction should be taken into account in the wall design. The value and distribution of the compaction stresses are dependent on the compaction equipment, procedures and applied energy, as well as, on the type and rigidity of the wall. When compaction equipment moves over the backfill adjacent to a wall, it induces added earth pressures on the wall. When the compaction equipment moves away, a portion of the added earth pressure continues to act on the wall due to the inelastic behavior of the soil. These residual horizontal pressures in a compacted soil mass can be considerably larger than the horizontal pressures in an uncompacted mass. A typical distribution of the residual horizontal pressures due to the compaction process is shown in Figure 3-25. The method proposed by Clough and Duncan (1991) was used to develop this figure. For this example, residual earth pressure increases sharply with depth in the upper 5 ft, but the rate of increase is reduced at greater depths. At depths below 25 ft, there is no residual earth pressure due to compaction, i.e., below 25 ft, the earth pressure is equal to the normal earth pressure at-rest. FHWA NHI-07-071 Earth Retaining Structures

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(a)

(b) Figure 3-23. (a) Retaining Wall with Uniform Surcharge Load and (b) Retaining Wall with Line Loads (see Railway Tracks) and Point Loads (see Catenary Structure). FHWA NHI-07-071 Earth Retaining Structures

3-38


Figure 3-24. Lateral Pressure Due to Surcharge Loadings (after USS Steel Sheet-Pile Manual, 1975). FHWA NHI-07-071 Earth Retaining Structures

3-39


EARTH PRESSURE AFTER COMPACTION (psf) 0

400

800

1200

1600

2000

st - re At

5

0

r th ea o

.4 =0

DEPTH (ft)

–K

15

ure ss pre

10

20

25

30

35

Roller Load = 500 lb/in. Roller 6 in. from wall Lift Thickness = 6 in. γ = 125 pcf φ= 35º c=0

Figure 3-25. Typical Residual Earth Pressure after Compaction of Backfill Behind an Unyielding Wall (after Clough and Duncan, 1991) It should be noted that the post compaction earth pressures estimated using, for example, the Clough and Duncan method, apply to conditions where the wall is stiff and unyielding. These pressures would provide a conservative estimate of pressures on flexible walls or massive walls whose foundation conditions allow them to shift laterally or tilt away from the backfill during compaction since such movements would reduce the earth pressures. The reduction would be expected to be less near the surface, where the compaction-induced loads tend to “follow” the wall as it is deflected or yielded (Clough and Duncan, 1991). Also, for MSE walls, compaction-induced pressures are already included in the design model (AASHTO, 2007).

3.9

EARTH PRESSURES RESULTING FROM SILO EFFECT

Occasionally, retaining walls are built in front of a stable rock face or an existing wall, and granular fill is placed between the new wall and the existing structure or the rock face (Figure 3-26). Due to the proximity of the two structures, the fill is partially supported by friction FHWA NHI-07-071 Earth Retaining Structures

3-40


between the soil and the confining “walls” similar to the way grain or other granular materials act in a silo. Consequently, the vertical stress in the fill and the horizontal stress acting on the wall are reduced. The lateral pressures in such wall configurations are commonly referred to as “silo” pressures. The design of these walls using typical active earth pressures based on an assumed planar failure surface may, therefore, be overconservative particularly for cases where the ratio of wall height to width of the fill zone is large. Figure 3-26 provides a procedure for estimating the lateral “silo” pressures on retaining walls.

3.10

EARTH PRESSURES RESULTING FROM SEISMIC FORCES

3.10.1 General

The seismic performance of earth retaining systems is most commonly evaluated using pseudo-static analysis, where the dynamic lateral earth pressure is estimated as a sum of the static lateral earth pressure and the increment in active pressure/passive pressure due to the seismic loading. For some cases, in particular where design ground accelerations are relatively high, displacement-based approaches are used. In this section, basic elements of seismic design of earth retaining systems are presented. Additional information on seismic design analyses for retaining walls can be found in Section 9.6 of GEC No. 3 (Kavazanjian et al., 1997). Specific LRFD-based information for seismic design is provided in NCHRP 20-07 (anticipated to be completed in October 2007)/ 3.10.2 Seismic Earth Pressures

The most commonly used method for seismic design of retaining structures is the pseudostatic method developed by Okabe (1926) and Mononobe (1929). The so-called Mononobe-Okabe (M-O) method is based on Coulomb earth pressure theory. In developing their method, Mononobe and Okabe assumed the following: •

the wall is free to move sufficiently to induce active earth pressure conditions;

•

the backfill is completely drained and cohesionless; and

•

the effect of the earthquake ground motion is represented by a horizontal pseudo-static inertia force, khWs and a vertical pseudo-static inertia force kvWs if the vertical force acts upward, or -kvWs, if the vertical force acts downward.


3-41


Figure 3-26. Estimation of “Silo” Pressures. FHWA NHI-07-071 Earth Retaining Structures

3-42


The limitations previously discussed for the Coulomb method also apply to the M-O method. One of the most significant limitations from a practical standpoint is the assumption that the soil is drained and cohesionless. While this situation can occur for fill walls, it is often not the case for cut walls where the natural geology includes some amount of cohesion. Even small amounts of cohesion reduce the seismic active earth pressure. Generalized limit equilibrium methods identified in the NCHRP 12-70 project can be used to quantify this reduction. Figure 3-27 shows the M-O representation for the case where soils behind the retaining wall are drained and cohesionless. In Figure 3-27, Ws is the weight of the sliding wedge and kh and kv are the horizontal and vertical seismic coefficients, respectively. The seismic coefficient kh and kv are expressed as a fraction of the acceleration of gravity g.

Figure 3-27. Seismic Forces Behind a Gravity Wall. Using M-O theory, the dynamic earth pressures in the active (PAE) and passive (PPE) state are given by the following:


PAE = 1 K AEγH 2 (1 − k v ) 2

(3-37)

PPE = 1 K PE γH 2 (1 − k v ) 2

(3-38)

3-43


K AE

cos 2 (φ − θ − β) = cos θ cos 2 β cos(β + δ + θ)D

  sin(φ + δ) sin(φ − θ − i)  12  D = 1 +      cos(δ + β + θ) cos(i − β)    

K PE =

(3-39)

2

cos 2 (φ − θ + β) cos θ cos 2 β cos(δ − β + θ)D′

  sin(φ + δ) sin(φ + i − θ)  12  D′ = 1 +      cos(δ − β + θ) cos(i − β)    

(3-41) 2

θ = tan-1 (kh/(1-kv)) where: γ H φ δ i β kh kv g

= = = = = = = = =

(3-40)

(3-42) (3-43)

effective unit weight of the backfill; height of the wall; angle of internal friction of the backfill; angle of friction of the wall/backfill interface; slope of the surface of the backfill; slope of the back of the wall; horizontal seismic coefficient expressed as a fraction of g; vertical seismic coefficient expressed as a fraction of g; and acceleration of gravity.

Figure 3-28 presents values for KAE for values of φ from 20 to 45 degrees for vertical walls with level backfill. The figure was derived for a wall/backfill interface friction angle δ=φ/2. The horizontal and vertical seismic coefficients (i.e., kh and kv) vary from 0 to 0.5 and from 0 to 0.2, respectively. The major challenges in applying the M-O theory are the selection of an appropriate seismic coefficient to determine the magnitude of the seismic earth pressure and the distribution of earth pressure or location of the seismic earth pressure resultant. For critical or rigid walls which cannot accommodate any deformation and for partially restrained abutments and walls restrained against lateral movements by, for example, batter piles, use of peak ground acceleration divided by the acceleration of gravity as the seismic coefficient may be warranted.


3-44


Seismic Active Pressure Coefficient (kAE)

1.0

0.8

= 20

o o

25

0.6

o

k v = 0.2 0.0 -0.2

30 o 35 o 40 o 45

0.4

0.2

o

= 35 =i=0 = /2

= i = kv= 0 = /2 0 0

0.1

0.2

0.3

0.4

0.5 0

0.1

0.2

0.3

0.4

0.5

Horizontal Seismic Coefficient (kh )

Figure 3-28. Effects of seismic coefficients and friction angle on seismic active pressure coefficient (after Lam and Martin, 1986). As noted in GEC No. 3 (Kavazanjian et al., 1997), use of a seismic coefficient from between one-half to two-thirds of the peak horizontal ground acceleration divided by gravity would appear to provide a wall design that will limit deformations in the design earthquake to small values acceptable for highway facilities. Evaluations being performed as part of NCHRP 1270 may recommend values even less than one-half for systems that can tolerate approximately up to 6 in. of movement. Similar to slope stability analyses, the vertical acceleration is usually ignored in practice in the design of earth retaining systems. Reference, however, should be made to applicable codes and local regulations which may provide specific recommendations for seismic coefficient values. The total seismic active earth pressure may be assumed to be uniformly distributed over the height of the wall, meaning that the earth pressure resultant acts at the mid-height of the wall. Therefore, place the resultant active earth pressure calculated using the M-O equations at mid-height of wall for design analysis. An alternate method of estimating seismic earth pressures has been suggested (see NCHRP 12-70) for cases where layered soil profiles exist. This method involves using a slope stability program. The total seismic earth pressure force is obtained by applying an external force to a vertical plane at the heel of the retaining wall. The external resisting force is FHWA NHI-07-071 Earth Retaining Structures

3-45


varied until a factor of safety of 1.0 is achieved during the pseudo-static seismic loading. The advantage of this approach is that it can accommodate layered soil profiles with cohesive soil content, various groundwater conditions, and non-uniform slopes. Undrained soil strengths would normally be used for these analyses to represent the short-term loading response that occurs during a seismic event. 3.10.3 Displacement Approach

To circumvent the shortcomings associated with selection of an appropriate seismic coefficient, deformation-based analyses have been developed for design of gravity retaining walls. Moreover, use of the M-O analysis may lead to overly conservative designs where design ground accelerations are relatively large. Several theories are available to account for the displacement and rotation of walls during an earthquake. Richards and Elms (1979) extended the work of Franklin and Chang (1977) on seismic deformation of earth dams to gravity retaining walls. Richards and Elms proposed the following simplified formula for the displacement of a gravity wall. d = 0.087 (V 2/A ⋅ g) ⋅ (N/A)-4

(3-44)

where d is the displacement in inches, V is the peak velocity of the earthquake record in in./s, N is the peak seismic resistance coefficient sustainable by the wall before it slides (equal to the yield acceleration of the retaining wall divided by gravity), and A is the maximum acceleration of the earthquake record or the maximum design acceleration value from a seismic hazard analysis. The Richards and Elms approach requires evaluation of the yield acceleration for the wall-backfill system. The yield acceleration is the level of acceleration that is just large enough to cause the wall to slide. A method to evaluate the yield acceleration for use in the Richards and Elms approach is provided in Kramer (1996). The NCHRP 12-70 project has updated the Richard and Elms approach based on a reevaluation of a large database of earthquake records and presented charts for estimating displacement for N/A ratios. These charts result in lower displacement estimates than obtained by Richard and Elms, particularly at lower N/A ratios. The NCHRP 12-70 project also developed a simple relationship between PGV and spectral acceleration at one second [PGV (in./sec) = 55 S1]. For highly seismic regions, where the methods discussed in this section result in wall systems which are conservative (e.g., very high anchor forces or excessively large-sized gravity structures), or where complex wall systems are used (e.g., high walls, stacked walls, etc.), dynamic stress-deformation (i.e., finite element or finite difference) analyses which account FHWA NHI-07-071 Earth Retaining Structures

3-46


for the nonlinear behavior of soil and soil-wall structure interfaces should be performed to evaluate earthquake-induced wall deformations. Several efforts are ongoing concerning LRFD seismic design of bridges and bridge substructures (e.g., TRC/Imbsen, 2006) and should be consulted when performing seismic analysis and design.

3.11

LATERAL EARTH PRESSURES IN COHESIVE BACKFILLS

Most DOTs involved in the design and procurement of fill wall systems (e.g., MSE walls) have well-defined backfill material requirements. In general, specifications for wall backfill require high-quality, granular, relatively free-draining backfills. However, in some cases a poorer quality on-site backfill material may be used, especially for temporary systems. These poorer quality backfills are generally more fine-grained (i.e., cohesive) and not freedraining. Previously, methods to calculate earth pressures in clayey soils were described. Herein, specific cautions are provided regarding the use of cohesive backfill soils. Lateral pressures can be caused by the volume expansion of ice in fine grained soils such as fine sand, silt and clay. These lateral pressures are difficult to predict and may achieve relatively high values. Since structures are usually not designed to withstand frost generated stresses, provisions should be made so that frost related stresses will not develop behind the structure or be kept to a minimum. The use of one or more of the following measures may be necessary: • • • • •

Isolate the backfill from underground sources of water either by providing a permeable drainage system or an impervious barrier; Use pervious backfill and provide weep holes in the structure; Provide an impervious soil layer near the ground surface, and grade the ground behind the wall to drain surface water away from the wall; Use a high-quality, granular, relatively free-draining backfill for a horizontal width behind the wall facing equal to the design frost depth; or Use of a thermal barrier (i.e., Styrofoam insulation) at the wall face.

Expansive clays can cause very high lateral pressures on the back of a retaining structure and should therefore be avoided whenever possible. In the cases where expansive clays are present behind the wall, swelling pressures may be evaluated based on laboratory tests and the wall may be designed to withstand these swelling pressures. Alternatively, one of the following measures can be taken: FHWA NHI-07-071 Earth Retaining Structures

3-47


•

A granular filter material can be provided between the clay backfill and the back of the wall. This material will drain the groundwater away from the expansive soil and, at the same time act as a buffer zone between the expansive soil and the structure.

•

The expansive soil can be treated with lime to significantly reduce, or eliminate, its swelling potential, if the soil does not contain gypsum. Expansive soils that contain gypsum should not be treated with lime because the combination of minerals in expansive soils, gypsum, and water may lead to ettringite, which has a much higher swelling potential than the expansive soils that are not treated.

The following is noted by Duncan et al. (1990) concerning the use of clayey soils as backfill for fill wall applications: •

Clayey backfills generally have lower drained shear strength than cohesionless soils. This results in: (1) larger lateral earth pressures against the back for the wall; (2) lower frictional resistance along the reinforcement for MSE walls which employ frictional reinforcement; and (3) lower bearing value for MSE walls which employ passive reinforcement (e.g., bar mats)

•

Clayey backfills are more plastic and contain more fines than cohesionless soils. This results in: (1) poor drainage and the potential for the development of water pressures behind the wall; (2) the potential for freezing of retained water and development of ice pressures on the back of the wall; and (3) greater potential for corrosion of metallic reinforcements for MSE walls.

•

Clayey backfills have the potential to undergo creep deformations that can lead to higher earth pressures and greater wall face deformations than for soils that do not exhibit significant creep potential. Earth pressures used for design of gravity walls employing clayey backfills should be based on past performance and field experience, as wall design methods do not consider the effects of creep.

Despite these problems, silts and clays may be used as backfill soils provided suitable design procedures are employed (including conservative estimates of lateral earth pressures) and construction control measures are incorporated into the contract documents. When such soils are used as backfills, walls may need to be designed for pressures between active and at-rest conditions. For soils which are deemed to have high swell potential, an earth-pressure coefficient as high as 1.0 may be used for design (Canadian Foundation Engineering Manual, 1992). In all cases, water pressures and appropriate surcharge loads also need to be added to these earth pressures.


3-48


In general, any permanent fill wall system which incorporates silty or clayey backfills must have an appropriately designed subsurface and surface drainage system to minimize pore pressure build-up and soil saturation. Such wall systems should also include periodic measurements of wall face movements.

3.12

LOADS FROM JOINTED ROCK MASSES

Rock is not a continuum but a regulated discontinuum and most rock slope failures occur along discontinuites or zones of weakness (Bell, 1992). For structures retaining rock, it is important to know the possible failure modes of the specific rock masses. The actual rupture or sliding plane behind a retaining wall depends on the spatial orientation, frequency and distribution of the discontinuities, and the involved shear and interlock resistance to shear among them (Jumikis, 1983). Depending on the type and structure of the rock, the lateral loads may be different. For example, if the rock is extensively weathered, highly disintegrated, densely jointed, loosely fragmented or decomposed, it behaves more or less like soil; in such cases the theories discussed previously are suitable using a shear strength representative of the rock mass. For other cases, the rupture or sliding plane behind a wall must be determined in order to calculate the lateral pressure. Detailed procedures on determining the lateral loads from jointed rock masses can be found in Brandl (1992) and FHWA HI-99-007 (Wyllie and Mah, 1998). Rock properties that are necessary to calculate earth pressures are provided in Chapter 2 of this Manual and GEC No. 5 (Sabatini et al., 2002).

3.13

EXAMPLE PROBLEMS

Three example problems are provided in this section: • • •

Constructing a lateral pressure diagram (Example Problem 1); Constructing a lateral earth and water pressure diagram and pressure diagram based on MEFP (Example Problem 2); and Constructing a lateral pressure diagram that includes a surcharge due to a line load (Example Problem 3).


3-49


3.13.1 Example Problem 1

For the wall configuration shown in Figure 3-29, construct the lateral pressure diagram. Assume the face of the wall to be smooth (δ = 0, cw = 0).

6 ft

Φ = 30º γ = 114 pcf

12 ft

β = 10º

Φ = 30º γ = 121 pcf

Ground Water Table

Figure 3-29. Example Problem 1 Geometry and Soil Conditions. Solution

Using the Coulomb method (Figure 3-5) for φ = 30o, β = 10o, θ = 0, and δ = 0: Ka = 0.374 The pressures at various depths can then be calculated as shown in Table 3-3. Table 3-3. Lateral Earth and Hydrostatic Pressures at Various Depths for Example 1. Effective Lateral Earth Pressures z, ft

σ 'VO , psf

Lateral Pressure, pa, psf

0

0

' K aσ VO =0

6

(114)(6) = 684

' K aσ VO = 0.374(684) = 255.8

18

(684+(121-62.4)(12) = 1387

' K aσ VO = 0.374(13870) = 518.8

Hydrostatic Pressure, pw

z, ft

zw, ft

σ w = z wγ w , psf

Lateral Pressure, pw, psf

0 6

0 0

0 0

18

12

(12)(62.4) = 748.8

0 0 K wσ w = 1.0(748.8) = 748.8


3-50


Based on Table 3-3, the lateral pressure diagrams due to earth and water can be constructed as shown in Figure 3-30a and b, respectively.

6 ft

β = 10º Φ = 30º γ = 114 pcf

Ground Water Table

12 ft

255.8 psf

Φ = 30º

+

γ = 121 pcf

518.8 psf

(a)

748.8 psf

(b)

Figure 3-30. (a) Lateral Effective Earth Pressure Diagram and (b) Water Pressure Diagram.

The total earth pressure is the summation of both the lateral earth pressure and water pressures shown in Figure 3-30. 3.13.2 Example Problem 2

For the wall geometry shown in Example Problem 1 above, assume that the retained soil is a silty sand to sandy silt material with φ′ = 30 degrees and c′ = 200 psf. Construct the design lateral pressure diagram. Assume the face of the wall to be smooth (δ = 0, cw = 0). Solution

As for Example 1, Ka = 0.374. The active earth pressure is calculated at various depths according to Equation 3-29. Figure 3-31 shows the total lateral pressure diagram and the pressure diagram calculated based on the minimum equivalent fluid pressure (MEFP). For design, the maximum envelope may be used.


3-51


Pressure (psf) -400

-200

0

200

400

600

800

1000

1200

0 Pa u Pa + u MEFP (= 30z) 6

Depth (ft)

DESIGN PRESSURE ENVELOPE

12

18

Figure 3-31. Pressure Diagrams for Example Problem 2. 3.13.3 Example Problem 3

Construct the lateral pressure diagram due to a line load of 210 psf located 15 ft behind the top of a 30 ft high unyielding wall shown in Figure 3-32.

15 ft

Q1 = 210 psf

30 ft

Figure 3-32. Geometry of Example Problem 3.


3-52


Solution

Lateral pressures due to surcharge loadings are given in Figure 3-24. From this figure the following can be found: −

m = 15 / 30 = 0.5 > 0.4 −

For m > 0.4 , the lateral pressure is given by:   −   − 2  Q1   m n  Ph = 1.28   2  H    − 2 −2    m + n       −

For m = 0.5, Q1 = 210 psf and H = 30 ft, the lateral pressure is given by:   −   2  210   0.5 n  Ph = 1.28  2  30    2 −2     0.5 + n           −   0.25 n   Ph = 8.96 2  −   0.25 + n 2         where all terms are defined in Figure 3-24. The computations for lateral pressures at various depths using the above formula are shown in Table 3-4 below.


3-53


Table 3-4. Computation of Lateral Earth Pressures Due to Line Load n =z/H

Depth below top of wall (ft)

Ph (psf)

0 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8 0.9 1.0

0 3 6 9 12 15 18 21 24 27 30

0.00 3.31 5.33 5.81 5.33 4.48 3.61 2.86 2.26 1.79 1.43

Using the information in Table 3-4, the lateral earth pressure with depth due to the line load is shown in Figure 3-33. 0

Depth below ground surface, ft

5

10

15

20

25

30 0.00

1.00

2.00

3.00

4.00

5.00

6.00

Lateral Pressure, psf

Figure 3-33. Lateral Pressure with Depth Due to Line Load.


3-54


CHAPTER 4 LOAD RESISTANCE FACTOR DESIGN FOR EARTH RETAINING SYSTEMS 4.1

INTRODUCTION

FHWA and the AASHTO subcommittee on Bridges and Substructures established an October 1, 2007 deadline for full implementation of load resistance factor design (LRFD). For many years, engineers have designed walls for highway and other applications using allowable stress design (ASD) methods. In ASD, all uncertainty in applied loads and material resistance are combined in a factor of safety or allowable material stress. In LRFD, uncertainty in load and material resistance are accounted for separately. The uncertainty in load is represented by a load factor and the uncertainty in material resistance is represented by a resistance factor. Sources of uncertainty include: • • • • • • • •

Material dimensions and location; Material strength; Failure mode and prediction method; Long term material performance; Material weights; Prediction of potential transient loads; Load analysis and distribution methods; and General uncertainty associated with structure function.

Although the implementation of LRFD requires a change in design procedures for engineers accustomed to ASD, many advantages do exist. LRFD accounts for variability in both resistance and load and can provide more consistent levels of safety in the superstructure and substructure as both are designed using the same loads. Section 11 of the LRFD AASHTO (2007) Specification provides information on LRFD for earth retaining structures including conventional retaining walls, nongravity cantilevered walls, anchored walls, mechanically stabilized earth (MSE) walls, and prefabricated modular walls. Section 10.4 of AASHTO (2007) provides detailed information on the evaluation of soil and rock properties to be used for design. NHI Course 130082A titled, “LRFD for Highway Bridge Substructures and Earth Retaining Structures” contains comprehensive information on LRFD for wall systems. In this section, LRFD-based concepts and computations are summarized. Information provided in Sections 4.2, 4.3, and 4.4 has been taken largely from Samtani (2007). The LRFD process and FHWA NHI-07-071 Earth Retaining Structures

4-1

4 - LRFD for ERS June 2008

example calculations for individual wall types are provided in the remaining chapters of this document. Also, in remaining chapters, wherever instructive, comparisons between ASD and LRFD computations are provided to assist design engineers in understanding the potential differences between ASD and LRFD designs.

4.2

CONCEPT OF LIMIT STATES

The AASHTO-LRFD approach uses reliability (probability) theory to quantify the uncertainty in loads, Q, and resistances, R. In the AASHTO-LRFD framework, once the load factors, γ, are established by reliability theory, the factored loads are combined to create a maximum load effect. A specific resistance factor, φ, is then developed corresponding to the load combination(s) based on measured resistances and their computed variances from nominal resistances predicted by numerical models for resistance. Similar to the loads, the uncertainties in the resistances are quantified based on reliability (probability) theory. The load and resistance factors include a consideration of the differences between measured and nominal values of the loads and resistances, respectively. By using factored loads, γQ, and factored resistances, φR, a limit state, g, can be defined. A limit state is a condition beyond which a structural component ceases to satisfy the provisions for which it was designed. A limit state, defined by combinations of factored loads and factored resistances, can be expressed as g = φR – γQ ≥ 0. From practical considerations, an acceptable risk level is determined for each limit state, i.e., the probability that φR – γQ < 0, because otherwise the design for the case φR – γQ ≥ 0 (i.e., no failure) will be very expensive. Thus, safety considerations in LRFD are addressed through load and resistance factors derived on the basis of an acceptable level of risk or acceptable probability of failure. This process is in contrast to the traditional ASD approach where safety is achieved with a single factor of safety applied to the resistance to obtain an allowable stress (or load). It is important to note that when load and resistance factors are developed using the limit state concept, they are completely tied to each other and form a pair. In other words, neither the load nor the resistance factor can be changed unilaterally in the LRFD framework. These factors can be changed based on local practice or previous successful projects, as long as, if one factor is changed, reliability-based computations are conducted to establish the other factor.


4-2


4.3

COMMON LIMIT STATES IN ERS DESIGN

In the AASHTO-LRFD framework, there are four distinct limit states: (1) strength limit states; (2) serviceability limit states; (3) extreme event limit states; and (4) fatigue limit states. For most earth retaining system designs, strength or service limit states control the design. Extreme limit states associated with typical earth retaining system design include earthquake loadings and vehicle impact.

4.4

•

Strength limit states are limit states that pertain to structural safety and the loss of load-carrying capacity. These limit states may be reached through either geotechnical or structural failure. Evaluation of strength limit states is based on the inelastic behavior of the structure, which is accomplished by using increased or factored loads (i.e., γ > 1.0) and on modification of soil behavior, which is accomplished by using reduced or factored strengths (i.e., φ < 1.0). From a geotechnical viewpoint, strength limit states are reached when they involve the partial or total collapse of the structure due to sliding, bearing resistance failure, pullout of reinforcements, etc. For welldesigned structures, strength limit states have a low probability of occurrence.

•

Serviceability limit states are the limiting conditions affecting the function of the structure under expected service conditions. Serviceability limit states occur before collapse. These include conditions that may restrict the intended use of the structure, e.g., excessive total or differential settlement. Evaluation of serviceability limit states is usually performed by using expected service loads (i.e., load factors = 1.0), nominal strengths (i.e., resistance factors = 1.0), and elastic analyses. Compared to strength limit states, the serviceability limit states have a higher probability of occurrence, but, if exceeded, involve less danger of loss of life.

LOAD COMBINATIONS IN LIMIT STATES

Because there are many different types of loads, the manner in which the loads are combined to create a limit state has sometimes been unclear in the traditional use of ASD. For instance, it is unlikely that the largest values of the live loads, wind load, stream load, and earthquake load will occur at the same time. The AASHTO-LRFD provides a solution to this problem by specifying several load combinations with load factors based on probability of occurrence. This is based on the observation that when one load component reaches its greatest value, the other load components are often acting at their average values, i.e., the probability of two or more load components acting at their greatest values simultaneously is so remote that it is negligible. For this reason, the AASHTO LRFD approach considers several load FHWA NHI-07-071 Earth Retaining Structures

4-3


combinations within each limit state with the intent that each load combination creates a maximum load effect. In the AASHTO-LRFD framework, each combination of the loads within a given limit state has an equal probability of occurrence. Since each combination of load has an equal probability of occurrence, all possible applicable load combinations in all limit states should be considered in design. Based on experience with specific designs, the user may realize that certain limit states may not control, however, it is prudent to check all possible load combinations. As previously introduced, factored loads or combinations of factored loads are compared to the available factored resistance of the earth retaining system. AASHTO LRFD load combinations for bridge substructures and earth retaining systems are provided in Table 3.4.1-1 of AASHTO (2007) and Table 4-1. Table 4-2 provides the load factors for permanent loads commonly used for earth retaining systems (i.e., Table 3.4.1-2 of AASHTO (2007) provides complete table). In general, use minimum load factors if permanent loads increase stability and use maximum load factors if permanent loads reduce stability.

4.5

EVALUATION OF RESISTANCE FACTORS

Load and resistance factors have been “calibrated” in an effort to obtain a more uniform level of safety for different limit states and types of materials and to provide reasonably consistent designs to those previously based on ASD. The most rigorous method for developing and adjusting resistance factors requires availability of statistical data but in most cases calibration is done by “fitting” using past experience and judgment. The majority of resistance factors for earth retaining system design in the AASHTO (2007) Specifications were developed by calibration by fitting to ASD. Calibration by fitting merely involves using parameters (e.g., resistance factors) that would result in the same minimum permissible physical dimensions of a structure as by ASD. Calibration by fitting does not achieve more uniform margins of safety than the ASD procedures it replaces; however, it ensures that designs based on LRFD will not lead to radically different designs from ASD. Calibration by fitting with ASD can be used where this is insufficient statistical data to perform a more formal process of calibration. Figure 4-1 depicts the equations used to calibrate a resistance factor in LRFD to an equivalent FS value in ASD. Equations on the left-side of Figure 4-1 generally describe ASD in which a nominal resistance is reduced by a factor of safety and the result is compared to the sum of the loads. Equations on the right-side depict the basic formulation of LRFD. By combining these, the resulting equation for resistance factor, φ, is found. FHWA NHI-07-071 Earth Retaining Structures

4-4


Table 4-1. Load Combinations and Load Factors (AASHTO, 2007). LOAD COMBINATIONS

Use one of these at a time

DC DD

LL

DW

IM

EH

CE

EV

BR

TU

ES

PL

CR

EL γP γP γP γP γP

LS 1.75 1.35 1.35

WA 1.00 1.00 1.00 1.00 1.00

WS 1.40 0.40

WL 1.0

FR 1.00 1.00 1.00 1.00 1.00

SH 0.50/1.20 0.50/1.20 0.50/1.20 0.50/1.20 0.50/1.20

TG γTG γTG γTG

SE γSE γSE γSE

γTG

γP

γEQ

1.00

-

-

1.00

-

γP

0.50

1.00

-

-

1.00

1.00 1.00 1.00 1.00

1.00 1.30 0.80 -

1.00 1.00 1.00 1.00

0.30 0.70

1.00 -

-

0.75

-

-

-

LIMIT STATE STRENGTH I STRENGTH II STRENGTH III STRENGTH IV STRENGTH V EXTREME EVENT I EXTREME EVENT II SERVICE I SERVICE II SERVICE III SERVICE IV FATIGUE – LL, IM, & CE ONLY

DC: dead load of structural components DD: downdrag

EQ

IC

CT

CV

γSE

-

-

-

-

-

-

1.00

-

-

-

-

-

-

1.00

1.00

1.00

1.00 1.00 1.00 1.00

1.00/1.20 1.00/1.20 1.00/1.20 1.00/1.20

γTG

γSE

γTG -

γSE 1.00

-

-

-

-

-

-

-

-

-

-

-

-

LL: vehicular live load

WS: wind load on structure WL: wind on live load

DW: dead load of wearing surfaces and utilities EH: horizontal earth pressure load EV: vertical pressure from dead load of fill ES: earth surcharge load

IM: vehicular dynamic load allowance CE: vehicular centrifugal force BR: vehicular braking force PL: pedestrian live load

TG: temperature gradient SE: settlement

FR: friction

EQ: earthquake

TU: uniform temperature

IC: ice load

CR: creep

LS: live load surcharge

SH: shrinkage

CT: vehicular collision force CV: vessel collision force

EL: accumulated locked-in force effects

WA: water load and stream pressure

See AASHTO (2007) Section 3.3 for complete definition of loads


4-5


Table 4-2. Load Factors for Permanent Loads, γP (modified after AASHTO, 2007). Load Factor Maximum Minimum 1.25 0.90 1.50 0.90 1.50 0.65

Type of Load DC: Component and Attachments DC: Strength IV only DW: Wearing Surfaces and Utilities EH: Horizontal Earth Pressures Active At-Rest AEP(1) for Anchored Walls EV: Vertical Earth Pressure Overall Stability Retaining Walls and Abutments ES: Earth Surcharge (1)

1.50 1.35 1.35

0.90 0.90 N/A

1.00 1.35 1.50

N/A 1.00 0.75

AEP refers to apparent earth pressure envelope. Anchored wall design may use earth pressures computed from apparent earth pressure envelopes.

Rn ≥ ∑ Qi FS

∑ γ Q ≤ ϕR i

i

∑γ Q

Rn ≥ FS ∑ Qi

i

ϕ

ϕ=

i

n

≤ Rn

∑γ Q FS ∑ Q i

i

i

Figure 4-1. Equations used to Relate LRFD Resistance Factor to ASD FS (after Samtani, 2007).


4-6


For example, if loads consist of only dead loads (QDC) and live loads (QLL), the resistance factor can be expressed as:

ϕ=

γ DC QDC + γ LL QLL FS (QDC + QLL )

(4-1)

Dividing the numerator and denominator by live loads, QLL, results in:

ϕ=

γ DC (QDC / QLL ) + γ LL FS (QDC / QLL + 1)

(4-2)

Therefore, for different ratios of dead load to live load, resistance factors can be calibrated to the ASD FS value, as shown in Table 4-3. It should be noted that these equations simply provide a way to obtain an average value for the load factor considering the use of two different load factors from two different sources (i.e., dead load and live load), and considering the relative magnitudes of each load. In the most basic terms, the ASD FS is simply the average load factor divided by the resistance factor. There is no consideration of the actual bias or variability of the load or resistance prediction methods when “calibration” by fitting to ASD is used, nor is there any consideration of the probability of failure. All that is being done is to calculate the magnitude of a resistance factor, for a given set of load factors, that when combined with the load factors, provides the same magnitude of FS as is currently used in ASD. Therefore, whatever margin of safety was implied by the ASD FS, the load and resistance factor combination that results from this type of analysis will have the same unknown margin of safety. Without conducting some type of reliability analysis based on statistical data for the loads and resistances under consideration, or without some type of quantification of the number of failures relative to the number of successes when the factor of safety in question is used in conjunction with a given design procedure, the margin of safety implied by the factor of safety, and the resistance factors derived using this approach, will be unknown (see Allen, 2005). Throughout this manual, the resistance factors used for various design checks for the wall types discussed are provided. In general, unless otherwise specified, resistance factors for design checks at Extreme Event limit states are equal to 1.00.


4-7


Table 4-3. Values of Resistance Factors Corresponding to Different Values of Factor of Safety and Dead to Live Load Ratios for γDC = 1.25 and γLL = 1.75. QDC/QLL 1.0

2.0

3.0

4.0

1.00 0.75 0.50 0.38

0.94 0.71 0.47 0.35

0.92 0.69 0.46 0.34

0.90 0.68 0.45 0.34

FS 1.5 2.0 3.0 4.0


4-8


CHAPTER 5 CAST-IN-PLACE (CIP) GRAVITY AND SEMI-GRAVITY WALLS 5.1

INTRODUCTION

The design and construction of Cast-In-Place (CIP) concrete gravity walls (also sometimes identified as “rigid gravity walls” or “mass gravity walls”) and CIP concrete semi-gravity walls (also identified as “conventional CIP walls”, or “concrete cantilever walls”) are discussed in this chapter. CIP gravity walls have been used as low-height retaining walls for both roadway cut and fill applications. CIP gravity walls are generally trapezoidal in shape (Figure 5-1) and are generally constructed of unreinforced or minimally reinforced mass concrete. This type of wall is also occasionally constructed of stone masonry. CIP gravity walls are rigid type walls that rely entirely on their self-weight to resist overturning and sliding, and are generally proportioned to avoid any tensile stresses within the structure. CIP semi-gravity walls are commonly used for earth retaining structures and bridge abutments in fill situations. They can also be used in cut situations, but for such applications a temporary support system is typically required. In addition to its own weight, this type of wall uses bending action to resist lateral forces on the wall. The CIP semi-gravity walls include the cantilever, counterfort, and buttress walls illustrated in Figure 5-2. At sites underlain by competent soils, the base of the CIP wall can be designed as a spread footing bearing directly on the foundation soils. At locations where foundation bearing resistance or settlement is a concern, the CIP wall can be provided with a deep foundation, using the wall base as the pile cap. Since there are substantial labor and material costs associated with CIP walls, they may not be economical in comparison with alternative types of walls, particularly for wall heights more than about 6 ft for CIP gravity walls or 15 ft for CIP semi-gravity walls. They also may not be economical for walls in cut applications because of the added costs associated with the construction of a temporary excavation support system. In evaluating the use of CIP walls, therefore, consideration should also be given to other wall alternatives as discussed in the following chapters of this manual.


5-1

5 - Gravity and Semi-Gravity Walls June 2008

Granular Soil Backfill H Mass Concrete

0.5H to 0.7H

Figure 5-1. Cast-In-Place Gravity Wall.

The primary advantages of CIP walls are: • • • • •

Control of all construction materials (e.g., cement, aggregate, reinforcing steel); Well-established design procedures and construction specifications; Compatibility with deep foundations; Minimal wall movements; and Resistance to degradation (corrosion).

However, the use of CIP walls has major limitations, including: • • •

• •

Construction is labor intensive; Relatively long construction time is required; A bottom-up sequence of construction is followed: o Temporary excavation support may be necessary; o Increased quantity of excavation; o Increased quantity of backfill; and o Wider work area required; this may be a significant problem at urban sites and on steep hillsides. Low tolerance to differential settlement; and Higher cost


5-2


Figure 5-2. Cast-In-Place (CIP) Concrete Retaining Walls and Terminology (a) Cantilever Wall (Bowles, 1988); (b) Counterfort Wall; and (c) Buttress Wall (Teng, 1962).

5.2

TYPES OF SEMI-GRAVITY CIP WALLS

5.2.1

Cantilever Walls

Cantilever walls are constructed in the form of an inverted T wherein the projecting members act as cantilever elements. Typical proportions of cantilever walls are shown in Figure 5-3. This type of wall is generally suitable for heights up to about 30 ft; higher walls may require an excessively dense arrangement of reinforcing bars at the base of the stem wall due to large lateral forces imposed by higher retained fill. Figure 5-4 shows an example of a CIP cantilever retaining wall.


5-3


8" min (12" preferable) Min. Batter (1H:48V) H

H/ 10

H/ 12

to H/8

varies

to H/10 H/ 12

2/ H 5

to H/10

to 3/5H

Note: The footing level should be below depth of seasonal volume change and frost line.

Figure 5-3. Common Proportions of Cantilever Walls (after Teng, 1962).

Figure 5-4. CIP Cantilever Retaining Wall.


5-4


5.2.2

Counterfort Walls

Counterfort walls can be used for structures higher than about 30 ft. This type of wall is a variation of the cantilever wall wherein both the base slab and wall face span horizontally between vertical brackets known as counterforts. The counterforts increase wall stability and reduce wall stem bending moments allowing the stem thickness to be reduced without excessive outward deflection. The proportions of counterfort walls, shown in Figure 5-5, vary to a greater extent than those for cantilever walls because the thicknesses of the face and base slab depend primarily on the spacing of the counterforts. For walls with a height of about 30 ft, the counterforts may be spaced as far as two-thirds of the height of the wall. As wall heights increase, the spacing of the counterforts is reduced, and may be as little as one-third of the wall height. For constructability considerations, counterforts should not be placed on spacing less than about 8 ft. The toe projection is generally smaller than that for cantilever walls. Figure 5-6 shows the construction of a counterfort retaining wall used in a cut application.

8 in min (12 in preferable)

8 in min

28 in min

Figure 5-5. Common Proportions of Counterfort Walls (after Teng, 1962).


5-5


Figure 5-6. CIP Counterfort Wall Construction in a Cut Application. 5.2.3

Buttress Walls

The buttressed wall illustrated in Figure 5-2(c) is another type of CIP semi-gravity wall. It is similar to the counterfort wall except that the vertical brackets (known as buttresses) are on the outside of the stem wall, and act in compression rather than tension. This type of wall is not commonly used for retaining walls because of the exposed buttresses. 5.2.4

Other CIP Semi-Gravity Walls

Other types of CIP semi-gravity walls which may be used for special applications are illustrated in Figure 5-7. The cantilever walls shown in Figures 5-7(a) and (b) could be considered for cut and fill applications, respectively, at sites with tight right-of-way restrictions. The U-walls shown in Figure 5-7(c) and Figure 5-8 are often used for construction of depressed roadways, particularly where the clear distance between the walls is less than about 30 ft, and for depressed roadways constructed below the groundwater level. Unless supported on a deep foundation, the cantilever walls shown in Figures 5-7(a) and (b) are structurally less efficient than the conventional cantilever wall shown in Figure 5-2(a), and typically would be limited to a lower height than a conventional cantilever wall. The maximum height of a U-wall, however, would be comparable to that of a conventional cantilever wall. FHWA NHI-07-071 Earth Retaining Structures

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Figure 5-7. Other Types of Cast-In-Place (CIP) Walls: (a) Fill Wall with Limited ROW, (b) Cut Wall with Limited ROW, and (c) U-Wall for Depressed Roadway.


5-7


Figure 5-8. CIP U-Wall for Depressed Roadway.

5.3

WALL CONSTRUCTION

The typical sequence of construction for CIP concrete walls is presented below. •

The first stage of construction consists of excavating to the wall foundation grade and preparing the wall foundation subgrade. If stable excavation slopes cannot be maintained during the period of wall construction, temporary excavation support needs to be provided. Foundation preparation includes removing unsuitable materials such as organic matter and vegetation from the area to be occupied by the wall and leveling and proof-rolling the foundation area. Since CIP walls cannot tolerate significant differential settlement, preparation of the wall foundation to provide a relatively stiff and uniform bearing surface is particularly important. For walls founded on compressible soils, foundation preparation may require ground improvement to increase bearing resistance and stiffness, or the construction of deep foundations for wall footing support. Where rock is encountered above the wall foundation grade, the rock will typically be removed to some nominal depth below the foundation grade (e.g., 6 in.) and backfilled with compacted granular material.


5-8


5.4

•

For CIP semi-gravity walls, the footing outline is formed and the reinforcing steel for the footing is placed and extended into the wall stem. For a counterfort (or buttress) wall, reinforcing steel is also extended into the counterforts (or buttresses) (Figure 56). The footing concrete is then poured.

•

The wall stem is formed, weep hole inserts placed, and concrete poured. For counterfort (or buttress) walls, reinforcing steel for the counterforts (or buttresses) is placed and the counterforts (or buttresses) are then formed and the concrete is poured. Typically, concrete is poured in sections between vertical expansion joints and, wherever possible, it is poured for the full wall height to eliminate cold joints.

•

Drainage systems are then constructed behind the wall. Concurrent with this activity is the placement and compaction of select backfill soil to finished grade. Care should be taken to ensure that the backfill soils are not overcompacted just behind the wall face. Overcompaction can induce large lateral earth pressures which may over stress the wall and, if prefabricated drainage material is installed against the back wall face, damage the drainage components.

COST

For CIP walls, costs for rigid gravity and semi-gravity walls are similar for wall heights less than 10 ft. A cost range for walls up to 10 ft is typically about $25 to $35/ft2 of wall face with the higher range associated with walls with surcharges or sloped fills behind them. As previously discussed, rigid gravity walls are not cost effective above 10 ft. For CIP cantilever walls greater than 10 ft and less than 20 ft high, costs are in the range of $20 to $35 /ft2 of wall face. For walls greater than 20 ft, costs will be greater than 35/ft2. It is noted that labor represents 30 to 40 percent of the unit cost for CIP walls so costs can be significantly affected by prevailing labor rates. The costs provided above for CIP walls do not include costs associated with temporary shoring and/or deep foundations. Costs for temporary shoring can be estimated based on information provided in Chapter 8 for sheet pile, soldier beam and lagging walls, and anchored wall systems by deducting the cost of permanent facing. As with most wall systems, unit costs will increase for high walls that are short in length due to lack of room for equipment to operate, where access is limited, and where ROW costs are significant.


5-9


5.5

WALL DESIGN

For the design of gravity and semi-gravity walls, external stability analyses are performed to evaluate the ability of the wall to resist lateral earth and water pressures. The possible modes of external stability failure include sliding, bearing resistance, and limiting eccentricity as illustrated in Figure 5-9.

Sliding

Limiting Eccentricity (Failure due to Overturning)

Bearing Figure 5-9. Strength Limit States for Rigid Gravity and Semi-Gravity Walls. Table 5-1 summarizes the major design steps for CIP gravity and semi-gravity walls. 5.5.1 Steps 1, 2, and 3 Herein, it is assumed that Step 1 has been completed and a CIP wall has been deemed appropriate for the project and Steps 2 and 3 have been completed to establish soil and/or rock parameters for design. In general, the required parameters for in situ soil and rock are the same as those required in support of the design of a spread footing, in particular foundation shear strength (to allow for bearing resistance evaluation) and compression parameters of the foundation materials (to allow for wall settlement computations). For gravity walls which require deep foundation support, the required parameters are the same as those required for the design of a driven pile or drilled shaft foundation.


5-10


Table 5-1. Design Steps for Gravity and Semi-gravity Walls. Step 1.

Establish project requirements including all geometry, external loading conditions (transient and/or permanent, seismic, etc.), performance criteria, and construction constraints.

Step 2.

Evaluate site subsurface conditions and relevant properties of in situ soil and rock parameters and wall backfill parameters.

Step 3.

Evaluate soil and rock parameters for design and establish resistance factors.

Step 4.

Select initial base dimension of wall for strength limit state (external stability) evaluation.

Step 5.

Select lateral earth pressure distribution. Evaluate water, surcharge, compaction, and seismic pressures.

Step 6.

Evaluate factored loads for all appropriate loading groups and limit states.

Step 7.

Evaluate bearing resistance.

Step 8.

Check eccentricity.

Step 9.

Check sliding.

Step 10.

Check overall stability at the service limit state and revise wall design if necessary.

Step 11.

Estimate maximum lateral wall movement, tilt, and wall settlement at the service limit state. Revise design if necessary.

Step 12.

Design wall drainage systems.

The drainage and shear strength characteristics of the wall backfill soil are assessed as part of Step 3. Guidelines for wall backfill material gradation and drainage behind gravity retaining walls can be found in the AASHTO (2002). Whenever possible, the backfill material should be free draining, nonexpansive, and noncorrosive. Silts and clays should not be used for backfill unless suitable design assumptions (e.g., possibility of saturation of backfill) are followed and specific measures are taken to account for their presence (e.g., robust surface drainage system). All backfill material should be free of organic material. Backfill gradation should follow the guidelines presented in Table 5-2.


5-11


Table 5-2. Suggested Gradation for Backfill for Cantilever Semi-Gravity and Gravity Retaining Walls Sieve Size 3 in. No. 4 No. 30 No. 200 5.5.2

Percent Passing 100 35 – 100 20 – 100 0 – 15

Step 4 – Select Base Dimension

Figures 5-3 and 5-5 show typical dimensions for a semi-gravity cantilever retaining wall and for a counterfort wall. These dimensions have been developed based on a range of backfill properties, geometries, and stable foundation soils. These typical dimensions can be used for preliminary design, however the final strength limit state calculations are made given the geometry requirements of the project (i.e. limited right-of-way) and specific conditions. 5.5.3

Step 5 – Select Lateral Earth Pressure Distribution

Lateral earth pressures for design of CIP walls are determined using the procedures presented in Chapter 3. Generally, Coulomb theory is used to compute earth pressures either directly on the back face of the wall (gravity wall case) or on a vertical plane passing through the heel of the base slab (semi-gravity wall case). The following should be considered in evaluating earth pressure loading for a CIP wall: •

•

• • •

Use at-rest earth pressures for rigid gravity retaining walls resting on rock or batter piles or for unyielding walls such as culverts, tunnels and rigid abutment U-walls (such as shown in Figure 5-10), where wall rotation and displacement are restrained. Use the average of at-rest and active earth pressures for CIP semi-gravity walls founded on rock or restrained from lateral movements (e.g., use of batter piles) and which are less than 16 ft in height. Use active earth pressures for CIP semi-gravity walls greater than 16 ft in height. Earth pressures due to compaction, water, and surcharges should be evaluated using the procedures described in Chapter 3. Passive resistance in front of the wall should not be used in the analyses unless the wall extends well below the depth of frost penetration, scour or other types of disturbance (e.g. utility trench excavation in front of the wall).


5-12


Figure 5-10. CIP Abutment with Integral Wingwalls. If adequate drainage measures are provided, the hydrostatic pressure due to groundwater behind the wall generally need not be considered; however, hydrostatic pressure must be considered for portions of wall below the level of the weep holes unless a deeper drainage pipe is provided behind the base of the wall. When it is necessary to maintain the groundwater level behind the wall, the wall must be designed for the full hydrostatic pressure. 5.5.4

Step 6 – Evaluate Factored Loads

Stability computations for gravity and semi-gravity retaining walls are made assuming that a triangular earth pressure distribution (plus additional pressures resulting from surcharges, compaction, water, or seismic forces) develops on a vertical or inclined pressure plane rising from the heel of the wall. Load combinations should be selected to obtain the most realistic “extreme load effects”. For example, as shown in Figure 5-11, the load factors applied for the evaluation of sliding resistance of the wall would be 0.90 DC, and 1.0 EV. These forces result in an increase in the contact stress and resistance of the wall base to sliding and therefore minimum load factors are used. Comparatively, as shown in Figure 5-12, the critical load combination and load factors applied for the evaluation of bearing resistance would be 1.25DC and 1.35EV since the use of maximum load factors results in the greatest value for bearing pressure which is a destabilizing force.


5-13


Figure 5-11. Typical Application of Load Factors for Eccentricity and Sliding.

Figure 5-12. Typical Application of Load Factors for Bearing Resistance.

A live load surcharge should be applied where vehicular load is expected to act on the surface of the backfill within a distance equal to one-half the wall height behind the back of the wall face of the wall. For this, the “back of the wall face” should be considered to be the pressure surface being considered for the computation. Where live load surcharge is applicable, the factored surcharge force is generally included over the backfill immediately above the wall only for evaluation of foundation bearing FHWA NHI-07-071 Earth Retaining Structures

5-14


resistance and structure design. The live load surcharge is not included over the backfill for evaluation of eccentricity, sliding, or other failure mechanisms for which such a surcharge would represent added resistance to failure. The load factor for live load surcharge (e.g., 1.75 for Strength I, 1.35 for Strength II, 1.0 for Service I, etc. (see Table 3.4.1-1 in AASHTO, 2007)) is the same for both vertical and horizontal load effects. When dealing with live load surcharges, it may be simplest to perform limit state checks for cases which include the live load surcharge and for cases which exclude the live load surcharge and design based on more critical result. Note that once a particular load is assigned a load factor, it retains that load factor. For example, the vertical component of load EH is assigned a maximum load factor (1.5) even though it acts to resist the overturning and sliding. It is stressed that the above figures are “typical” applications. The designer is responsible to identify the load and load factor combination that produces the maximum force effect. Loads from traffic barriers are considered at the Extreme II limit state. In this case, the load factor for the traffic load, CT, is 1.00. For a CIP wall, this loading is used for evaluating eccentricity, bearing resistance, and sliding of the wall. The vehicle collision load (CT) is an instantaneous load applied in the same direction as LS and EH. Section A13.2 of AASHTO (2007) describes the procedure used to evaluate the vehicle collision loading and geometric requirements for a traffic railing. A yield line analysis is used to check whether a particular rail or barrier type is applicable to resist the loading. For example, for the CIP wall shown in Figure 5-13, assume that structural calculations (i.e., yield line analysis) have shown that a proposed rail meets or exceeds requirements for a TL-4 rail test level (see Table A13.2-1 of AASHTO, 2007). The evaluation of sliding and eccentricity would consider a horizontal load of 54 kips located at a distance of 32 in. (2’-8”) (the effective height of the vehicle rollover force) above the base of the rail (see Figure 5-14). In the analysis, it is assumed that the 54 kip load acts over a longitudinal distance of 5 ft. The load is then assumed to reduce using a 45 degree distribution angle to the elevation of the bottom of the footing. The resulting 2.44 kip/ft width of wall force would then be assumed to be applied at 17.17 ft above the base of the footing (i.e., 1.5 ft + 13 ft + 2.67 ft) for sliding and eccentricity calculations.


5-15


1’6’’ 2’8’’

Barrier Load = 54 kips

1/2

14’6’’

2’0’’

4’6’’

2’9’’

13’0’’

12

1’6’’ 9’3’’

Figure 5-13. Example CIP Wall with Traffic Barrier. 5’0’’

2’8’’

13’0’’

1’6’’

BARRIER LOAD DISTRIBUTION WIDTH

ASSUMED DISTRIBUTION TO RETAINING WALL END OF RETAINING WALL PANEL

1 1

22’2’’ LIMITED BY LENGTH OF RETAINING WALL PANEL

Barrier Load @ Bottom of Footing PCT = 54 kips / (5 ft + 2.67 ft + 13 ft + 1.5 ft) = 2.44 kips/ ft width

Figure 5-14. Computation of Barrier Load at Footing Level. FHWA NHI-07-071 Earth Retaining Structures

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5.5.5

Step 7 - Evaluate Bearing Resistance

The computed vertical stress, σv, at the base of the wall footing must be checked against the factored bearing resistance of the soil. The generalized distribution of bearing vertical stress at the wall base is illustrated in Figures 5-15 and 5-16 for foundations on soil and rock, respectively. The procedures for evaluating the nominal bearing resistance of the foundation are found in Geotechnical Engineering Circular No. 6 (Kimmerling, 2002) and in AASHTO (2007) Section 10.6.3.1 and 10.6.3.2. Student Exercise 3 illustrates the LRFD evaluation of bearing resistance for a CIP wall. Bearing resistance is evaluated at the strength limit state using factored loads and resistances as illustrated in Figure 5-12. The live load surcharge is considered for bearing resistance. The nominal bearing resistance, qn, is calculated based on methods for shallow foundations, and uses the resistance factor from Table 10.5.5.2.1-1 of AASHTO (2007). For walls on shallow foundations, a resistance factor of 0.55 is used. This resistance factor has been calibrated to a ASD factor of safety value of 2.5.

q r = φq n ≥

∑V B − 2e

(5-1)

For walls on soil foundations, the vertical stress, σv, is calculated assuming a uniform distribution of pressure over an effective base width, B′ where B′ = B - 2 e. σv =

∑V

B − 2e

(5-2)

where e is defined in Figure 5-15 and V is the summation of all factored vertical forces (including weight of wall stem, wall base, and soil above the heel and vertical component of earth pressure loading (if applicable)). If the computed e is less than zero, assume e = 0. Where deep foundation elements are used, resistance is calculated using methods for resistance computations for driven piles (see AASHTO (2007) Section 10.7) and drilled shafts (see AASHTO (2007) Section 10.8).


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Figure 5-15. Loading and Eccentricity for Walls Founded on Soil. For walls founded on rock, the vertical stress shall be calculated assuming a linearly distributed pressure over an effective base area as shown in Figure 5-16. If the resultant vertical force is within the middle one-third of the wall base, σ v max =

∑ V 1+ 6 e 

(5-3)

σ v min =

∑ V 1− 6 e 

(5-4)

B 

B 

B

B

If the resultant is outside the middle one-third of the wall base, then σ v max =


2∑V

3[(B / 2) − e]

(5-5)

σ v min = 0

(5-6)

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Figure 5-16. Loading and Eccentricity for Walls Founded on Rock.

5.5.6

Step 8 – Check Eccentricity

The criteria for evaluating eccentricity in ASD for CIP walls requires that the resultant force be maintained within B/6 of the foundation centroid for foundations on soil, and within B/4 of the centroid for foundations on rock. For LRFD, these criteria were revised to reflect the factoring of loads and therefore the eccentricity limits have been increased compared to ASD. As a result, the location of the resultant of the resultant force should be within B/4 of the foundation centroid for foundations on soil, and within 3B/8 of the foundation centroid for foundations on rock. These limits were developed by direct calibration with ASD and, along with the check for bearing resistance, replace the check on overturning previously used for ASD designs. If there is inadequate resistance to overturning (eccentricity value greater than limits given above), consideration should be given to either increasing the width of the wall base, or providing a deep foundation. FHWA NHI-07-071 Earth Retaining Structures

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5.5.7 Step 9 – Check Sliding

Sliding resistance along the base of the wall is evaluated using the same procedures as for spread footing design. The driving forces in a sliding evaluation will generally included factored horizontal loads due to earth, water, seismic, and surcharge pressures and the resisting force is provided by the shear resistance between the foundation base and foundation soil. Student Exercise 3 includes calculations for evaluating sliding. The factored resistance against sliding can be written as: R R = ϕ Rn = ϕτ Rτ + ϕ ep Rep

(5-7)

where Rn = nominal sliding resistance against failure by sliding ϕτ = resistance factor for shear resistance between soil and foundation (provided in Table 10.5.5.2.2-1 of AASHTO (2007)) Rτ = nominal sliding resistance between soil and foundation ϕep = resistance factor for passive resistance (provided in Table 10.5.5.2.2-1 of AASHTO (2007)) Rep = nominal passive resistance of soil available throughout design life Note that any passive resistance provided by soil at the toe of the wall by embedment is typically ignored due to the potential for the soil to be removed through natural or manmade processes during the service life of the structure. If the soil beneath the footing is cohesionless, the nominal sliding resistance between soil and foundation is given as: Rτ = V tan δ

(5-8)

for which (with φf = the friction angle of the foundation soil and V = total vertical force): tan δ = tan φf (for concrete cast against soil) tan δ = 0.8 tan φf (for precast concrete footing) For cohesive soil foundations, the procedure provided in AASHTO (2007) Section 10.6.3.4 should be used. If adequate sliding resistance cannot be achieved, design modifications may include: (1) increasing the width of the wall base; (2) using an inclined wall base or battering the wall to FHWA NHI-07-071 Earth Retaining Structures

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decrease the horizontal load; (3) incorporating deep foundation support; (4) constructing a shear key; and (5) embedding the wall base to a sufficient depth so that passive resistance can be relied upon. If the wall is supported by rock, granular soils or stiff clay, a key may be installed below the foundation to provide additional resistance to sliding 5.5.8 Step 10 – Evaluate Overall Stability

Where retaining walls are underlain by weak soils (Figure 5-17), the overall stability of the soil mass must be checked with respect to the most critical failure surface. Analyses considering both circular and non-circular slip surfaces must be considered. Overall stability is evaluated at the Service I limit state with typical resistance factors of either 0.65 or 0.75. Since the overall stability is evaluated at Service I limit state it is possible to derive a straightforward relationship between the FS in the ASD approach and the φ in LRFD. This relationship is given by FS = 1/φ. Thus, a value of 0.75 and 0.65 mentioned above correspond to a FS value of 1.33 and 1.54, respectively. The LRFD specifications do not intend to increase the traditional safety factors of 1.30 and 1.50 to 1.33 and 1.54, respectively. For the LRFD specifications in general, resistance factors are rounded to the nearest 0.05 because not doing so would imply an accuracy to the resistance factors that is really not justifiable. Thus, in a practical sense, a resistance factor of 0.65 is intended to be the same as using a FS=1.50, for example, and rounding from 1.54 to 1.50 is appropriate. Similarly, rounding from 1.33 to 1.3 is appropriate. In practice, the resistance factor should be 0.65 if the wall supports a bridge foundation, or the foundation for a similarly important structure such as a noise wall, water tower/tank, pipeline or critical utility that cannot tolerate movement. On the other hand, a resistance factor of 0.75 may be more appropriate for a sign foundation, minor retaining wall, etc. If overall stability is found to be a problem, deep foundations or use of lightweight backfill may be considered, or measures can be taken to improve the shear strength of the weak soil stratum. Other wall types, such as an anchored soldier pile and lagging wall, or tangent or secant pile wall, should also be considered. 5.5.9 Step 11 – Estimate Maximum Lateral Wall Movement, Tilt, and Wall Settlement at the Service I Limit State

Foundation settlement and tilt of the wall can be computed using methods in Geotechnical Engineering Circular No. 6 (Kimmerling, 2002), FHWA NHI-06-088 (Samtani and Nowatzki, 2006), or AASHTO (2007) Section 10.6.2. CIP walls can generally accommodate FHWA NHI-07-071 Earth Retaining Structures

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Figure 5-17. Typical Modes of Global Stability (after Bowles, 1988).

a differential settlement of up to about 1/500 (ratio of differential settlement of two points along the wall to the horizontal distance between the points). In general, tolerable total settlements of CIP walls are limited to 1 in. (as a means to control differential settlement). If the computed settlement and tilt exceed acceptable limits, the wall dimensions can be modified to shift the resultant force closer to the center of the base and thereby reduce the load eccentricity and differential settlement. In some cases, use of lightweight backfill material may solve the problem. The use of deep foundations can also be considered. Unless CIP walls are provided with a deep foundation, a small amount of wall tilting should be anticipated. It is therefore advisable to provide the face of the wall with a small batter to compensate for the forward tilting (see Figure 5-3). In general then, an acceptable level of tilt should be no more than the front batter of the wall, however, less allowable tilt may be necessary to avoid overstressing the junction of the stem and wall footing. In unusual cases, where the foundation materials are stiffer or firmer at the toe of the base than at the heel, the resulting settlement may cause the wall to rotate backwards, towards the retained soil. Such wall movements could substantially increase the lateral pressures on the wall since the wall is now pushing against the soil similar to a passive pressure condition. This case can be avoided by reproportioning the wall, supporting the wall on a deep foundation, or treating the foundation soils. CIP walls founded on a deep foundation may be subject to potentially damaging ground and structure displacements at sites underlain by cohesive soils. This may occur if the weight of the backfill material exceeds the bearing resistance of the cohesive subsoils, causing plastic displacement of the ground beneath the retaining structure and heave of the ground surface in FHWA NHI-07-071 Earth Retaining Structures

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front of the wall. As a general rule, if the ratio of the weight of the embankment plus any surcharge to the cohesive strength (su) of the foundation is greater than 2.5, then progressive lateral movements of the retaining structure are likely to occur (Peck et al., 1974). As this ratio gets larger, the rate of movement will increase until failure occurs when the ratio is at about 5. For CIP walls founded on vertical piles or drilled shafts, this progressive ground movement would be reflected by an outward displacement of the wall. CIP walls founded on battered piles typically experience an outward displacement of the wall base and a backward tilt of the wall face (Figure 5-18).

Figure 5-18. Typical Movement of Pile Supported Cast-In-Place (CIP) Wall with Soft Foundation.

5.5.10 Step 12 – Design Wall Drainage Systems

Water can have detrimental effects on earth retaining structures. Subsurface water and surface water can cause damage during construction and/or on a constructed structure. Control of water is a key component of the design of earth retaining structures. A subsurface drainage system serves to prevent the accumulation of destabilizing hydrostatic pressures, which may develop as a result of groundwater seepage and/or infiltration of FHWA NHI-07-071 Earth Retaining Structures

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surface water. Subsurface drainage is addressed in Section 5.5.10.1. There may be several soil zones behind an earth retaining structure. Groundwater flow from one zone to another, and then to a drain and outlet feature, should be unimpeded. If impeded, water will backup at the interface of the two adjacent zones increasing hydrostatic pressures and decreasing the stability of the wall structure. Soil filtration and permeability requirements must be met between the two adjacent zones of (different) soils to prevent impeded flow. Soil and geotextile filter design, and water collection components, are discussed in Section 5.5.10.2. Surface water runoff can destabilize a structure under construction by inundating the backfill. Surface water can also destabilize a constructed structure by erosion or infiltrate into the backfill. Design for surface water runoff is discussed in Section 5.5.10.3. In most cases, and especially for fill walls, it is preferable to provide backfill drainage rather than design the wall for the large hydrostatic water pressure resulting from a saturated backfill. Saturation of the backfill may result from either a high static water table from direct and/or indirect rainfall infiltrations, or other wet conditions. 5.5.10.1

Subsurface Drainage

Potential sources of subsurface water are groundwater and surface water infiltration, as illustrated in Figure 5-19. Groundwater may be present at an elevation above the bottom of the wall and would flow to the backfill from an excavation backcut; or it may be present beneath the bottom of the wall. A groundwater surface beneath a wall may rise into the structure, depending on the hydrogeology of the site. Surface water may infiltrate into the wall backfill from above, or from the front face of the wall for the case of flowing water in front of the structure (after Collin et al., 2002). Drainage system design depends on wall type, backfill and/or retained soil type, and groundwater conditions. Drainage system components (e.g., granular soils, prefabricated drainage elements, filters) are usually sized and selected based on local experience, site geometry, and estimated flows, although detailed design is only occasionally performed. Drainage systems may be omitted if the wall is designed to resist full water pressure. Drainage measures for fill wall systems, such as CIP walls, (and cut wall systems as well) typically consist of the use of a free-draining material at the back face, with “weep holes” and/or longitudinal collector drains (perforated pipes or gravel drains) along the back face as shown in Figure 5-20. This minimum amount of drainage should be sufficient if the wall backfill is relatively free-draining and allows the entire backfill to serve as a drain. It may be costly to fully backfill with free-draining or relatively free-draining material for some project applications therefore, it may be necessary to construct other types of drainage systems. FHWA NHI-07-071 Earth Retaining Structures

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Reinforced Fill

Retained Fill Groundwater

Drainage Aggregate

Foundation Soil

Figure 5-19. Potential Sources of Subsurface Water.

Backfill Soil Drainage Blanket

Backfill Soil Weephole

Longitudinal Drain Pipe

Prefabricated Drainage Element

Figure 5-20. Typical Retaining Wall Drainage Alternatives (after Sabatini et al., 1997).

Fill wall drains may be placed (1) immediately behind the concrete facing or wall stem; (2) between wall backfill and embankment fill; (3) along a backcut; and (4) as a blanket drain beneath the wall. Examples of drains behind a wall stem are shown in Figure 5-18. This drain primarily serves to collect surface water that has infiltrated immediately behind the wall, and transport it to an outlet. It may also serve to drain the wall backfill, if the backfill soil is relatively free-draining. It is noted, however, that even with vertical drains immediately behind the wall, appreciable water pressures can develop within the fill behind FHWA NHI-07-071 Earth Retaining Structures

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the drain leading to an increase in pressure on the wall itself. Inclined drains such as that shown in Figure 5-21 result in near vertical seepage, thus eliminating excess pressures in the sliding wedge that enters into earth pressure computations.

Figure 5-21. Inclined Drain for Reducing Water Pressure Behind Wall (after Cedergren, 1989).

A drain behind the wall backfill should be used when the backfill is not relatively freedraining. This may be a drain noted as (2) or as (3), above, and illustrated in Figure 5-22. A granular blanket drain, with collection pipes and outlets, should be used beneath fill wall structures where a high or seasonally high groundwater tables exist. 5.5.10.2

Drainage System Components

The drainage systems for fill walls may include: • • • • •

column(s) or zone(s) of free-draining gravel or coarse sand to collect water seepage from the backfill; perforated pipe(s) to collect water in the granular column(s) or zone(s); conveyance piping; outlet(s); and filter(s) between backfill soil(s) and granular column(s) or zone(s).

Longitudinal pipes transport collected water to outlet pipes that discharge at appropriate points in front of and/or below the wall. Outlets may be via weep holes through the wall facing and discharging in front of the structure to grade; via conveyance piping to storm


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Weephole

Wall Backfill

Retained Backfill

Face chimney drain

Chimney drain

Collection and Drain Pipes

Outlet Pipe

Foundation Soil

Figure 5-22. Drains Behind Backfill in Cantilever Wall in a Cut Situation. sewers (urban applications); or via conveyance piping to a slope beneath the wall structure. Weep holes generally consist of 1½ to 3 in. diameter holes that extend through the wall facing and are closely spaced (typically less than 10 ft. apart) horizontally along the wall. If weep holes are used with a counterfort wall, at least one weep hole should be located between counterforts. A screen and/or filter are used to prevent soil piping through a weep hole. The collection and conveyance pipes need to be large enough and sufficiently sloped to effectively drain water from behind the wall while maintaining sufficient pipe flow velocity to prevent sediment buildup in the pipe. Use of 3 to 4 in. diameter pipes is typical and practical (the diameter is usually much greater than that required for flow capacity). The design of pipe perforations (“holes”) and slots is provided in Section 5.2 of Cedergren (1989). Pipe outlets to slope areas beneath wall structures should be detailed similar to pavement drain outlets. If the outlet is to a grass area, it should have a concrete apron, vertical post marking location (for maintenance), and a screen to prevent animal ingress.


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Filters are required for water flowing between zones of different soils. A filter must prevent piping of the retained soil while providing sufficient permeability for unimpeded flow. The filter may be a soil or a geotextile. A geotextile is not required if the two adjacent soils meet soil filtration criteria. An open graded aggregate will generally not allow the development of a soil filter at its interface with the backfill soil, thus, requiring a geotextile filter. Soil and geotextile filter design criteria are summarized in Section 2.5.8. Geocomposite drains may be used in lieu of clean gravel or coarse sand and a geotextile. A geocomposite, or prefabricated, drain consists of a geotextile filter and a water collection and conveyance core. Geocomposites consist of cores that may be plastic waffles, threedimensional meshes or mats, extruded and fluted plastic sheets, or nets to convey water, which are covered by a geotextile filter. A wide variety of geocomposites are readily available. However, the filtration and flow properties, detailing requirements, and installation recommendations vary and may be poorly defined for some products. The flow capacity of geocomposite drains can be determined with ASTM D 4716, Test Method for Determining the (In-plane) Flow Rate per Unit Width and Hydraulic Transmissivity of a Geosynthetic Using a Constant Head. Long-term compressive stress and eccentric loadings on the geocomposite core should be considered during design and selection. The geotextile of the geocomposite should be designed to meet filter and permeability requirements. Installation details, such as joining adjacent sections of the geocomposite and connections to outlets, are usually product specific. These variances should be considered and addressed in the design, specification, detailing and construction phases of a project. Post installation examination of the drainage core/path with a cameras scope should be considered for critical applications. 5.5.10.3

Surface Water Runoff

Surface drainage is an important aspect of ensuring wall performance and must be addressed during design. Appropriate drainage measures to prevent surface water from infiltrating into the wall backfill should be included in the design of an earth retaining structure, whether of fill or cut construction. During construction of a fill wall, the backfill surface should be graded away from the wall face at the end of each day of construction to prevent water from ponding behind the wall and saturating the soil. Surface water running onto a partially completed backfill can carry fine-grained soils into and contaminate a free-draining granular backfill with fines. If a finegrained backfill soil is being utilized, saturation can cause movements of the partially constructed wall facing. FHWA NHI-07-071 Earth Retaining Structures

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Finish grading at the top of a wall structure should provide positive drainage away from the wall, when possible, to prevent or minimize infiltration of surface water into the backfill. If the area above the wall is paved, a curb and gutter is typically used to direct the flow. Concrete, asphalt or vegetation lined drainage swales may be used where a vegetated finished grade slopes to the wall. Water runoff over the top of a wall where the backfill slopes towards it and without a collection swale can lead to erosion and can cause staining of the wall face as soil is carried with the water. Runoff flow will concentrate at grading low points behind the face. The collection and conveyance swales should prevent overtopping of the wall for the design storm event. Extreme events (i.e., heavy rainfalls in short duration) have been known to cause substantial damage to earth retaining structures both during construction and during the life of the structure due to erosion, flooding, and/or increased hydrostatic pressures. This is particularly true for sites where surface drainage flows toward the wall structure and where finer-grained backfills are used. Site drainage features are designed for an assumed or prescribed design storm event (e.g., the 24-hour, 25 year storm event). However, extreme events do occur which result in short duration (i.e., 1 to 3 hours) flows that significantly exceed the design capacity of the stormwater management system. When this does occur, site flooding can result, causing overtopping of the wall, erosion, and an increase in hydrostatic forces within and behind the reinforced soil mass. If surface water flows toward an earth retaining structure, it is likely picked up in a gutter or other collection feature; and this feature is often sized based upon the design storm event. Similar to designing an overflow spillway with dam structures, the site layout and wall structure should include features for handling flows greater than the design event. The wall designer should address potential excess flows, and coordinate work with other project designers. Consideration should be given to incorporating overflow (i.e., spillway) details into the wall design for sites where surface water flows towards the wall structure.

5.6

CONSTRUCTION INSPECTION

Construction inspection requirements for CIP walls are similar to those for other concrete structures. In some cases, state agencies may have developed inspector checklists for this type of construction. Table 5-3 provides a summary of construction inspection requirements for CIP walls.


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Table 5-3. Inspector Responsibilities for a Typical CIP Gravity and Semi Gravity Wall Project CONTRACTOR SET UP Review Plans and Specifications Review Contractor’s schedule Review test results and certifications for preapproved materials (e.g., cement, coarse and fine aggregate) Confirm that Contractor stockpile and staging area are consistent with locations shown on Plans Discuss anticipated ground conditions and potential problems with Contractor Review Contractor’s survey results against Plans EXCAVATION Verify that excavation slopes and/or structural excavation support is consistent with the Plans Confirm that limits of any required excavations are within right-of-way limits shown on Plans Confirm all unsuitable materials (e.g., sod, snow, frost, topsoil, soft/muddy soil) are removed to the limits and depths shown on the Plans and the excavation is backfilled with granular material and properly compacted Confirm that leveling and proof-rolling of the foundation area is consistent with requirements of Specifications Confirm that Contractor’s excavation operations do not result in significant water ponding Confirm that existing drainage features, utilities, and other features are protected Identify areas not shown on Plans where unsuitable material exists and notify Engineer FOOTING Approve footing foundation condition before concrete is poured Confirm reinforcement strength, size, and type consistent with Specifications Confirm consistency of Contractor’s outline of the footing (footing size and bottom of footing depth) with Plans Confirm location and spacing of reinforcing steel consistent with Plans Confirm water/cement ratio and concrete mix design consistent with Specifications Record concrete volumes poured for the footing Confirm appropriate concrete curing times and methods as provided in the Specifications Confirm that concrete is not placed on ice, snow, or otherwise unsuitable ground Confirm that concrete is being placed in continuous horizontal layers and time between successive layers is consistent with Specifications


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STEM If used, confirm the placement of weep hole inserts (number, elevation, and specific locations) with Plans Confirm concrete is poured in section lengths consistent with Specifications Record concrete volumes used to form the stem Confirm that all wall face depressions, air pockets, gaps, rough spots, etc. are repaired Confirm storage of reinforcing bars is consistent with Specifications (e.g.- use of platform or supports) Perform preliminary check of condition of epoxy-coated reinforcing bars Confirm that forms are clean and appropriately braced during concrete pour operations Confirm that all reinforcing bars are held securely in place and being rigidly supported at the face of forms and in the bottom of wall footings Confirm that construction joints are being made only at locations shown on the Plans or approved by the Engineer DRAINAGE SYSTEMS AND BACKFILL Confirm that installation of the drainage system consistent with Specifications and Plans Confirm backfill material used is approved by the Engineer Confirm placement of backfill is performed as lifts consistent with Specifications Confirm that minimum concrete strength is achieved before backfill and compaction against back of wall Confirm that backfill placement method used by Contractor does not cause damage to prefabricated drainage material or drain pipe Confirm that earth cover over drainage pipes is sufficient to prevent damage from heavy equipment (minimum cover based on ground pressure from equipment should be provided in the Specifications) Perform required backfill density tests (at the frequencies specified), especially for areas that are compacted with lightweight equipment (e.g.- areas just behind the wall) Check that the drainage backfill just behind weep holes is the correct gradation and that it is properly installed POST INSTALLATION Verify pay quantities Note: Throughout project, check submittals for completeness before transmittal to Engineer


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CHAPTER 6 MODULAR GRAVITY WALLS 6.1

INTRODUCTION

This chapter describes types of modular gravity wall systems used in highway practice, including crib walls, concrete module walls, bin walls and gabion walls (Figure 6-1). Also presented are the specific design considerations and construction requirements for each of these wall types.

Figure 6-1. Modular Gravity Walls (a) Metal Bin Wall; (b) Precast Concrete Crib Wall; (c) Precast Concrete Module Wall; and (d) Gabion Wall (after AASHTO, 2007). A system that combines features of a modular gravity wall and an MSE wall is the so-called “T-Wall”. This is a proprietary system and, while other similar systems may be available in the U.S. market, in Section 6.9, specific information on this wall system is provided. Modular gravity walls comprise two major components: modular structural elements and the fill material placed within these elements. The structural elements, which may be proprietary, may consist of steel modules or bins, prefabricated concrete modules, timber units, wire baskets or other configurations and materials. The fill material used within the FHWA NHI-07-071 Earth Retaining Structures

6-1

6 – Modular Gravity Walls June 2008

modular units is recommended to be free draining granular soils, gravel, or rock fragments. Modular gravity walls rely on their own weight and the weight of the fill material within and above the wall elements to resist the applied loads and moments. Modular gravity walls have been used in a variety of project applications including highways, bridges, railroads, channels, dikes, and others (Figure 6-2). Modular gravity walls may be used where conventional cast-in-place concrete retaining walls are considered. Because of their modular configuration, they are easy to assemble, dismantle and transport; therefore, they are ideally suited for use in remote areas for construction of permanent or temporary fill wall systems. In 2007, most modular gravity walls constructed are either precast concrete module walls or gabion walls. Metal bin walls are rarely used unless for rehabilitation of existing metal bin walls. The primary advantages of modular gravity walls are: • • • • • • • •

Low cost; Fast and easy construction; Low labor requirement (except for gabion walls); Erection unaffected by temperature; “Plant-controlled” quality of prefabricated units; Flexibility and tolerance to differential settlements; No need for architectural finish; and Units can be disassembled and re-used economically.

The limitations of this type of wall system are: • • • • • • • •

Requires bottom-up construction; Potential corrosion of exposed metallic elements including metallic bins and wire used for gabion baskets; Potential decay of timber units exposed to water action; Required storage space at the site for relatively large prefabricated units; Possible unavailability of required fill material (especially required fill for gabion walls); Susceptibility of some systems to vandalism (wire baskets can be damaged more easily than concrete structures); Not suitable for use with deep foundations; and Many of the available systems are proprietary.


6-2


Figure 6-2. Examples of Modular Gravity Wall Applications (after Contech, 1997). FHWA NHI-07-071 Earth Retaining Structures

6-3


The principal types of modular gravity walls are the crib wall, the concrete module wall, the bin wall and the gabion wall. Although the same basic design considerations and procedures generally apply to all of these wall systems, each type of wall has distinctive features and construction requirements.

6.2

CRIB WALLS

6.2.1

General

Crib walls are built using prefabricated units which are stacked and interlocked and filled with free-draining granular material (Figure 6-3). The crib units are made of either reinforced concrete or timber. Timber units require treatment for long-term durability against decay, particularly in walls with extensive contact with water. The wall face can either be open or closed. In closed-face cribs, the stretchers are placed in contact with each other. In open-faced cribs, the stretchers are placed apart but at close intervals so that the infill material does not escape through the face. When subject to wave or current action, such as in stream or waterfront applications, a geotextile filter layer is usually placed behind the wall face to prevent the loss of fines through the spaces between stretchers, even in closed-face cribs. Walls higher than about 6 ft are usually built with a batter to improve stability. Crib walls of constant width are commonly used for heights up to about 15 ft (Figure 6-3a). For higher walls, stepped wall cross-sections (Figure 6-3b) are often used to increase stability and reduce the cost. 6.2.2

Wall Construction

The first stage of construction consists of excavating to the wall foundation grade and preparing the wall foundation. If stable excavation slopes cannot be maintained during this period of wall construction, temporary excavation support may be needed. Foundation preparation includes removing unsuitable materials such as organic matter and vegetation from the area to be occupied by the wall and leveling and proof-rolling the foundation area. A cast-in-place concrete leveling pad is normally provided at the base of the wall. This slab usually extends beyond the front and back faces of the crib and, on sloping ground, it is often stepped to follow the slope.


6-4


Note:

Type I is an open-face wall system. Types II and III are both closed-face wall systems

Figure 6-3. Crib Walls (a) Uniform Cross-Section; (b) Stepped Cross-Section; and (c) Typical Details of a Reinforced Concrete Crib Wall (after HKGEO, 1993). FHWA NHI-07-071 Earth Retaining Structures

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Figure 6-4. Setting Precast Elements for an Open Faced Crib Wall. Timber and concrete crib members are placed in successive tiers in accordance with the design spacing and arrangement (Figure 6-4). At the intersection of concrete header and stretcher members, asphalt felt shims, or other suitable material, are used to obtain uniform bearing between the members. After each course of stretchers is assembled, the void within the crib is filled with infill material. Backfilling progresses simultaneously with the erection of the crib members, and the backfill is carefully placed and compacted to avoid displacing or damaging the crib members. Unless the infill material can serve as a filter, a geotextile layer is usually provided behind the rear face of the crib wall to prevent migration of fines from the backfill. Drainage systems similar to those used in conventional CIP concrete walls may be needed, particularly for closed-face cribs. In open-face cribs, the interspaces can be planted with vegetation to help blend the wall with the surrounding environment. 6.2.3

Materials

Precast concrete should comply with the standard specifications for reinforced concrete. The precast concrete members should be constructed of Portland cement concrete with a FHWA NHI-07-071 Earth Retaining Structures

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minimum compressive strength at 28 days of 4,000 psi. For high walls (say greater than 20 ft) stresses in the lower members may require that concrete with higher compressive strength (say 5,000 psi) be considered. If the crib wall is constructed of timber, the material should be of structural grade, properly treated and should comply with standard specifications for structural timber (see Section 8 of AASHTO, 2007). Crib members should be fabricated so that they are fully interchangeable, without the need to drill, cut or offset the members to correct for non-uniform sections. Bolts, nuts and miscellaneous hardware should be galvanized. The infill material should be well graded, free-draining, granular soil which will not sift or flow through openings in the wall. Infill should have a maximum particle size of 3 in. and should have no more than 15 percent of particles passing the No. 200 sieve. For wall heights greater than say 20 ft, infill material with even less material passing the No. 200 sieve should be considered. The backfill material should consists of either well graded crushed rock that is at least 1 in. in diameter, or free draining nonexpansive soil free of organic and deleterious materials and confirming to the gradation limits specified in AASHTO T-27. Where cohesive soil is used, backfill soil must be a mixture of cohesive and cohesionless soils. The infill and backfill material should be placed in uniform layers not exceeding 12 in. in thickness, and compacted with a manual vibratory tamping device to at least 95 percent of the maximum density as determined in accordance with AASHTO T-99. Backfill behind the crib wall shall be placed following the erection of the wall as closely as possible. The wall height should never be greater than 3 ft above the backfill. 6.2.4

Cost

A cost range of crib walls up to 35 ft in height is typically about $25 - $35/ft2 of exposed wall face with the higher range associated with higher walls. Crib walls are typically not cost effective above 35 ft. Crib walls do not require skilled labor or specialized equipment therefore the cost range is most affected by wall height and cost to obtain and transport prefabricated units to the project site.


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6.3

CONCRETE MODULE WALLS

6.3.1

General

Concrete module walls use interlocking precast concrete cells erected at the site and filled with compacted earth (Figure 6-5). A common concrete module (Figure 6-5b) has a face height of about 4 ft, a face length about 8 ft, and a width ranging from 4 ft to 20 ft. The wall units can be assembled vertically or with a batter. A variety of surface treatments (striations, exposed aggregate, etc.) are available to meet specific aesthetic requirements. On top of the wall, a parapet module can be placed and held rigidly by a cast-in-place concrete slab. A reinforced cast-in-place or precast concrete footing is usually placed at the toe and heel of the wall. Assembly of the interlocking wall units does not require bolts, nuts, pins, or fasteners. Accordingly, corrosion problems are minimized. Since all the units are manufactured in the plant and hauled to the site, their fabrication is standardized and performed to shop-applied quality control procedures. 6.3.2

Wall Construction

Construction of a concrete module wall is relatively fast and simple, requires minimum labor and can be accomplished with a small crane. The first stage of construction consists of excavating to the wall foundation grade and preparing the wall foundation. If stable excavation slopes cannot be maintained during the period of wall construction, temporary excavation support may be needed. Foundation preparation includes removing unsuitable materials such as organic matter and vegetation from the area to be occupied by the wall and leveling and proof-rolling the foundation area. Removal of weak and/or compressible materials and replacement with stable soil would also be part of foundation preparation. Next, precast or CIP concrete footings are installed. The footings set the line and grade, establish the batter of the wall if it is not vertical, and help distribute the load to the foundation soils. The concrete modules arrive at the site ready to be installed. A crane equipped with a special handling device picks up the units and places them in their proper positions within the structure (Figure 6-6a). Using a building-block concept, the interlocking modules can quickly be aligned as they are placed. Between each module course, rubber pads or other suitable material are placed in the front horizontal joints, and at other bearing areas. Alignment and elevations are corrected by adding or removing pads. FHWA NHI-07-071 Earth Retaining Structures

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Figure 6-5. Concrete Module Wall (a) Typical Section; (b) Typical Module; and (c) Precast Parapet (after Doublewal Corporation). FHWA NHI-07-071 Earth Retaining Structures

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(a)

(b)

(c) Figure 6-6. Construction of Concrete Module Wall (a) Placement of Precast Modules; (b) Placing Fill Within the Modules; and (c) Compacting the Infill Material (after Doublewal Corp.). FHWA NHI-07-071 Earth Retaining Structures

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Once the modules are properly set, shims are used to fill any space between the keys to prevent movement of the modules during backfilling. A geotextile filter layer is placed behind the front vertical joints to prevent the migration of fines and allow the joints to act as weep holes permitting the passage of water that might otherwise build up behind the wall. After each course is set, the modules are filled with select material (Figure 6-6b), and backfill is placed behind them. The infill and backfill are carefully placed and compacted to avoid displacing or damaging the concrete modules (Figure 6-6c). The parapet units, if required, are placed on top of the upper course, and adjusted to the proper alignment by hardwood wedges. A concrete slab is then cast-in-place to hold the wall parapet in its permanent position. 6.3.3

Materials

Precast concrete should comply with the standard specifications for reinforced concrete. The precast concrete modules should be constructed of Portland cement concrete with a minimum compressive strength at 28 days of 4,000 psi. The concrete modules should be fabricated so that they are fully interchangeable, without the need to drill, cut or offset the modules to correct for non-uniform sections. The infill material should be well graded, free-draining, granular soil which will not sift or flow through openings in the wall. Infill should have a maximum particle size of 3 in. and should have no more than 15 percent of particles passing the No. 200 sieve. For wall heights greater than say 20 ft, infill material with even less material passing the No. 200 sieve should be considered. The infill should be placed in uniform layers not exceeding 12 in. in thickness, and compacted with a manual vibratory tamping device to at least 95 percent of the maximum density as determined in accordance with AASHTO T-99. The backfill behind the wall is placed and compacted following normal retaining wall construction procedures. 6.3.4

Cost

A cost range of concrete module walls up to 25 ft in height is typically about $30 - $35 ft2 exposed wall face with the higher range associated with higher walls. Concrete module walls are typically not cost effective above 35 ft. Just like crib walls, concrete module walls do not require skilled labor or specialized equipment therefore the cost range typically varies due to wall height and material costs.


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6.4

GABION WALLS

6.4.1

General

Gabion walls are composed of rows and tiers of orthogonal wire cages or baskets filled with rock fragments and tied together. They are widely used for channel and river bank protection, but are also used for earth retaining structures, particularly in rugged terrain. The finished wall appearance easily blends with the natural landscape, especially if local rock is used as infill material. Gabion walls are simple and easy to construct. The construction of gabion walls becomes more economical for sites where suitable infill rock is available. Gabion walls are free draining and, if the backfill does not trap water, they will not be subjected to hydrostatic water pressures. Because they are free-draining, they also are frost resistant. Since gabion walls are flexible structures, they are suitable for construction over compressible soils. They are also well suited for remote areas that cannot be easily accessed by heavy machinery. In these situations, the wire baskets are assembled by hand and filled with local stone. Another advantage of gabion walls is that they allow penetration by protruding objects such as pipes. Gabion walls are very labor intensive; thus, they may not be cost-effective in certain regions. The wire baskets may be subject to vandalism, and the long-term durability of the wire may be questionable, particularly in corrosive environments. Some systems are available with PVC coated wires. The front and rear faces of a gabion wall may be vertical or stepped. A batter is usually provided for walls higher than about 10 ft to improve stability. The wall batters commonly used are 1 in 10, 1 in 6, and 1 in 4. A variety of cage sizes can be produced to suit the terrain with a standard width of 3 ft and lengths ranging from 3 to 12 ft. Available basket heights generally include 1 ft, 1.5 ft, and 3 ft. The longer gabions may be divided into cells by diaphragms made of the same mesh as the gabion basket itself, and directly joined to the base panel during manufacture. These diaphragms reinforce the structure and make assembly and erection easier (Figure 6-7).


6-12


1.5 ft 3 ft

3 ft 12 ft 6 ft 3 ft

Figure 6-7. Gabion Baskets (a) Module without Diaphragms and (b) Module with Diaphragms.

6.4.2

Wall Construction

The first stage of construction consists of excavating to the wall foundation grade and preparing the wall foundation. If stable excavation slopes cannot be maintained during the period of wall construction, temporary excavation support may be needed. Foundation preparation includes removing unsuitable materials such as organic matter and vegetation from the area to be occupied by the wall and leveling and proof-rolling the foundation area. Removal of weak and/or compressible materials and replacement with stable soil would also be part of foundation preparation. Gravel bedding layer or concrete leveling pad is often placed to provide a working surface and to help establish the alignment and elevation of the wall. The hexagonal woven-mesh gabions are supplied folded flat and in a bundle. Each gabion is erected at the site by folding up the sides and lacing together all vertical edges with the lacing wire. The four corners of the box are laced first, followed by lacing the edges of internal diaphragms to the sides of the box. The erected units are placed in their proper locations and adjoining empty gabions are laced along the perimeter of their contact surfaces to obtain a monolithic structure. The connected units are then stretched taut along their proper alignment before they are filled. The first gabion in the line is usually partially filled to provide anchorage during stretching. The welded wire gabions are either assembled in the factory and shipped flat to the project site similar to the woven-mesh gabions, or the wire mesh panels are cut in the field to the FHWA NHI-07-071 Earth Retaining Structures

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dimensions of the sides, top and base of the units in separate pieces which are joined together by spiral binders or tie wires. No stretching is required for the welded wire gabions. For example, for a 3-ft high basket, the baskets are tightly filled in three equal layers to ensure minimum voids (Figure 6-8a). Tie wires between the front and back faces of the gabion basket are typically installed at the top of the first and second infill layers to restrain bulging of the front face. Adjoining units are filled to the same elevation during each stage of filling. The last layers of stone are leveled with the top of the gabion to allow proper closing of the lid, and to provide a level surface for the following course (Figure 6-8b). The mesh of the lid is tied down to the sides and ends of the basket as well as to any internal diaphragm panels. Well packed filling (without undue bulging) and secure connections are essential in all gabion structures. Gabions are filled by any type of earth-handling equipment such as a backhoe, clam shell, etc. Care must be taken to avoid damaging the wire mesh or its protective coating by sharp particles of crushed rock. The gabion wall can be constructed on curves or angles by cutting and folding the wire-mesh to make units of special shapes and sizes. When building under water, the gabion units are usually assembled and filled at the surface then lowered into position by a crane. Prior to placing backfill, a geotextile filter material or graded soil filter is placed to fully cover the back faces and steps of the gabion wall. 6.4.3

Materials

Gabion baskets are made from a range of materials including steel, nylon, polypropylene or polyethylene. The polymer-type materials have the advantage of being lightweight and corrosion resistant. However, they are susceptible to attack by fire or ultraviolet light. Therefore, if these materials are used in making the baskets, it is advisable to cover the exposed grids with a non-flammable material such as shotcrete. The material most commonly used in commercial production of gabions is steel wire-mesh, which is manufactured in two types: hexagonal woven with an approximate opening size of 3.25 in. by 4.5 in., and square welded with an approximate opening side dimension of 3 in.


6-14


(a)

(b) Figure 6-8. Gabion Wall Construction (a) Filling Gabion Baskets with Stone and (b) Closing Gabion Lid for Tying. FHWA NHI-07-071 Earth Retaining Structures

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The hexagonal wire mesh is mechanically woven in a continuous sheet. The wires are double twisted to form the mesh. The diameter of the steel wires is between 0.08 and 0.15 in. The gabion base, top and sides are usually formed from a single piece of mesh, with its edges made of a wire with a diameter of about 1.5 times that of the wire mesh to prevent unraveling. High-tensile zinc-coated, steel wires are used, with a minimum tensile strength of 50,000 psi. For permanent applications, the wires should be at least 0.1 in. in diameter and galvanized before weaving. The welded wire mesh is manufactured from high tensile steel wire, electrically welded at each intersection. The welded mesh should be hot-dip galvanized after welding. The fabrication of panels by welding the wires after they have been galvanized is not recommended as the welds are left unprotected (HKGEO, 1993). Since the wires are welded at each intersection, the welded wire mesh has less ability to stretch and contract, and thus the assembled baskets are less flexible than comparable woven wire mesh gabions. In highly corrosive environments, such as sea water or polluted water, PVC (polyvinyl chloride) coating is usually provided for the wires. A black, PVC coat, 0.016 to 0.024 in. thick, is applied to the woven mesh by hot dipping or by extrusion onto the galvanized wire before weaving. For welded mesh, the PVC coating is applied electrostatically to the welded panel. The PVC coat should be bonded sufficiently to the wire core to prevent capillary flow of water between the wire and the PVC, which may cause corrosion. The durability of the PVC coating against contact with acidic, salt or polluted water, and exposure to ultra violet light and abrasion, is tested in accordance with ASTM E 42-65 specifications. All wires used in the construction of gabion walls (lacing wires, etc.) are made of the same quality wire used in the wire mesh. In waterfront structures, armor units, such as rip rap, are usually provided to protect the wire baskets from wave action or heavy water-borne material. The infill material should consist of rock fragments that are sound, durable and well-graded. The inclusion of objectionable quantities of shale, dirt, sand, clay, rock fines, and other deleterious materials should not be permitted with rock fragments. The rock fragments should be between 4 in and 10 in. in diameter, based on U.S. Standard square mesh sieves. No rock fragment should have a minimum dimension less than 4 in. The backfill material should consist of open-graded granular soil similar to that placed behind other gravity-type structures. The backfill material should be compacted to at least 95 percent of the maximum dry density as defined in AASHTO T 99 within 1 ft of the top of the wall. The top 1 ft should be compacted to at least 100 percent of the maximum laboratory dry density. In partially submerged walls, the backfill material should be free-draining to prevent the build up of water pressure behind the wall when the water level in front of the FHWA NHI-07-071 Earth Retaining Structures

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wall is lowered. Drainage behind the wall should be similar to that provided for crib walls and other types of modular walls. A filter layer (soil or geotextile) is usually provided between the gabion baskets and the backfill to prevent migration of the backfill soil through the relatively large voids of the infill material. 6.4.4

Cost

A cost range of gabion walls up to 25 ft is typically about $30 - $50/ft2 of exposed wall face. Gabion walls are typically not cost effective above 25 ft. Gabion walls do not require skilled labor or specialized equipment however, they require significant labor effort. The labor cost can be up to 50 percent of the total cost. Gabion walls require select stone material for infill and depending on the location of the source relative to the project site, the cost of gabion walls can vary significantly. Typically, if the source of the infill rock is not local, the cost of a gabion wall can be so high that it may not be economical to consider this wall type for the project.

6.5

BIN WALLS

6.5.1

General

Bin walls consist of adjoining closed-face cells filled with compacted select backfill to form a gravity type retaining structure. The cells are generally constructed of sturdy, lightweight, steel members that are bolted together at the site. The flexibility of the steel structure allows the wall to flex against minor ground movements that might damage rigid-type walls. Figure 6-9 illustrates the typical geometry of a bin wall. Basically, the wall is assembled using modular panels called stringers and spacers. The stringers constitute the front and back faces of the bin and the spacers its sides. Vertical connectors, grade plates, and stringer stiffeners act to hold the wall together (Figure 6-10). The stringers are approximately 10 ft long while shorter stringers are sometimes used to allow curvature in the wall. The length of the spacer (wall depth) is variable, and is determined by stability requirements. Wall heights may vary in 1.33 ft increments. The walls of the bin (stringers and spacers) are formed of S-shaped steel members. As the wall height and depth increase, the thickness of these members is also increased to resist the internal pressure from the fill material. Table 6-1 shows the typical relationship between thickness of steel panel members and wall height, as recommended by one wall manufacturer. FHWA NHI-07-071 Earth Retaining Structures

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Table 6-1. Typical Height-Thickness Relationship for Bin Walls (after Contech). Wall Height (ft)

Wall Depth (ft)

Steel Thickness (in.)

Gage

0 - 12 12 - 15 15 - 25 25 - 35

5.5 – 7.7 7.7 – 9.9 9.9 – 14.3 14.3 – 16.5

.064 .079 .109 .138

16 14 12 10

The advantages of bin walls include flexibility, versatility, ease of installation and dismantling for reuse. Their main disadvantage is the potential corrosion of the steel elements when the wall units are exposed to corrosive materials or installed in aggressive electrochemical environments. 6.5.2

Wall Construction

The lightweight wall components are easily assembled by a small crew of unskilled labor, using a small crane and ordinary hand tools. The first stage of construction consists of excavating to the wall foundation grade and preparing the wall foundation. If stable excavation slopes cannot be maintained during the period of wall construction, temporary excavation support may be needed. Foundation preparation includes removing unsuitable materials such as organic matter and vegetation from the area to be occupied by the wall and leveling and proof-rolling the foundation area. Removal of weak and/or compressible materials and replacement with stable soil would also be part of foundation preparation. A gravel working pad can be placed to facilitate bin construction. Generally, the bins are erected at the site by bolting panels to vertical connector elements. However, on larger projects, or where unusual working conditions are encountered, panel and transverse sections may be preassembled at the shop, or they may be assembled off-site then transported to the site (Figure 6-11). If the infill has a high percentage of fine sand, the coarser material is usually placed adjacent to the vertical connectors in the bin corner to prevent the loss of fine material through normal small openings in the corners. Caulking or sealing these critical areas with geotextile filter fabric material can also be done.


6-18


Figure 6-9. Typical Geometry of Type 2 Bin Wall (a) Plan, (b) Elevation and (c) Section AA (after Contech) FHWA NHI-07-071 Earth Retaining Structures

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Figure 6-10. Elements of Bin Walls (a) T-Shaped Vertical Connector for Bin Wall Type 2 and (b) Channel Shaped Vertical Connector for Bin Wall Type 1(after Contech). FHWA NHI-07-071 Earth Retaining Structures

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Figure 6-11a. Construction of a Bin Wall (a) Setting Preassembled Panels and (b) Filling the Completed Bins (after Contech).

Figure 6-12b. Construction of a Bin Wall (a) Setting Preassembled Panels and (b) Filling the Completed Bins (after Contech). FHWA NHI-07-071 Earth Retaining Structures

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As the bins are filled, the backfill material is placed behind the wall and compacted using standard earthwork construction procedures. At all times, the level of the infill material should be kept higher than the fill behind the wall, to maintain stability. Construction of bin walls on curves requires special attention. Installation of the wall can be accomplished through one of two methods (Figure 6-12) as follows: 1. Shorter stringers can be used at the front and the rear faces of the wall. These stringers can either be manufactured in shorter sizes, or can be cut in the field to fit any wall configuration. When stringers are cut in the field, new bolt holes must be drilled or burnt in the field to match the standard hole pattern in the ends of the cut stringers. 2. Shop-fabricated corner plates are used to provide angularity and curvature.

Figure 6-13. Construction of Bin Walls at Curves (a) Typical Outside Corner; and (b) Typical Inside Corner (after Contech).


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6.5.3

Materials

The wall units are usually formed from galvanized steel sheets. Where the electrochemical environment is unusually severe, the wall may be built using polymeric coated steel components. Aluminized steel coated components have also been used, but for stringers and spacers only. The exposed stringer, or front panel, typically consists of corrugated steel or aluminized steel panels (Figure 6-13a), but precast concrete panels with variable architectural treatments (Figure 6-13b) can also be used. The infill material should be well graded, free-draining, granular soil which will not sift or flow through openings in the wall. Infill should have a maximum particle size of 3 in. and should have no more than 15 percent of particles passing the No. 200 sieve. For wall heights greater than say 20 ft, infill material with even less material passing the No. 200 sieve should be considered. The infill material should be placed in uniform layers not exceeding 12 in. in thickness and compacted using manual tamping equipment to at least 95 percent of the maximum density as determined in accordance with AASHTO T-99. The backfill material may be placed in slightly thicker loose lifts and should be compacted to a minimum 90 percent of the maximum density as determined in accordance with AASHTO T-99. 6.5.4

Cost

A cost range of bin walls up to 35 ft is typically about $25 - $35/ft2 of exposed wall face with the higher range associated with higher walls. Bin walls are typically not cost effective above 35 ft. Bin walls do not require skilled labor or specialized equipment therefore, like other modular walls their cost is typically affected by the cost of materials.

6.6

MODULAR GRAVITY WALL DESIGN

6.6.1

General

The design of crib walls, and other types of modular gravity walls, requires an evaluation of sliding, overturning (i.e., limiting eccentricity), and bearing resistance at the strength limit state and wall settlement at the service limit state. Drainage is also a major design consideration. These design issues should be evaluated using the same procedures described in wall design of CIP concrete walls in Chapter 5. Specific considerations and procedures for design of modular gravity walls are provided herein. FHWA NHI-07-071 Earth Retaining Structures

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(a)

(b) Figure 6-14. Bin wall with (a) Corrugated Steel Face Panels; and (b) Precast Concrete Face Panels (after Contech). FHWA NHI-07-071 Earth Retaining Structures

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6.6.2

Lateral Earth Pressures

The magnitude and location of resultant loads and moments for modular walls may be calculated using the earth pressure distributions shown in Figures 6-14 and 6-15. Where the back of the modules forms an irregular, stepped surface, the earth pressure should be computed on a plane surface drawn from the upper back corner of the top of the module to the lower back heel of the bottom module using Coulomb earth pressure theory. If the wall is founded on a relatively incompressible foundation material, the angle of wall friction (δ) may be assumed to be equal to φ′/2, where φ′ is the friction angle of the compacted backfill soil. It is important to adequately compact the infill material for the assumption of δ = φ′/2 to be appropriate. If the back face of the wall is continuous as shown in Figure 6-14, the angle of wall friction should be assumed as δ = φ′/2 and if the back face of the wall is stepped as shown in Figure 6-15, the angle of friction should be assumed as δ = 3/4φ′ for the incompressible foundation situation. For walls on relatively compressible foundation material, the wall friction should be assumed as δ = 0 as the wall may settle relative to the backfill (See Table C3.11.5.9-1 in AASHTO (2007)). In addition to the lateral earth pressure, the wall must be designed for lateral pressure due to surcharge loads. Positive measures must be provided to drain the material within and behind the modules or bins. Drainage is typically provided by installing perforated drain pipes behind and below the rear base of the bin or crib wall. The perforated pipe should be surrounded by pervious backfill material wrapped in a geotextile filter material, and provided with suitable drainage outlets. Filtration criteria between the retained backfill and the granular wall fill should be checked and a filter used if needed to prevent potential piping of the backfill soils. 6.6.3

Sliding

Resistance to sliding for a prefabricated modular wall is provided by infill to foundation material interface friction, friction at the base of the modules, and, for walls with footings, the frictional resistance developed between the bottom of the footings and the foundation materials. The coefficient of sliding friction between the infill and foundation material at the wall base should be the lesser of the friction angle of the infill and the friction angle of the foundation soil. Calculation methods are similar to CIP walls as described in Section 5.5.7.


6-25


θ W1

H

δ

H

Pa

W1 Pa

δ

H/3

H/3

β

β θ

θ=90º

W2

W2 H

H

W1

δ

W1

Pa

δ

H/3

Pa H/3

Figure 6-15. Earth Pressure Distribution for Modular Walls with Continuous Pressure Surfaces (Figure 3.11.5.9-1 from AASHTO, 2007). As with other gravity retaining structures, the potential benefit derived from passive resistance on front of the wall due to wall embedment is in most cases conservatively neglected. Additional shear resistance can be developed for modular walls by inclining the base of the wall or by increasing the base width of the wall. Possible erosion of the soil at the wall toe by an act of nature, or removal by human activities, should be taken into account in the design. In performing sliding analyses for modular walls with footings (e.g., concrete module walls), it is assumed that 80 percent of the weight of the soil in the modules is transferred to the footing supports (through soil arching) with the remaining soil weight being transferred to the area of the wall between the footings.


6-26


θ

θ

W1

W1 H

Pa

δ

W2

H/3

W3

θ

H

θ

β W5

W2 W3 W4

W4

Pa

δ H/3

W3

β

W1

W1 H

W2

δ

Pa

H

H/3

W2 W3

δ

Pa H/3

Figure 6-16. Earth Pressure Distribution for Modular Walls with Irregular Pressure Surfaces (Figure 3.11.5.9-2 from AASHTO, 2007). 6.6.4

Limiting Eccentricity

Because the interior of modular walls are backfilled with soil to complete their construction and because the interior of the backfill soil can move with respect to the retaining module (for open bottom modules) if the module is uplifted or overturned, the full weight of the backfilled structure is assumed to not be effective in resisting eccentric (overturning) loads. Therefore, the soil load resisting overturning is limited to a maximum of 80 percent of the weight of the soil in the modules. However, if a structural bottom is provided to the module, no reduction of soil weight is taken. As an example, if the infill soil in a module has a unit weight of 110 pcf and the volume of fill per unit length of wall is 9.6 ft3/ft, the total weight of fill in the module, WF, is WF = Volume × Unit Weight = 9.6 FHWA NHI-07-071 Earth Retaining Structures

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ft 3 (110 pcf ) = 1056 lb . ft ft

(6-1)


To evaluate a Strength I load group for eccentricity evaluations, 80 percent of the weight of the fill, WF, is factored by the load factor, γEV of 1.35. The factored weight of the fill is:  lb  lb Factored WF = γ EV × (0.8 × WF ) = 1.35 ×  0.8 ×1056  = 1140.5 . ft  ft 

(6-2)

The same limits on eccentricity as were provided in Chapter 5 for CIP concrete walls are used for modular walls. Calculation methods are similar to CIP walls as described in Section 5.5.6. 6.6.5

Bearing Resistance

Bearing resistance is evaluated by assuming that a minimum of 80 percent of dead loads and earth pressure loads are transferred to point supports per unit length at the rear and front of the modules or at the location of the bottom legs (see Figure 6-16). If foundation conditions require a footing under the total area of the module, all of the soil weight inside the modules should be considered. Calculation methods are similar to CIP walls as described in Section 5.5.5.

B = 2 ft CONCRETE BEARING PAD

Figure 6-17. Load Distribution for Modular Wall with Footings.


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6.6.6

Structural Capacity of Wall Modules

Structural design of the module members and base slab, if any, is performed using the factored loads developed for the geotechnical design of the wall foundation and the factored structural resistance of reinforced concrete (for concrete systems), and the factored structural resistance of steel (for steel systems). Structural design of the modules is based on the difference between pressures developed inside the module (bin pressures) and those resulting from the backfill. Rear facing surfaces are often designed for this condition as well as for the factored earth pressures developed inside the module during construction (module has been filled, but backfill has not yet been placed). The computed compression, shear and bending stresses can be compared with the manufacturer specified structural capacity (see Section 11.11.5.1 of AASHTO (2007)). 6.6.7

Lateral and Vertical Displacements at the Service Limit State

The vertical and lateral displacements of prefabricated modular retaining walls must be evaluated at the Service I limit state for all applicable load combinations and compared with tolerable movement criteria. Similar to conventional retaining walls, prefabricated modular retaining walls must deflect (either by foundation movement or structural deformation) a sufficient magnitude to permit mobilization of the shear strength of the backfill soil and development of the design (usually active) earth pressure on the wall. In general, lateral movements of walls on shallow foundations can be estimated assuming the wall rotates or translates as a rigid body due to the effects of applied loads and the corresponding differential settlements along the base of the wall. Tolerable movement criteria for prefabricated modular walls should be developed with consideration of the function and type of wall, anticipated service life, and consequences of unacceptable movements (e.g., the specific wall application and the affect of wall movements on adjacent facilities). Most prefabricated modular walls have tolerable differential movements, expressed as a ratio of differential settlement to horizontal distance on the order of 1/300 along the alignment of the wall, except for gabion walls which may tolerate differential settlements on the order of 1/50. 6.6.8

Overall Stability at the Service Limit State

The overall stability of a modular wall, including the retained ground and foundation, should be evaluated using limit equilibrium methods of slope stability analysis. Calculation methods are similar to CIP walls as described in Section 5.5.8.


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6.6.9

Gabion Walls

Design issues specific to gabion walls are provided below: •

Since the infill and backfill materials are well-drained, no hydrostatic pressure is used in the design.

•

The angle of wall friction, δ, between the gabion structure and the backfill material may be reduced if a geotextile layer is used as a filter on the back face of the gabion structure.

•

For stepped walls, analyses for sliding and eccentricity should be conducted at each change in wall section, ignoring the resistance contributed by the wire mesh and the connection between the baskets.

•

The unit weight of the wall infill material used in design analyses is dependent on the type and porosity of the infill material. The weight of the wire basket is relatively small and is usually ignored in the design. The unit weight, γg, of the wall is calculated as follows:

γ g = (1 − nr )G S γ W

(6-3)

where nr is the porosity of the rock fill, Gs is the specific gravity of the rock, and γw is the unit weight of water. The porosity of the infill material is generally between 0.3 and 0.4 (HKGEO, 1993), and is dependent on the grading and angularity of the rock fragments, as well as the method of installation and degree of compaction of the infill material.

6.7

EXAMPLE PROBLEM

Design a gabion wall 12 ft high, supporting a soil backfill with a uniform surcharge of 200 psf. The back face of the wall is vertical and a geotextile filter layer will be placed between the back face of the wall and the retained fill soil. Properties of the backfill are γ = 120 pcf and φ = 34o. The rockfill within the wire basket will be compacted to a unit weight of 100 pcf. The undrained shear strength of the foundation soil is 1,500 psf. The geometry and parameters of the wall are given in Figure 6-17 below.


6-30


q = 200 psf 3 ft

4 @ 3ft = 12 ft

4 3

γ = 120 pcf φbackfill = 34º

2 1 3 @ 3ft = 9 ft

Figure 6-18. Geometry and Parameters of Example Problem. Solution

Step 1: Select a trial section Consider using a front-stepped vertical back gabion wall, consisting of 3 ft high baskets, placed for a cumulative base width of 9 ft as shown in Figure 6-17. Evaluate the following Strength I limit states: (1) eccentricity; (2) bearing resistance; and (3) sliding. Step 2: Calculate weight of fill in each level of wall (WF) Section(1) 1 2 3 4 (1)

Height / Width (ft) 3/9 3/6 3/6 3/3

Volume of Fill (ft3/ft) 27 18 18 9

WF (lb/ft) 2,700 1,800 1,800 900

Section numbers labeled on Figure 6-17

Step 3: Calculate horizontal earth force due to earth pressure (EH) and earth surcharge (ES) Assume δ equal to zero and calculate Ka based on Rankine theory Ka = tan2 (45 – φbackfill/2) = tan2 (45 – 34/2) = 0.28 PEH = ½ KaγH2 = 0.5 (0.28) (100 pcf) (12 ft)2 = 2,016 lb/ft PES = KaqH = 0.28 (200 psf) (12 ft) = 672 lb/ft FHWA NHI-07-071 Earth Retaining Structures

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Step 4: Summarize unfactored vertical and horizontal forces. Summary of Unfactored Vertical Loads Item WF1 WF2 WF3 WF4

V (lb/ft) 2,700 1,800 1,800 900

Moment Arm (ft) 4.5 6 6 7.5

Moment (lb-ft/ft) 12,150 10,800 10,800 6,750

Moment Arm (ft) 4 6

Moment (lb-ft/ft) 8,064 4,032

Summary of Unfactored Horizontal Loads Item PEH PES

H (lb/ft) 2,016 672

Step 5: Select Load Factors For Sliding and Eccentricity γEV = 1.0 (minimum load factor typically appropriate for sliding) γES = 1.5 γEH = 1.5 For Bearing Resistance γEV = 1.35 (maximum load factor typically appropriate for bearing resistance) γES = 1.5 γEH = 1.5 Step 6: Evaluate eccentricity e=

1.5 (8,064) + 1.5 (4,032) −1.0 (1,800) (1.5 ft ) − 1.0 (1,800) (1.5 ft ) − 1.0 (900) (3 ft ) =1.395 ft 1.0 (2,700 + 1,800 + 1,800 + 900)

e = 1.395 ft < B/6 (= 9ft /6 ) = 1.5 ft ∴OK Step 7: Check bearing resistance (a) Calculate eccentricity using load factors for bearing resistance calculations e=

1.5 (8,064) + 1.5 (4,032) − 1.35 (1,800) (1.5 ft ) − 1.35 (1,800) (1.5 ft ) − 1.35 (900) (3 ft ) = 0.74 ft 1.35 (2,700 + 1,800 + 1,800 + 900)


6-32


(b) Calculate vertical stress, σv

σv =

∑V

1.35 (2,700 + 1,800 + 1,800 + 900) lb / ft = 1,292 psf B − 2e (9 ft − 2 (0.74 ft )) =

(c) Calculate factored bearing resistance Undrained shear strength = 1,500 psf Assume embedment effects are conservatively neglected

qn = c N cm N cm = N c sγ iγ

N c = 5.14 ( for φ = 0)

sγ = 1 (assumes long wall ) iγ = 1.0 qn =1,500 psf (5.14) = 7,710 psf Resistance factor, φ, is 0.65 (see Table 10.5.5.2.2-1 in AASHTO (2007)) qR = φ qn = 0.65 (7,710 psf) = 5,011 psf qR = 5,011 psf > σv = 1,292 psf ∴OK

Step 8: Check sliding (a) Calculate factored resistance against sliding (assume passive resistance neglected) RR = φτ Rτ Rτ = V tan φfill = (1,200 + 1,800 + 1,800 + 900) lb/ft tan (32º) = 4,499 lb/ft For soil on soil, φτ = 1.0 (Table 10.5.5.2.2-1 in AASHTO (2007)) RR = 1.0 (4,499 lb/ft) = 4,499 lb/ft FHWA NHI-07-071 Earth Retaining Structures

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(b) Calculate total factored horizontal load, HTOT HTOT = 1.5 (2,016 lb/ft) + 1.5 (672 lb/ft) = 4,032 lb/ft RR = 4,499 lb/ft > HTOT = 4,032 lb/ft ∴OK Summary

All Strength I limit states have been satisfied. Service I limit states (i.e., overall stability and wall lateral and vertical displacement) would be checked.

6.8

CONSTRUCTION INSPECTION

Inspector responsibilities of typical modular gravity walls are presented in Table 6-2.

Table 6-2. Inspector Responsibilities for a Typical Modular Gravity Wall Project. CONTRACTOR SET UP

Review Plans and Specifications Review Contractor’s schedule Review corrosion protection requirements of metallic units Review infill and backfill soil test results (i.e., grain size, unit weight, and shear strength etc.) Confirm that Contractor stockpile area consistent with locations shown in Plans Discuss anticipated ground conditions and potential problems with Contractor Review Contractor’s survey results according to Plans Check site conditions and observe site accessibility, construction dewatering, and drainage features such as seeps, adjacent streams, etc. Check submittals of the Contractor for obvious deficiencies and act accordingly before sending it to Engineer for review Confirm that Contractor’s length, size, and unit arrangements of the wall is consistent with Plans Inspect wall units for cracks, chips or unsightliness before the beginning of installation. EXCAVATION Verify that Contractor’s slope excavation and/or structural excavation support are consistent with the Plans Confirm that limits of any required excavations are within right-of-way limits shown on the Plans Confirm all unsuitable materials are removed to the limits and depths shown on the Plans


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SUBGRADE Confirm stability of slopes and report to the Engineer any temporary excavation support systems being used Confirm limits of excavation and wall footing as shown in the Plans Conform all unsuitable materials are removed from the area to be occupied by the wall Confirm construction dewatering consistent with Specifications Confirm bedrock (or competent ground) according to the Specifications Conform leveling, alignment, and proof-rolling of the subgrade consistent with Specifications and Plans Confirm that the layouts, dimensions, and elevations of the leveling pads are consistent with Plans (3 and 4) WALL FACE Confirm that Contractor’s arrangement of wall face and alignment consistent with Plans Confirm that shear keys are wrapped with joint material and placed as shown in Plans Confirm erection of gabions and lacing of all vertical edges consistent with Specifications(2) Confirm that gabion basket are coated with specified material(2) Confirm erection and bolting of the bin wall panels consistent with Specifications and Plans(1) Confirm placement of wall elements consistent with Plans(3, 4) Confirm installation of stretcher members and minimum edge distances consistent with Specifications(3) Confirm placement of bearing pads for horizontal joints consistent with Plans(4) If required, confirm installation of parapet units consistent with Plans DRAINAGE SYSTEMS, INFILL AND BACKFIILL Confirm that soils used as backfill are approved by Engineer Confirm that soils used for infill are approved by Engineer Confirm that stone filling operations are performed carefully without damaging galvanized wire coating (2) Confirm that gabion basket lids are stretched tight over the stone fill using crowbars or lid closing tools (2) Confirm that backfilling behind the modular gravity system follows erection of the wall as closely as possible and the wall height is never greater than three feet above the backfill. Confirm placement of infill and backfill is performed as lifts consistent with Specifications Confirm compaction of infill and backfill material consistent with Specifications Confirm the geosynthetic material for drainage consistent with Specifications Confirm placement methods and limits of coverage of drainage materials consistent with Specifications and Plans POST INSTALLATION Verify pay quantities Note: (1)bin wall, (2)gabion wall, (3)crib wall, and (4)concrete modular wall.


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6.9

T-WALLS

6.9.1

Overview

T-WALL® (T-wall) is a proprietary wall system that consists of precast concrete units that resemble a “T” shape. T-wall is considered a gravity structure that combines the weight of externally stabilized retaining walls (e.g., concrete module wall) with the frictional resistance of internally stabilized systems (e.g., MSE wall). The concrete module has a face height of 2.5 ft (single unit) or 5.0 ft (double unit), a face width of 5.0 ft and stem lengths ranging from 4 to 20 ft. The units can be installed either vertically or on a batter (Figures 6-18 and 6-19). In general, vertical walls can more easily accommodate bends and curves. Battered and tiered walls permit shorter stem lengths, resulting in less excavation and backfill.

Figure 6-19. Vertically installed T-wall units (after the Neel Company, 2004).


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Figure 6-20. Battered T-wall units (after the Neel Company, 2004). A variety of architectural finishes are available. Since all the units are manufactured in the plant and hauled to the site, their fabrication is standardized and performed to shop-applied quality control procedures. The advantages of T-walls include: • • • • •

rapid construction; no skilled labor or specialized equipment is necessary for wall construction; requires smaller quantity of select backfill as compared to MSE walls; can be built on a batter; and no metallic components that could be subject to corrosion.


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The limitations of T-walls include: • • •

6.9.2

requires high quality granular backfill; a LRFD-based design methodology has not been developed for this system; and difficult to make height adjustments in the field as the units are prefabricated to specific heights. T-Wall Construction

Precast concrete units are shipped to site ready to be installed. The units are designed in accordance with AASHTO specifications for reinforced precast concrete with a minimum compressive strength at 28 days of 4,000 psi. In general, the concrete units are fabricated so that they are interchangeable. The units are unloaded using a small crane and a special lifting device provided by the T-Wall supplier. The basic steps for T-wall construction are similar to modular gravity and MSE walls except the specific steps listed below: •

Once the first course of concrete T-wall units is placed on the leveling pad, cover the vertical joint with filter fabric.

•

After compacting the select backfill material between and behind the stems, place the horizontal joint materials and the shear keys.

•

Continue placing units and backfilling. Normally a complete course is placed before beginning the next course.

6.9.3

T-Wall Cost

T-walls are cost effective to heights between 7 and 40 ft. Within the cost effective height range their cost per ft2 of exposed wall face is between $30 and $40, where the higher cost is associated with higher wall heights. 6.9.4

T-Wall Design

The design for a T-Wall system includes evaluation of external stability and internal stability, LRFD-based design computations for T-walls are not available in AASHTO (2007). For ASD, external stability computations are made by assuming that the system acts as a rigid body. Methods used to evaluate eccentricity, sliding, and bearing resistance for MSE walls can be applied to T-Walls. FHWA NHI-07-071 Earth Retaining Structures

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For internal stability evaluation, computations are made at each level to verify that the frictional forces gripping the stem (caused by pressure of the compacted, confined soil between the stems) exceeds any horizontal force acting on the face of the unit. The vertical joints in a T-wall act as slip planes, which can accommodate differential movements in the foundation. This flexibility can minimize the expense of foundation preparation.


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6-40


CHAPTER 7 MECHANICALLY STABILIZED EARTH WALLS 7.1

INTRODUCTION

Mechanically Stabilized Earth (MSE) walls are constructed by placing alternating layers of reinforcing elements and compacted backfill behind a facing. The soil and the structural elements act in unity to form a composite structure that constitute the wall. This composite structure is flexible and can generally accommodate relatively large horizontal and vertical movements without excessive structural distress. MSE walls are typically constructed in fill situations. MSE Wall is also used as a generic term that includes reinforced soil (a term used when multiple layers of inclusions act as reinforcement in soils placed as fill). Reinforced Earth is a trademark for a specific reinforced soil system. The modern methods of soil reinforcement for retaining wall construction were pioneered by the French architect and engineer Henri Vidal in the early 1960s. Since the late 1960s, MSE walls have been used increasingly worldwide. The first wall to use this technology in the U.S. was built in 1972 on California State Highway 39, northeast of Los Angeles. Since its first use in U.S., applications, design methods, and construction specifications for MSE walls have evolved significantly and are described in the following documents: •

AASHTO LRFD Bridge Design Specifications, 4th Edition (2007) Section 11.

•

Mechanically Stabilized Earth Walls and Reinforced Soil Slopes, Design and Construction Guidelines. FHWA-NHI-00-043, (Elias et al., 2001).

•

Corrosion/Degradation of Soil Reinforcements for Mechanically Stabilized Earth Walls and Reinforced Soil Slopes, FHWA-NHI-00-044, (Elias, 2000).

•

Geosynthetic Design and Construction Guidelines. FHWA NHI-07-092, (Holtz et al., 2007).

Another system referred to as “Shored Mechanically Stabilized Earth (SMSE)” walls combine features of an MSE wall and shoring walls (such as soil nail walls). The design of this system considers the long-term retaining benefits provided by the shoring wall, including reduction of lateral loads on the MSE wall mass and contributions to global stability. Detailed information about this system is provided in Shored Mechanically Stabilized Earth (SMSE) Wall Systems Design Guidelines, FHWA-CFL/TD-06-001 (Morrison et al., 2006). FHWA NHI-07-071 Earth Retaining Structures

7-1

7 – MSE Walls June 2008

7.2

PRINCIPAL COMPONENTS OF MSE WALLS

The principal components of an MSE wall consists of retained and select backfills, reinforcing elements within the select backfill, and a facing (Figure 7-1).

Figure 7-1. Principal Components of a Mechanically Stabilized Earth Wall (after Christopher, et al., 1990). Retained backfill is the fill material located between the mechanically stabilized soil mass and the natural soil. The select backfill is used to construct the mechanically stabilized earth mass shown in Figure 7-1 and is required for durability, good drainage, constructability, and good soil reinforcement interaction. Backfill material requirements for MSE walls are discussed in detail in Section 7.4.1. The reinforcing elements within the select backfill may consist of steel strips or bars, welded wire mats, grids, or geotextile sheets. Details of the reinforcing elements (i.e., geometry, materials, and extensibility) are presented in Section 7.4.2. Facing is not really part of the composite structure, but it prevents backfill erosion at the wall face. Common facings include precast concrete panels, dry cast modular blocks, metal sheets and plates, gabions, welded wire mesh, shotcrete, wood lagging and panels, and wrapped sheets of geosynthetics. Details of the facing elements are presented in Section 7.4.3. MSE walls do not require a structural foundation. A leveling pad of unreinforced concrete or gravel is used to serve as a guide for facing panel erection and is not intended as a structural foundation support. FHWA NHI-07-071 Earth Retaining Structures

7-2


Subsurface and surface drainage systems for MSE walls are discussed in Section 7.8.5. Table 7-1 lists many of the MSE wall systems currently available on the market. The primary differences between these systems are the materials, geometries and arrangements of the reinforcing elements, and the materials and details of the facing elements.

Table 7-1. Summary of Reinforcement and Face Panel Details for Selected MSE Wall Systems (after Elias et al., 2004). Reinforcement Detail

Typical Face Panel Detail1

Stabilized Earth Wall T&B Structural Systems LLC 6800 Manhattan Blvd, Suite 304 Ft. Worth Texas 76120

Galvanized welded steel wire mesh with W7 to W20 bars. Mesh width and spacing can vary. Epoxy-coated meshes also available.

Reinforced Earth® The Reinforced Earth Company 8614 Westwood Center Drive Suite 1100 Vienna, VA 22182-2233

Ribbed galvanized steel strips, 0.157 in. thick, 2 in. wide. Or galvanized steel ladder strips, W10 wire, two longitudinal wires and cross bars spaced at 6 in.

Retained Earth The Reinforced Earth Company 8614 Westwood Center Drive Suite 1100 Vienna, VA 22182-2233

Rectangular grid of W11, W15 or W20 galvanized steel wire, 24 x 6 in. grid. 2, 4, 5 or 6 longitudinal bars. Stainless steel mesh used in marine and corrosive environments.

Precast concrete panels 5 ft x 5 ft x 6 in. thick or 5 ft x 10 ft x 6 in. thick. Different size panels used at top and bottom to match project requirements. Cruciform and square shaped precast concrete 5 ft x 5 ft x 5.5 in. thick. Also rectangular shaped precast concrete 5 ft x 10 ft x 5.5 in. thick. Variable height panels used at top and bottom of wall. Hexagonal and square precast concrete 5 ft x 5 ft x 5.5 in. thick. Also rectangular shaped precast concrete 5 ft x 10 ft x 5.5 in. thick. Variable height panels used at top and bottom of wall.

Mechanically Stabilized Embankment Dept. of Transportation, Division of Engineering Services 5900 Folsom Blvd. P.O. Box 19128 Sacramento, CA 95819

Rectangular grid, nine 3/8 in. diameter plain steel bars on 24 x 6 in. grid. Two bar mats per panel (connected to the panel at four points).

Precast concrete; rectangular 12.5 ft long, 24 in. high, 8 in. thick.

ARES Tensar Earth Technologies 5883 Glenridge Drive, Suite 200 Atlanta, GA 30328

HDPE Geogrid

Precast concrete panel; rectangular 9 ft wide, 5 ft high, 5.5 in. thick.

Wire Faced Wall T&B Structural Systems LLC 6800 Manhattan Blvd Ste 304 Ft. Worth Texas 76120

4 ft wide welded steel wire mesh. Mesh is 8 in. x 12, 18 or 24 in., of W4.5 to W20 bars. Size and configuration are variable.

Welded steel wire mesh facing. Several veneer facing options available.

Welded Wire Wall Hilfiker Retaining Walls 1902 Hilfiker Lane Eureka, CA 95503

Welded steel wire mesh, Galvanized or Non-Galvanized. Mesh reinforcements vary in spacing and gauges to meet project design specifications.

Welded steel wire mesh, 2’ tall x 8’ wide typical. Backing mat, Hardware Cloth or Filter Fabric depending on project. (With geotextile or shotcrete, if desired).

System Name


7-3


Table 7-1 (cont’d). Summary of Reinforcement and Face Panel Details For Selected MSE Wall Systems Reinforcement Detail

Typical Face Panel Detail1

Reinforced Soil Embankment Hilfiker Retaining Walls 1902 Hilfiker Lane Eureka, CA 95503


Precast concrete unit 12.5 ft long, 24 in. high.

ArtWeld Gabions Hilfiker Retaining Walls 1902 Hilfiker Lane Eureka, CA 95503


ArtWeld Gabion baskets of various sizes and heights designed per project requirements.

Gabion Faced M.S.E. Hilfiker Retaining Walls 1902 Hilfiker Lane Eureka, CA 95503


ArtWeld Gabions of various sizes and heights connected to reinforcing mesh by spiral binders.

Eureka Reinforced Soil Hilfiker Retaining Walls 1902 Hilfiker Lane Eureka, CA 95503


Precast or cast-in-place concrete facing panels, shotcrete, sculpted shotcrete, or stacked stone.

Steepened Slope Hilfiker Retaining Walls 1902 Hilfiker Lane Eureka, CA 95503


Welded steel wire mesh, 1 to 1 slope typical. Hardware Cloth or Filter Fabric depending on project. (With geotextile or shotcrete, if desired).

INTER-LOK Atlantic Concrete Industries P.O. Box 129 Tullytown, PA 19007

0.63 or 0.75 in. reinforcing steel bars fitted with 5 x 10 x 0.4 in. anchor plates and connected to a keyplate, and galvanized after fabrication.

Precast concrete panel; cross-shaped 6 ft wide and 3 ft high, 8 and 10 in. thick.

Rectangular grid of W11 x W11 4 bars per grid.

Diamond shaped precast concrete units, 5 ft x 8 ft, 5.5 in. thick.

Rectangular 4 ft wide welded steel wire mesh of W7 to W20 steel bars.

Dry cast concrete block 8 in high x 16 in long x 12 in deep.

HDPE Geogrid

MESA HP (high performance), DOT OR Standard units (8 in. high by 18 in. long face, 10.8 in. nominal depth). (dry cast concrete)

Galvanized welded wire ladders. Size varies with design requirements.

Dry cast concrete units, 8 in. high, 16 in. nominal length at face, 10 in. nominal depth.

Omega The Reinforced Earth Company 8614 Westwood Center Drive Suite 1100 Vienna, VA 22182-2233

Geostraps composed of high tenacity polyester with polyethylene sheathing. Reinforcement used in marine and corrosive environments only.

Cruciform and square shaped precast concrete 5 ft x 5 ft x 5.5 in. thick. Also rectangular shaped precast concrete 5 ft x 10 ft x 5.5 in. thick. Variable height panels used at top and bottom of wall.

Terratrel™ The Reinforced Earth Company 8614 Westwood Center Drive Suite 1100 Vienna, VA 22182-2233 Maccaferri Terramesh System Maccaferri Gabions, Inc. 43A Governor Lane Blvd. Williamsport, MD 21795

Ribbed galvanized steel strips, 0.157 in. thick, 2 in. wide. Or, galvanized steel ladder strips or mesh. Size varies with design requirements.

Welded steel wire mesh with geotextile backing or stone fill at wall face.

Continuous sheets of galvanized double twisted woven wire mesh with PVC coating.

Rock filled gabion baskets laced to reinforcement.

System Name

ISOGRID Neel Co. 6520 Deepford Street Springfield, VA 22150 T-Block Wall System T&B Structural Systems LLC 6800 Manhattan Blvd Ste 304 Ft. Worth Texas 76120 MESA Tensar Earth Technologies, Inc. 5883 Glenridge Drive, Suite 200 Atlanta, GA 30328 Pyramid™ The Reinforced Earth Company 8614 Westwood Center Drive Suite 1100 Vienna, VA 22182-2233

Strengthened Earth Gifford-Hill & Co. 2515 McKinney Ave. Dallas, Texas 75201


3

Rectangular grid, W7, W9.5 and W14, transverse bars at 9 and 18 in.

7-4

Precast concrete units, rectangular or wing shaped, 6 ft x 7 ft x 5.5 in.


Table 7-1 (cont’d). Summary of Reinforcement and Face Panel Details For Selected MSE Wall Systems MSE Plus SSL 4740 Scotts Valley Drive Scotts Valley, CA 95066

Rectangular grid with 8 to W24 longitudinal bars and W8 transverse. Mesh may have 2 – 6 longitudinal bars spaced at 6 or 8 in.

KeySystem - Inextensible Galvanized welded wire ladder mat of Keystone Retaining Wall Systems W7.5 to W17 bars with crossbars at 6 – 24 4444 W. 78th Street in. Minneapolis, MN 55435 KeySystem - Extensible Stratagrid high-tenacity knit polyester Keystone Retaining Wall Systems geogrid soil reinforcement by Strata 4444 W. 78th Street Systems, Inc. PVC coated. Minneapolis, MN 55435 Tricon System Galvanized welded-wire. Tricon Precast Ltd. 15055 Henry Road Houston, TX 77060 Versa-Lok Retaining Wall Systems PVC coated PET or HDPE geogrids. 6348 Highway 36 Blvd. Oakdale, MN 55128 Anchor Wall Systems PVC coated PET geogrid. 5959 Baker Road Minnetonka, MN 55345 1 Additional facing types are possible with most systems.

7.3

Rectangular precast concrete panels 5 ft high, 5 and 12 ft wide, with a thickness of 6 or 7 in. KeySystem concrete facing unit is 8 in high x 18 in. wide x 12 in. deep (dry cast concrete). Keystone Standard and Compac concrete facing units are 8 in. high x 18 in. wide x 18 in. or 12 in. deep (dry cast concrete). Rectangular precast concrete panels with a face area of 45 sq. ft. Versa-Lok concrete unit 6 in. high x 16 in. long x 12 in. deep (dry cast concrete) Anchor Landmark concrete unit 15 in. high x 8 in. long x 12 (small unit) or 12.5 (large unit) in. deep (dry cast concrete).

APPLICATIONS OF MSE WALLS

Figure 7-2 illustrates various applications of MSE walls. Figure 7-3 shows a variety of constructed MSE walls. MSE walls have many advantages compared with conventional reinforced concrete and concrete gravity retaining walls, including: • • • • •

use simple and rapid construction procedures and do not require large construction equipment; require less site preparation; need less space in front of the structure for construction operations; do not need rigid, unyielding foundation support because MSE structures are tolerant to deformations; and are technically feasible and cost-effective to heights in excess of 100 ft (with 150 ft being achieved for tiered MSE wall at Seattle-Tacoma International Airport).


7-5


Figure 7-2. Examples of MSE Wall Applications (a) Retaining Wall; (b) Access Ramp; (c) Waterfront Structure; and (d) Bridge Abutment.


7-6


(a)

(b)

(c)

(d)

(e) Figure 7-3. Examples of MSE Walls. FHWA NHI-07-071 Earth Retaining Structures

7-7


The relatively small quantities of manufactured materials required, rapid construction, and competition among the developers of different proprietary systems have resulted in a cost reduction relative to other types of retaining walls. MSE walls are likely to be more economical than other wall systems for walls higher than about 10 ft or where deep foundations would be required for a CIP wall. Another advantage of MSE walls is their flexibility and capability to absorb deformations due to poor subsoil conditions in the foundations. Also, based on observations in seismically active zones, these structures have demonstrated a higher resistance to seismic loading than have rigid concrete structures. Precast concrete facing elements for MSE walls can be made with various shapes and textures (with little extra cost) for aesthetic considerations. Masonry units, timber, and gabions also can be used with advantage to blend into the environment. The following general limitations may be associated with MSE walls, although these limitations may be addressed as part of detailed design: • • •

•

they require a relatively large space behind the wall face to obtain enough wall width for internal and external stability; MSE walls require select granular fill; at sites where there is a lack of granular soils, the cost of importing suitable fill material may render the system uneconomical; suitable design criteria are required to address corrosion of steel reinforcing elements, deterioration of certain types of exposed facing elements such as geosynthetics by ultra violet rays, and potential degradation of polymer reinforcement in the ground; and the design of soil-reinforced systems often requires a shared design responsibility between material suppliers and owners, and greater input from agencies’ geotechnical specialists.

MSE walls should not be used under the following conditions (see AASHTO (2007) Section 11.10.1): • •

•

when utilities other than highway drainage must be constructed within the reinforced zone where future access for repair would require the reinforcement layers to be cut; with galvanized metallic reinforcements exposed to surface or ground water contaminated by acid mine drainage or other industrial pollutants, as indicted by low pH and high chlorides and sulfates; and when floodplain erosion may undermine the reinforced fill zone, or where the depth to scour cannot be reliably determined.


7-8


7.4

MATERIALS FOR MSE WALLS

7.4.1

Backfill Material

The backfill material for an MSE wall has a major impact on: • •

Short-term stability during construction; and Long-term stability and deformation of the completed structure.

MSE walls require high quality backfill for durability, good drainage, constructability, and good soil reinforcement interaction, which can be obtained from well graded, granular materials. Many MSE systems depend on friction between the reinforcing elements and the soil. In such cases, a material with high friction characteristics is specified and required. Some systems rely on passive pressure on reinforcing elements, and, in those cases, the quality of backfill is still critical. These performance requirements generally eliminate soils with high clay contents. From a reinforcement capacity point of view, lower quality backfills could be used for MSE structures; however, a high quality granular backfill has the advantages of being more free draining, providing better durability for metallic reinforcement, and requiring less reinforcement. There are also significant handling, placement, and compaction advantages in using granular soils. These include an increased rate of wall erection and improved maintenance of wall alignment tolerances. The following requirements are consistent with current practice: Select Granular Fill Material for the Reinforced Zone. All backfill material used in the structure volume for MSE structures shall be reasonably free from organic or other deleterious materials and shall conform to the following gradation limits as determined by AASHTO T-27. 1.

Gradation Limits Sieve Size 4 in.(a) No. 40 No. 200

Percent Passing 100 0-60 0-15

Plasticity Index (PI) shall not exceed 6. (a)

Cu should be greater than or equal to 4. As a result of recent research on construction survivability of geosynthetics and epoxy coated reinforcements, it is recommended that the maximum particle size for these materials be reduced to ¾-inch for geosynthetics and epoxy and PVC coated reinforcements FHWA NHI-07-071 Earth Retaining Structures

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unless tests are or have been performed to evaluate the extent of construction damage anticipated for the specific fill material and reinforcement combination.

It should be noted that granular fill containing even a few percent fines may not be free draining, and drainage requirements should always be carefully evaluated on an individual project basis. Information on drainage systems for MSE walls is provided in Section 7.8.5. 2.

Soundness. The materials shall be substantially free of shale or other soft, poor durability particles. The material shall have a magnesium sulfate soundness loss (or a sodium sulfate value less than 15% after five cycles) of less than 30% after four cycles. Testing shall be in accordance with AASHTO T-104.

The design of buried steel elements of MSE structures is predicated on backfills exhibiting minimum or maximum electrochemical index properties, and then designing the structure for maximum corrosion rates associated with these properties. These recommended index properties and their corresponding limits are shown in Table 7-2. Reinforced fill soils must meet the indicated criteria to be qualified for use in MSE construction using steel reinforcements.

Table 7-2. Recommended Limits of Electrochemical Properties for Select Backfills When Using Steel Reinforcement. Property Resistivity pH Chlorides Sulfates Organic Content

Criteria >3000 ohm-cm >5<10 <100 PPM <200 PPM 1% max.

Test Method AASHTO T-288-91 AASHTO T-289-91 AASHTO T-291-91 AASHTO T-290-91 AASHTO T-267-86

Where geosynthetic reinforcements are proposed to be used, the limits for electrochemical criteria would vary depending on the polymer. Limits based on current research are shown in Table 7-3.


7-10


Table 7-3. Recommended Limits of Electrochemical Properties For Select Backfills When Using Geosynthetic Reinforcements. Property

Criteria

Test Method

Polyester (PET)

pH

>3 and <9

AASHTO T-289-91

Polyolefin (PP & HDPE)

pH

>3

AASHTO T-289-91

Base Polymer

The select reinforced fill criteria outlined above represent materials that have been successfully used throughout the United States and have resulted in excellent wall performance. Peak shear strength parameters are used in the analysis. For MSE walls, a lower bound frictional strength of 34 degrees would be consistent with the specified fill, although some nearly uniform fine sands meeting the specifications limits may exhibit friction angles of 31 to 32 degrees. Higher values may be used if substantiated by laboratory direct shear or triaxial test results for the site-specific material used or proposed. However, extreme caution is advised for use of friction angles above 40 degrees for design, due to a lack of field performance data and questions concerning mobilization of shear strength above that value. Fill materials outside of these gradation and plasticity index requirements have been used successfully (when appropriate design and construction procedures have been implemented); however, problems including significant distortion and structural failure have also been observed. While there may be a significant savings in using lower quality backfill, property values must be carefully evaluated with respect to influence on both internal and external stability and constructability. For MSE walls constructed with reinforced fill containing more than 15% passing a #200 sieve and/or PI exceeding 6, both total and effective shear strength parameters should be evaluated in order to obtain an accurate assessment of horizontal stresses, sliding, compound failure (behind and through the reinforced zone), and the influence of drainage on the analysis. Both long-term and short-term pullout tests, as well as soil/reinforcement interface friction tests, should be performed. Settlement characteristics (both magnitude and time) must be carefully evaluated, especially in relation to downdrag stresses imposed on connections at the face and settlement of supported structures. Drainage requirements at the back, face, and beneath the reinforced zone must be carefully evaluated (e.g., use flow nets to evaluate influence of seepage forces and hydrostatic pressure). Also, it is more difficult to compact and moisture-condition more clayey soils (which may lead to longer required construction time).


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7.4.2 7.4.2.1

Reinforcing Elements Reinforcement Material

Distinction can be made between the characteristics of metallic and nonmetallic reinforcements: • •

7.4.2.2

Metallic reinforcements (Figure 7-4). Typically of mild steel. The steel is usually galvanized or may be epoxy coated. Nonmetallic reinforcements (Figure 7-5). Generally, polymeric materials consisting of polypropylene, polyethylene, or polyester.

Reinforcement Geometry

Three types of reinforcement geometry can be considered: • • •

7.4.2.3

Linear unidirectional. Strips, including smooth or ribbed steel strips, or coated geosynthetic strips over a load-carrying fiber. Composite unidirectional. Grids or bar mats characterized by grid spacing greater than 6 in. Planar bidirectional. Continuous sheets of geosynthetics, welded wire mesh, and woven wire mesh. The mesh is characterized by element spacing of less than 6 in.

Reinforcement Extensibility

There are two classes of extensibility: •

•

Inextensible. The deformation of the reinforcement at failure is much less than the deformability of the soil. Examples of inextensible reinforcement include steel reinforcements. Extensible. The deformation of the reinforcement at failure is comparable to or even greater than the deformability of the soil. Examples of extensible reinforcement include geosynthetic reinforcements.


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(a)

(b) Figure 7-4. Metallic Reinforcements (a) Ribbed Metal Strip and (b) Welded Bar Mat.


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(a)

(b) Figure 7-5. Geosynthetic Reinforcements (a) Geogrid and (b) Geotextile Facing.


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The performance and durability considerations for these two classes of reinforcement vary considerably and are detailed in FHWA-NHI-00-043, Mechanically Stabilized Earth Walls and Reinforced Steepened Slopes, Design and Construction Guidelines (Elias et al., 2001) and FHWA-NHI-00-044 Corrosion/Degradation of Soil Reinforcements for Mechanically Stabilized Earth Walls and Reinforced Soil Slopes (Elias, 2000). 7.4.2.4

Reinforcement Types

Most, although not all, systems using precast concrete panels use steel reinforcements that are typically galvanized, but may be epoxy coated. Two types of steel reinforcements are in current use: •

•

Steel strips. The currently commercially available strips are ribbed top and bottom, 2 in. wide and 0.156-in-thick. Smooth strips 2.375 to 4.75-in-wide, 0.125 to 0.156 in-thick have been used. Steel grids. Welded wire grid using 2 to 6 W7.5 to W24 longitudinal wire spaced at either 6 or 8 in. The transverse wire may vary from W11 to W20 and are spaced based on design requirements from 9 to 24 in. Welded steel wire mesh spaced at 2 by 2-in. (of thinner wire) has been used in conjunction with a welded wire facing. Some modular block wall (MBW) systems use steel grids with 2 longitudinal wires.

Most MBW systems use geosynthetic reinforcement, principally geogrids. The following types are widely used and available: • •

•

7.4.3

High Density Polyethylene (HDPE) geogrid. These are of uniaxial manufacture and are available in many styles. PVC coated polyester (PET) geogrid. Available from a number of manufacturers. They are characterized by bundled high-tenacity PET fibers in the longitudinal load carrying direction. For longevity, the PET is supplied as a high molecular weight fiber and is further characterized by a low carboxyl end group number. Geotextiles. High strength geotextiles can be used principally in connection with wrap-around wall construction. Both polyester (PET) and polypropylene (PP) geotextiles have been used. Wall Facing Elements

The types of facing elements used in the different MSE systems control their aesthetics because they are the only visible parts of the completed structure. A wide range of finishes and colors can be provided in the facing. In addition, the facing provides protection against FHWA NHI-07-071 Earth Retaining Structures

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backfill sloughing and erosion, and provides a drainage path in certain cases. The type of facing influences settlement tolerances. Figure 7-6 illustrates a variety of wall facings. •

•

Segmental precast concrete panels, summarized in Table 7-1 and illustrated in Figure 7-7. The precast concrete panels have a minimum thickness of 5.5 in. and are of a cruciform, square, rectangular, diamond, or hexagonal geometry. Temperature and tensile reinforcement are required, but will vary with the size of the panel. Vertically adjacent units are usually connected with shear pins. Dry cast MBW units. These are relatively small concrete units that have been specially designed and manufactured for retaining wall applications. The weight of these units commonly ranges from 30 to 110 lbs, with units of 75 to 110 lbs routinely used for highway projects. Unit heights typically range from 4 to 8 in. for the various manufacturers. Exposed face length usually varies from 8 to 18 in. Nominal width (dimension perpendicular to the wall face) of units typically ranges between 8 to 24 in. Units may be manufactured solid or with cores. Full height cores are filled with aggregate during erection. Units are normally dry-stacked (i.e., without mortar) and in a running bond configuration. Vertically adjacent units may be connected with shear pins, lips, or keys. They are referred to by trademarked names, such as Keystone®, Versa-Lok®, Allan®, etc. They are illustrated in Figure 7-8.

Major facing types are: •

•

•

•

Metallic Facings. The original Reinforced Earth® system had facing elements of galvanized steel sheet formed into half cylinders. Although precast concrete panels are now commonly used in Reinforced Earth walls, metallic facings may be appropriate in structures where difficult access or difficult handling requires lighter facing elements. Welded Wire Grids. Wire grid can be bent up at the front of the wall to form the wall face. This type of facing is used in the Hilfiker, Tensar, and Reinforced Earth wire retaining wall systems. Gabion Facing. Gabions (rock-filled wire baskets) can be used as facing with reinforcing elements consisting of welded wire mesh, welded bar-mats, geogrids, geotextiles, or the double-twisted woven mesh placed between or connected to the gabion baskets. Geosynthetic Facing. Various types of geotextile reinforcement are looped around at the facing to form the exposed face of the retaining wall. These faces are susceptible to ultraviolet light degradation, vandalism, and damage due to fire. Alternately, a geosynthetic grid used for soil reinforcement can be looped around to form the face of the completed retaining structure in a similar manner to welded wire


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Figure 7-6. Various Types of Wall Facing (after Wu, 1994).


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Figure 7-7. MSE Wall Surface Textures.

mesh and fabric facing. Vegetation can grow through the grid structure and can provide both ultraviolet light protection for the geogrid and a pleasing appearance. •

Post-Construction Facing. For wrapped faced walls, the facing, whether geotextile, geogrid, or wire mesh, can be attached after construction of the wall by shotcreting, guniting, cast-in-place, or attaching prefabricated facing panels made of concrete,


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wood, or other materials. This multi-staging facing approach adds cost, but is advantageous where significant settlement is anticipated. Precast elements can be cast in several shapes and provided with facing textures to match environmental requirements that blend aesthetically into the environment. Retaining structures using precast concrete elements as the facings can have surface finishes similar to any reinforced concrete structure. Retaining structures with metal facings have the disadvantage of shorter life because of corrosion unless provision is made to compensate for it. Facings using welded wire or gabions have the disadvantages of an uneven surface, exposed backfill materials, more tendency for erosion of the retained soil, possible shorter life from corrosion of the wires, and more susceptibility to vandalism. These disadvantages can be mitigated by providing shotcrete or by hanging facing panels on the exposed face and compensating for possible corrosion. The greatest advantages of such facings are low cost, ease of installation, design flexibility, good drainage (depending on the type of backfill), and possible treatment of the face for vegetative and other architectural effects. The facing can easily be adapted and well blended with natural country environment. These facings, as well as geosynthetic wrapped facings, are especially advantageous for construction of temporary or other structures with a short-term design life. Dry cast MBW facings may raise some concerns as to durability in aggressive freeze-thaw environments when produced with water absorption capacity significantly higher than that of wet-cast concrete. Historical data provide little insight, as their usage history is less than two decades. Further, because the cement is not completely hydrated during the dry cast process, (as is often evidenced by efflorescence on the surface of units), a highly alkaline regime may establish itself at or near the face area, and may limit the use of some geosynthetic products as reinforcements. Freeze-thaw durability is enhanced for products produced at higher compressive strengths and low water absorption ratios.

7.5

MSE WALL CONSTRUCTION

Construction of MSE walls involves placement of alternating layers of reinforcing elements and compacted backfill behind a wall facing. Although the construction sequence is basically the same for most MSE wall types, certain construction details may differ from one system to another due to differences in the reinforcing elements, the wall facings, the labor/equipment requirement, and the experience of the contractors. The following is an outline of the principal sequence of construction for MSE walls. Specific systems, special FHWA NHI-07-071 Earth Retaining Structures

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appurtenances, and specific project requirements may vary from the general sequence indicated. Details for such can be obtained from their vendors.

Figure 7-8. Examples of Commercially Available MBW Units (after Simac et al, 1996). FHWA NHI-07-071 Earth Retaining Structures

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7.5.1

Construction of MSE Walls with Precast Facings

A complete sequence of construction of MSE wall systems with a precast facing is illustrated in Figure 7-9 and a description of the primary construction steps is provided below: •

Preparation of subgrade. This step involves removal of unsuitable materials from the area to be occupied by the retaining structure. All organic matter, vegetation, slide debris, and other unstable materials should be removed and the subgrade compacted. In unstable foundation areas, ground improvement methods, such as dynamic compaction, stone columns, wick drains, or other foundation stabilization/ improvement methods, would be completed prior to wall erection.

•

Placement of a leveling pad for the erection of the facing elements. This generally unreinforced concrete pad is often only 1 ft wide and 6 in. thick and is used for MSE wall construction only, where concrete panels are subsequently erected. A gravel pad has been often substituted for MBW construction. The purpose of this pad is to serve as a guide for facing panel erection and is not intended as a structural foundation support.

•

Erection of the first row of facing panels on the prepared leveling pad. Facings may consist of either precast concrete panels, metal facing panels, or dry cast modular blocks. The first row of facing panels may be full, or half-height panels, depending upon the type of facing used. The first tier of panels must be temporarily braced to maintain stability and alignment. A geotextile filter is glued over the vertical and horizontal panel joints (with glue applied to both of the panels) to provide for drainage and prevent soil piping. For construction with modular dry-cast blocks, full sized blocks are used throughout with no shoring. A graded soil on geotextile filter is used between the gravel block or drain fill and the wall fill to prevent soil piping. The erection of facing panels and placement of the soil backfill proceed simultaneously.

•

Placement and compaction of backfill on the subgrade to the level of the


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first layer of reinforcement and its compaction. The fill should be compacted to the specified density, usually 95 to 100% of AASHTO T-99 maximum density and within the specified range of optimum moisture content. Compaction moisture contents dry of optimum are recommended. A key to good performance is consistent placement and compaction. Wall fill lift thickness must be controlled, based on specification requirements and vertical distribution of reinforcement elements. The uniform loose lift thickness of the reinforced backfill should not exceed 12 in. Place fill on the reinforcements away from the wall face and on top of the reinforcements to secure them in place. Windrow fill behind face panels to fill the gap left behind the bottom of the facing panels. Thinner lifts and lighter compaction equipment should be used within 3 ft of the back of the wall face. Retained fill placement behind the reinforced volume should proceed simultaneously.

7.5.2

•

Placement of the first layer of reinforcing elements on the backfill. The reinforcements are placed and connected to the facing panels when the compacted fill has been brought up to the level of the connection. They are generally placed perpendicular to back of the facing panels. Refer to Chapter 9 of FHWA NHI-00-043 Mechanically Stabilized Earth Walls and Reinforced Soil Slopes Design and Construction Guidelines (Elias et al., 2001) for more detailed construction controls procedures.

•

Placement of the backfill over the reinforcing elements to the level of the next reinforcement layer and compaction of the backfill. The previously outlined steps are repeated for each successive layer.

•

Construction of traffic barriers and copings. This final construction sequence is undertaken after the final panels have been placed, and the backfill has been completed to its final grade.

Construction of MSE Walls with Flexible Facings

Construction of flexible-faced MSE walls, where the reinforcing material also serves as facing material, is similar to that for walls with precast facing elements. For flexible facing types such as welded wire mesh, geotextiles, geogrids or gabions, the erection of the first level facing element requires only a level grade. A concrete footing or leveling pad is not usually required unless precast elements are to be attached to the system after construction. The construction steps for this type of wall are illustrated in Figures 7-10, and 7-11.


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(a)

(b)

(c)

(d)

(e)

(f)

Figure 7-9. Construction of MSE Wall With Inextensible Strip Reinforcements: (a) Concrete Leveling Pad Construction, (b) Erection of Facing Elements, (c) Placement of Reinforcements (d) Placement of Backfill, (e) Spreading of Backfill, and (f) Backfill Compaction. FHWA NHI-07-071 Earth Retaining Structures

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Figure 7-10. Construction Procedures of Geotextile Retained Earth Wall with Wrap-Around Facing.

Figure 7-11. Geotextile Retained Earth Wall Details. FHWA NHI-07-071 Earth Retaining Structures

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Construction proceeds as outlined for segmental facings with the following exceptions: •

Placement of first reinforcing layer. Reinforcement with anisotropic strength properties (i.e., many geosynthetics) should be placed with the principal strength direction perpendicular to face of structure. It is often convenient to unroll the reinforcement with the roll or machine direction parallel to the face. If this is done, then the cross machine tensile strength must be greater than the design tension requirements. Secure reinforcement with retaining pins to prevent movement during reinforced fill placement. Overlap adjacent sheets a minimum of 6 in. along the edges perpendicular to the face. Alternatively, with geogrid or wire mesh reinforcement, the edges may be butted and clipped or tied together.

•

Face Construction. Place the geosynthetic layers using face forms. For temporary support of forms at the face, form holders should be placed at the base of each layer at 4 ft horizontal intervals. These supports are essential for achieving good compaction. When using geogrids or wire mesh, it may be necessary to use a geotextile to retain the backfill material at the wall face. When compacting backfill within 3 ft of the wall face, a hand-operated vibratory compactor is recommended. The return-type method or successive layer tie method can be used for facing support. In the return method, the reinforcement is folded at the face over the backfill material, with a minimum return length of 4 ft to ensure adequate pullout resistance. Consistency in face construction and compaction is essential to produce a wrapped facing with satisfactory appearance. Apply facing treatment (shotcrete, precast facing panels, etc.).

Walls with geotextile sheet reinforcements or continuous mesh reinforcements (metallic or synthetic) can be constructed using the sheet or mesh as the wall facing element. 7.5.3

Backfill Placement and Compaction

Proper backfill placement and compaction is extremely important to the overall success of a MSE wall. Following are typical requirements for backfill placement:


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•

•

• •

•

•

• •

Backfill placement should closely follow erection of each course of facing units. Place fill on the reinforcements away from the wall face and on top of the reinforcements to secure them in place. Windrow fill behind face panels to fill the gap left behind the bottom of the facing panels. Backfill placement methods near the facing must assure that no voids exist directly beneath the reinforcing elements. In general, the backfill should be placed in such a manner as to avoid any damage to the wall materials or misalignment of the facing units or reinforcing elements. At no time should any construction equipment be in direct contact with the reinforcements because the reinforcements and their protective coatings can be damaged. The soil layers should be compacted up to or even slightly above the elevation of each level of reinforcement connections prior to placing that layer of reinforcements. Moisture and density control is imperative. Even when using high-quality granular materials, problems can occur if compaction control is not exercised. The backfill material should be placed and compacted at or within 2 percent dry of the optimum moisture content. If the reinforced fill is free draining with less than 5 percent fines passing the No. 200 U.S. sieve, water content of the fill may be within ±3 percent of the optimum. The moisture content during placement can have a significant effect on the soil-reinforcement interaction. Moisture contents wet of optimum makes it increasingly difficult to maintain an acceptable facing alignment, especially if the fines content is high. Moisture contents that are too dry result in significant settlement during periods of precipitation. A density of 95 percent of the maximum density as determined by AASHTO T-99, Method C or D (with oversize corrections as outlined in Note 7 of that test) is recommended for MSE walls. For applications where spread footings are used to support bridge or other structural loads, the top 5 ft below the footing elevation should be compacted to 100 percent AASHTO T-99. A procedural specification is preferable where a significant percentage of coarse material, generally 30 percent or greater retained on the ¾ in. sieve, prevents the use of AASHTO T-99 or T-180 test methods. In this situation, typically three to five passes with conventional vibratory roller compaction equipment is adequate to attain the maximum practical density. The actual requirements should be evaluated based on field trials. The maximum lift thickness after compaction should not exceed 12 in. The lift thickness may be decreased, if necessary, to obtain the specified density. With the exception of the 3 ft zone directly behind the facing elements, large smoothdrum, vibratory rollers should generally be used to obtain the desired compaction. Sheepsfoot rollers should not be permitted because of possible damage to the reinforcements. When compacting uniform medium to fine sands (in excess of 60


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•

•

•

7.6

percent passing the No. 40 sieve), a smooth-drum static roller or lightweight (walk behind) vibratory roller should be used. The use of large vibratory compaction equipment with this type of backfill material will make wall alignment control difficult. Within 3 ft of the wall, small, single or double drum, walk-behind vibratory rollers or vibratory plate compactors should be used (Figure 7-12). Placement of the reinforced backfill near the front should not lag behind backfill placement for the remainder of the structure by more than one lift. Poor fill placement and compaction in this area has in some cases resulted in a chimney-shaped vertical void immediately behind the facing elements. Within this 3 ft zone, quality control should be maintained by a methods specification such as three passes of a light drum compactor. Higher quality fill is sometimes used in this zone so that the desired properties can be achieved with less compactive effort. In some cases, the contractor should be required to use a rod to densify the coarse aggregate at the wall face. Excessive compactive effort or use of too heavy equipment near the wall face could result in excessive face panel movement (modular panels) or structural damage (full-height, precast panels), and overstressing of reinforcing materials. Inconsistent compaction and undercompaction caused by insufficient compactive effort or allowing the contractor to “compact” backfill with trucks and dozers will lead to gross misalignments and settlement problems and should not be permitted. At the end of each day’s operation, the last level of backfill should be sloped away from the wall facing to rapidly direct runoff away from the wall face. In addition, surface runoff from adjacent areas must not be allowed to enter the wall’s construction site.

MSE WALL COST

Site-specific costs of a soil-reinforced structure are a function of many factors, including cutfill requirements, wall/slope size and type, in-situ soil type, available backfill materials, facing finish, temporary or permanent application. It has been found that MSE walls with precast concrete facings are usually less expensive than reinforced concrete retaining walls for heights greater than about 10 ft and average foundation conditions. Modular block walls (MBW) are competitive with concrete walls at heights of less than 15 ft.


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(a)

(b) Figure 7-12. Lightweight Compaction Adjacent to Wall Facing. FHWA NHI-07-071 Earth Retaining Structures

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In general, the use of MSE walls results in savings on the order of 25 to 50% and possibly more in comparison with a conventional reinforced concrete retaining structure, especially when the latter is supported on a deep foundation system (poor foundation condition). A substantial savings is obtained by elimination of the deep foundations, which is usually possible because reinforced soil structures can accommodate relatively large total and differential settlements. Other cost saving features include ease of construction and speed of construction.

7.7

CONSTRUCTION INSPECTION OF MSE WALLS

Prior to construction of any MSE structure, personnel responsible for construction control should become familiar with the following items: • • • •

Plans, specifications and testing requirements; Site conditions relevant to construction conditions; Material requirements and allowable tolerances; and Construction sequencing.

Quality assurance measures must be implemented to: •

• •

Inspect reinforcement for damage and availability of mill test certificates to certify grade and corrosion protection, where required for metallic reinforcement and compliance with submitted manufacturer's specifications for geosynthetics; Inspect precast elements for damage and casting tolerances; and Inspect and test reinforced backfill for conformance to gradation and electrochemical properties.

During erection, construction control is focused on assuring that: • • • •

Leveling pads are correctly constructed with respect to elevation; Facing elements are erected within the specified vertical tolerances; The fill is compacted to the required uniform density within the required moisture content range; and The reinforcements are properly secured to the facing elements.

Detailed discussion on MSE wall construction and component inspection can be found in FHWA Manual on Mechanically Stabilized Earth Walls and Reinforced Steepened Slopes,


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Design and Construction Guidelines (Elias et al., 2001) and in the NHI Course Inspection of MSE Walls and RSS (see document FHWA-NHI-07-094). A checklist of general requirements for monitoring and inspecting MSE systems is provided in Table 7-4.

Table 7-4. MSE Wall Field Inspection Checklist CONTRACTOR SET UP Review Specifications and become familiar with - material requirements - construction procedures - soil compaction procedures - alignment tolerances - acceptance/rejection criteria Review construction Plans and become familiar with - construction sequence - corrosion protection systems - special placement to reduce damage - soil compaction restrictions - details for drainage requirements - details for utility construction - construction of slope face - contractor’s documents Review Contractor’s schedule Review material (retained and select backfills, reinforcement elements etc) requirements and approval submittals Review corrosion protection requirements of metallic units, ultraviolet light degradation protection of nonmetallic units, and construction sequence requirements for the reinforcement system Review select backfill property test results against the requirements Discuss anticipated ground conditions and potential problems with Contractor Check overall condition of Contractor’s equipment Confirm that the layouts, dimensions, and elevations of the leveling pads are consistent with Plans Review Contractor’s stockpile area Check site conditions and observe site accessibility, construction dewatering, and drainage features such as seeps, adjacent streams, etc. Check submittals of the Contractor for deficiencies and act accordingly before sending it to Engineer for review SUBGRADE Confirm stability of slopes and report to the Engineer any temporary excavation support systems being used Confirm limits of excavation and wall footing as shown in the Plans Conform all unsuitable materials are removed from the area to be occupied by the wall Confirm construction dewatering consistent with Specifications FHWA NHI-07-071 Earth Retaining Structures

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Confirm bedrock (or competent ground) according to the Specifications Conform leveling, alignment, and proof-rolling of the subgrade consistent with Specifications and Plans Confirm that facing pad construction is performed according to Specifications FACING AND REINFORCING ELEMENTS Confirm that facing elements meet the Specification requirements and reject them for: - compressive strength < specification requirements - imperfect molding - honey-combing - severe cracking, chipping or spalling - color of finish variation - out-of-tolerance dimensions - misaligned connections Inspect materials in batch of reinforcements to ensure that they are all the same Check reinforcement labels to verify whether or not they match certification documents Observe construction of the reinforcements to confirm that Contractor complies with Specification requirements Confirm that overlapping layers of reinforcement are separated by select backfill(1) Confirm that the reinforcement is turned up at the face of the wall(2) Confirm that minimum overlap is observed along the edges perpendicular to the wall face(2) Obtain test samples from the reinforcement elements according to the Specifications from randomly selected locations If possible, inspect reinforcement elements after placement of select backfill by constructing a trial installation or by removing a small section of select backfill for inspection Check all reinforcement and prefabricated facing elements against the initial approved shipment and collect additional test samples Observe and confirm the consistency of first row of facing elements with Plans Monitor facing alignment: - adjacent facing panel joints - precast face elements - wrapped face walls - line and grade Confirm that first tier of facing elements are braced to maintain stability and alignment(1) Confirm that facing elements are not placed on top of each other until they are completely backfilled(1) If geogrid is used for reinforcement, confirm that a wire mesh or geotextile is used at the wall face to retain select backfill(2) Confirm proper face protection is observed by the Contractor after placement of the final select backfill(2) Confirm the line and grade of the wall face according to Plans


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DRAINAGE SYSTEMS, INFILL AND BACKFIILL Confirm consistency of retained and select backfill materials based on Specifications Confirm placement and compaction of retained and select backfill materials according to Specifications Confirm alignment and elevation of internal drainage systems, and method of passing reinforcing elements around the drainage systems according to Plans Confirm alignment and elevation of external drainage systems according to Plans POST INSTALLATION Verify pay quantities Note: (1)MSE wall with strip reinforcement and (2)MSE wall with geotextile or mesh facing.

7.8

MSE WALL DESIGN

The procedure for design of MSE walls using LRFD is very similar to that followed using ASD. In LRFD, the external and internal stability of the MSE wall is evaluated at all appropriate strength limit states and overall stability and lateral/vertical wall movement are evaluated at the service limit state. Extreme event load combinations are used to design and analyze for conditions such as vehicle impact and seismic loading. The specific checks for the strength and service limit states required for MSE wall design are listed below. Strength Limit States for MSE walls •

External Stability (Figure 7-13) 1. Limiting Eccentricity; 2. Sliding; and 3. Bearing Resistance.

•

Internal Stability 1. 2. 3. 4.

Tensile Resistance of Reinforcement; Pullout Resistance of Reinforcement; Structural Resistance of Face Elements; and Structural Resistance of Face Element Connections.

Service Limit States for MSE walls 1. Overall Stability; and 2. Lateral and Vertical Wall Movements.


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Sliding

Limiting Eccentricity (Overturning)

Bearing

Figure 7-13. Potential External Failure Mechanisms for MSE Walls. The external stability of an MSE wall is evaluated assuming that the reinforced soil mass acts as a rigid body. This is because, when properly designed, the wall facing and the reinforced soil act as a coherent block with lateral earth pressures acting on the back face of that block. The internal stability of the reinforced soil mass is dependent on three fundamental characteristics: • • •

the soil-reinforcement interaction; the tensile strength of the reinforcement; and the durability of the reinforcing material.


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Therefore, the internal stability analyses of an MSE wall in LRFD is evaluated by (a) determining the maximum factored load in each reinforcement and (b) comparing this maximum factored load to the factored pullout resistance and to the factored tensile resistance of the reinforcement for all applicable strength, service, and extreme event limit states. The major design steps for MSE walls are summarized in Table 7-5.

Table 7-5. Design Steps for MSE Walls Step 1.

Establish project requirements including all geometry, loading conditions (transient and/or permanent, seismic, etc.), performance criteria, and construction constraints.

Step 2.

Evaluate existing topography, site subsurface conditions, in-situ soil/rock properties, and wall backfill parameters.

Step 3.

Based on initial wall geometry, estimate wall embedment depth and length of reinforcement.

Step 4.

Estimate unfactored loads.

Step 5.

Calculate factored loads for all appropriate limit states (external stability).

Step 6.

Check eccentricity.

Step 7.

Check sliding resistance.

Step 8.

Check bearing resistance.

Step 9.

Estimate critical failure surface based on reinforcement type (i.e., extensible or inextensible) for internal stability design at all appropriate strength limit states.

Step 10.

Calculate factored horizontal stress at each reinforcement level.

Step 11.

Calculate maximum factored tensile stress in each reinforcement.

Step 12.

Check reinforcement pullout resistance.

Step 13.

Calculate nominal long-term reinforcement design strength and check reinforcement tensile resistance.

Step 14.

Check overall stability at the service limit state. Revise design if necessary.

Step 15.

Estimate vertical and lateral wall movements at the service limit state. Revise design if necessary.

Step 16.

Design wall drainage systems.


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7.8.1 7.8.1.1

Initial Design Steps (Steps 1 through 3) Steps 1 and 2 – Establish Project Requirements and Evaluate Site Conditions

Herein, it is assumed that Step 1 has been completed and a MSE wall has been deemed appropriate for the project and Step 2 has been completed to establish site subsurface conditions, in-situ soil/rock properties, and wall backfill parameters. Guidelines for site evaluation for an MSE wall project (i.e., site exploration, field reconnaissance, subsurface exploration, and laboratory testing) can be found in FHWA-NHI-00-043 (Elias et al., 2001). The drainage and shear strength characteristics of the wall backfill soil are also assessed as part of Step 2. Guidelines for select backfill material gradation are provided in Section 7.4.1. 7.8.1.2

Step 3 – Select Initial Wall Geometry

Based on the required wall height, the length of the reinforcement and embedment depth of the wall are assumed. These assumptions are verified later in the external and internal stability checks. The design height of the MSE wall (for external stability computations) is the sum of the required embedment depth and the wall height. To prevent exceeding local bearing resistance, the minimum embedment depths shown in Table 7-6 should be used in the design unless the wall is constructed on a rock foundation. The wall embedment is provided as a function of the height of the structure (H) above the leveling pad. Final wall embedment for design is based on computations considering overall stability and bearing resistance (see AASHTO Section 11.10.2.2). Larger values for embedment depth may be required, depending on depth of frost penetration, shrinkage and swelling of foundation soils, seismic activity, potential for erosion, and scour. The final embedment depth should also be based on bearing resistance and overall stability computations (conducted in later design steps). The minimum wall embedment, in any case, should be 1.5 ft, except for structures founded on rock at the surface, where no embedment may be used. Alternatively, frost-susceptible soils could be over excavated and replaced with non frost susceptible backfill, hence reducing the overall wall height. Additional discussion on wall embedment depth and local bearing resistance for walls founded on slopes is provided in Section 11.10.2.2 of AASHTO (2007). The minimum reinforcement length (L) should be 0.7H (or 8 ft, whichever is greater), where H is the design height of the structure as measured from the top of the leveling pad (see Figure 7-14 for definition of H). MSE walls with sloping surcharge fills or other external FHWA NHI-07-071 Earth Retaining Structures

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loads, such as abutment footings or surcharges, generally require longer reinforcements for stability, often on the order of 0.8H to as much as 1.1H. Shorter lengths may be required for individual reinforcement levels. Shorter minimum reinforcement lengths, on the order of 6 ft, but no less than 70 percent of the wall height, can be considered if smaller compaction equipment is used, facing panel alignment can be maintained, and minimum requirements for wall external stability are met. Additional discussion is provided in Section C11.10.2.1 (AASHTO, 2007).

Table 7-6. Minimum Embedment Requirements for MSE Walls (after AASHTO, 2007) Slope in Front of Structure

Minimum Embedment

Horizontal (for walls) Horizontal (for abutments) 3H:1V (for walls) 2H:1V (for walls) 1.5H:1V (for walls)

H/20 H/10 H/10 H/7 H/5

Notes: Near rivers choose minimum embedment depth 2 ft below maximum scour depth.

A minimum horizontal bench width of 4 ft (measured from bottom of wall horizontally to slope face) should be provided in front of walls founded on slopes. This minimum bench width is required to protect against local instability near the toe of the wall. 7.8.2 7.8.2.1

Strength Limit States – External Stability (Steps 4 through 8) Step 4 – Estimate Unfactored Loads

The primary sources of loading for an MSE wall will be from earth pressures resulting from ground behind the reinforced zone and from surcharge loadings at the ground surface (e.g., vehicle live loads, line loads from a bridge seat, etc.). The unfactored loads for MSE walls may include loads due to horizontal earth pressure (EH), vertical earth pressure (EV), live load surcharge (LS), and earth surcharge (ES). Water, seismic, and vehicle impact loads should also be evaluated. Estimation of earth pressures on MSE walls for three different conditions (i.e., horizontal backslope with traffic surcharge, sloping backslope, and broken backslope) are presented below.


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Assumed for bearing resistance

Assumed for sliding and eccentricity

Figure 7-14. Pressure Diagram for MSE Walls With Horizontal Backslope and Traffic Surcharge (after AASHTO, 2007).

Horizontal backslope with traffic surcharge Figure 7-14 shows the procedure to estimate the earth pressure acting on the back of the reinforced zone for the case of a horizontal backslope with traffic surcharge. The active earth pressure coefficient (Ka) for vertical walls (i.e., walls with less than an 8 degree batter) with horizontal backslope is calculated from a simplified version of the Coulomb equation as presented in AASHTO (2007) Equation C11.10.6.2.1-1 and shown below:  φ'f K a = tan 2  45 − 2 

where:

  

(7-1)

φ’f = friction angle of retained fill.

Figure 7-15 illustrates the typical application of live load surcharge over the retained fill for an MSE wall in LRFD. When a live load surcharge is considered in design, the factored surcharge force is generally included over the backfill immediately above the wall only for evaluation of foundation bearing resistance and tensile resistance of the reinforcement FHWA NHI-07-071 Earth Retaining Structures

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Figure 7-15. Typical application of live load surcharge for MSE walls (after AASHTO, 2007). elements. The live load surcharge is not included over the backfill for evaluation of eccentricity, sliding, reinforcement pullout, or other failure mechanisms for which the surcharge load would increase resistance to failure. In addition to the uniform surcharge loads shown in Figure 7-14, any concentrated vertical and horizontal loads at the surface may influence both the internal and external stability of the wall. The stresses from concentrated vertical and horizontal loads can be calculated using simplified approaches illustrated in Figures 7-16 and 7-17 respectively. Sloping backslope Figure 7-18 shows the procedure to estimate the earth pressure acting on the back of the reinforced zone for the case of a sloping backslope. The active earth pressure coefficient (Ka) to determine the pressure on vertical walls with a surcharge slope is calculated as shown below: Ka =


2 sin (θ + φ ′) Γ sin 2 θ sin (θ - δ )

7-38

(7-2)


feet

Figure 7-16. Distribution of Stress from Concentrated Vertical Load (after AASHTO, 2007)


7-39


Figure 7-17. Distribution of Stress from Concentrated Horizontal Loads for External and Internal Stability Calculations (after AASHTO, 2007).


7-40


where:  sin (φ ′ + δ ) sin (φ ′ - β )  Γ = 1 +  sin (θ - δ ) sin (θ + β )  

2

β = Nominal slope of backfill behind wall (deg) δ = Angle of wall friction (deg)

φ′ = friction angle of retained fill θ = 90o for vertical wall The earth pressure force, (Po) in Figure 7-18, is oriented at the same angle as the backslope. This orientation is not due to wall friction because wall friction is assumed to be zero for this case. Broken backslope Figure 7-19 shows the procedure to estimate the earth pressure acting on the back of the reinforced zone for the case of a broken backslope. The active earth pressure coefficient (Ka) for this condition is computed using Equation 7-2. It should be kept in mind that soil components (e.g., retained backfill, reinforced (i.e., select) backfill) may have different soil parameters. 7.8.2.2

Step 5 – Calculate Factored Loads

In this design step, the unfactored loads from Step 4 are multiplied by load factors to obtain the factored loads for each appropriate limit state. The load factors for the limit states are provided in Table 4-1. It is specifically noted that, in certain states, the Strength II limit state is more critical than the Strength I limit state because owner-prescribed legal loads are greater than those provided in the AASHTO specifications. Load factors for permanent loads are selected to produce the maximum destabilizing effect for the design check being considered. For example, to produce the maximum destabilizing effect, when checking sliding resistance, γEV is selected as the minimum value from Table 42 (i.e., γEV = 1.00) and when checking bearing resistance, γEV is selected as the maximum value from Table 4-2 (i.e., γEV = 1.35).


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h–H

h H

Retained Fill

β

Pv

Reinforced Soil Mass

Po β

Sv

Ph

Le h/3

Figure 7-18. Pressure Diagram for MSE Walls With Sloping Backslope (after AASHTO, 2007). 2H β

h–H

Retained Fill

β

h H

Reinforced Soil Mass

Po

Pv

β Ph

h/3

Figure 7-19. Pressure Diagram for MSE Walls With Broken Backslope (after AASHTO, 2007).


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7.8.2.3

Step 6 – Check Eccentricity

The eccentricity of the wall (eB) can be calculated for each load group as:

eB =

B − Xo 2

(7-3)

where: B = base width (length of reinforcement elements) and Xo = location of the resultant from the toe of the wall. The parameter Xo is calculated as: Xo =

(M EV − M HTOT ) PEV

(7-4)

where: MEV = resisting moment due to factored vertical earth pressure calculated about the toe of the wall; MHTOT = driving moment due to factored horizontal earth pressure from ground and from factored live load surcharge calculated about the toe of the wall; and PEV = factored resultant force from vertical earth pressure due to the weight of reinforced soil.

MEV and MHTOT are calculated using the factored loads. Example calculations for two different backslopes are given below: For horizontal backslope with traffic surcharge (Figure 7-20) condition: M EV = PEV

L 2

H  H  M HTOT =  PEH  +  PLSH  3  2 

where

PEH = factored resultant force from horizontal earth pressure; and PLSH = factored resultant force from horizontal component of surcharge load.


7-43


PLSH PEH PEV

Figure 7-20. Calculation of Eccentricity for Horizontal Backslope with Traffic Surcharge Condition (after Elias and Christopher 1996).

It should be noted that the effect of external loadings on the MSE mass, which increases sliding resistance, should only be included if the loadings are permanent. For example, live load traffic surcharges should be excluded. For sloping backslope (Figure 7-21) condition: M EV = PEV 1

2L L + PEV 2 + PEH sin β L 2 3

M HTOT = PEH cos β

where

h 3

PEV2 = factored resultant force from earth pressure due to the weight of soil of sloping backslope; PEH sin β = factored resultant force from vertical component of the earth Pressure; and PEH cos β = factored resultant force from horizontal component of the earth pressure.


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PEV2

P EH

PEV1

Figure 7-21. Calculation of Eccentricity for Sloping Backslope Condition (after Elias and Christopher 1996). For eccentricity to be considered acceptable, the calculated location of the resultant vertical force (based on factored loads) should be within the middle one-half of the base width for soil foundations (i.e., emax = B / 4) and middle three-fourths of the base width for rock foundations (i.e., . emax = 3/8 B). Therefore, for each load group, eB must be less than emax. If eB is greater than emax, a longer length of reinforcement is required. 7.8.2.4 Step 7 – Check Sliding Resistance Sliding and overall stability usually govern the design of structures greater than about 30 ft high, structures constructed on weak foundation soils, or structures with a sloping surcharge. The live load surcharge is not considered as a stabilizing force when checking sliding. The driving forces in a sliding evaluation will generally include factored horizontal loads due to earth, water, seismic, and surcharge pressures and the resisting force is provided by the minimum shear resistance between the base of the MSE wall and foundation soil. FHWA NHI-07-071 Earth Retaining Structures

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Sliding resistance along the base of the wall is evaluated using the same procedures in AASHTO (2007) Section 10.6.3.4 for spread footings on soil. The factored resistance against failure by sliding (RR) can be estimated by: RR = φ Rn= φτ Rτ + φep Rep

(7-5)

where: Rn = nominal sliding resistance against failure by sliding;

φτ = resistance factor for shear resistance between soil and foundation (provided in Table 10.5.5.2.2-1 of AASHTO (2007) as 1.0 for sliding of soil-on-soil); Rτ = nominal sliding resistance between soil and foundation;

φep = resistance factor for passive resistance (provided in Table 10.5.5.2.2-1 of AASHTO (2007)); and Rep = nominal passive resistance of soil available throughout design life It should be noted that any passive resistance provided by soil at the toe of the wall by embedment (i.e., φep Rep ) is ignored due to the potential for the soil to be removed through natural or manmade processes during the service life of the structure. The shear strength of the facing system is also conservatively neglected in most cases. If the soil beneath the wall is cohesionless, the nominal sliding resistance between soil and foundation is:

Rτ = PEV tan δ

(7-6)

where PEV = minimum factored vertical load for the strength limit state being considered; and δ = coefficient of sliding friction at the base of the reinforced soil mass.

For continuous sheet reinforcement, δ is selected as the minimum of: (1) friction angle of reinforced fill; (2) friction angle of foundation soil; or (3) interface friction angle between the reinforcement and soil. FHWA NHI-07-071 Earth Retaining Structures

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For discontinuous reinforcement (e.g., geogrid), δ is selected as the minimum value of (1) or (2). Sliding resistance of the MSE wall is considered adequate if RR is equal to or greater than the maximum factored horizontal earth pressure force from the ground and from the factored live load surcharge calculated in Step 5. 7.8.2.5 Step 8 – Check Bearing Resistance Due to the flexibility of MSE walls and the inability of the flexible reinforcement to transmit moment, a uniform base pressure distribution is assumed over an equivalent footing width. Unlike the bearing resistance check for CIP walls founded on rock, the assumption of a uniform base pressure is used for MSE walls founded on rock (see Section 11.10.5.4 of AASHTO (2007)). The effect of eccentricity, load inclination, and live load surcharges must be included in this check. Effects of live load surcharges are included because they increase the loading on foundation. The factored bearing resistance (qR) is given as:

q R = φ qn

(7-7)

where:

φ = resistance factor (see AASHTO (2007) Table 11.5.6-1) (for walls with flexible footings such as MSE walls, this factor is 0.65 in AASHTO (2007)); and qn = nominal bearing resistance (see AASHTO (2007) Equation 10.6.3.1.1.2a1). The design example provided in Section 7.9 illustrates the evaluation of qn for MSE walls. To check whether the bearing resistance of the MSE wall is adequate, the qR computed in Equation 7-7 is compared against the following criterion: q R ≥ quniform

where:

quniform = vertical stress for walls on soil foundations, which is calculated assuming a uniform distribution of pressure over an effective base FHWA NHI-07-071 Earth Retaining Structures

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width (B’ = B – 2e) quniform =

VTOT B − 2e B

(7-8)

where:

VTOT = sum of all factored vertical forces acting at the base of the wall (e.g., weight of reinforced fill, live and dead load surcharges); B = base width (length of reinforcement strips); and eB = eccentricity determined from Equation 7-4; however for bearing resistance calculations, Xo is defined as:

Xo =

(M VTOT − M HTOT ) VTOT

where:

MVTOT = resisting moment due to factored total vertical load based on earth pressure and live load surcharge calculated about the toe of the wall. Design example in Section 7.9 illustrates the calculation of MVTOT for horizontal backslope with traffic surcharge condition. MHTOT = driving moment due to factored lateral load based on earth pressure and live load surcharge calculated about the toe of the wall. The calculation of MHTOT is provided in Section 7.8.2.3. 7.8.3

Strength Limit States – Internal Stability (Steps 9 through 13)

To be internally stable, the MSE structure must be coherent and self supporting under the action of its own weight and any externally applied forces. This is accomplished through stress transfer from the soil to the reinforcement. This interaction between the soil and reinforcement improves the tensile properties and creates a composite material with the following characteristics:

• •

Stress transfer between the soil and reinforcement takes place continuously along the reinforcement; and Reinforcements are distributed throughout the soil mass with a degree of regularity and must not be localized.


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Figure 7-22 illustrates the internal failure mechanisms for MSE walls. At each reinforcement level, the reinforcement must be sized and spaced to preclude rupture under the stress it is required to carry and to prevent pullout from the soil mass. The process of sizing and designing to preclude internal failure, therefore, consists of determining the maximum developed tension forces in the reinforcements (i.e., maximum load), the location of this load along a critical failure surface, and comparing this maximum load against the resistance provided by the reinforcements both in pullout and tensile strength.

(a)

(b)

Figure 7-22. Mechanisms of Internal Failure in MSE Walls (a) Tension Failure and (b) Pullout Failure (after Christopher et al., 1990).

7.8.3.1 Step 9 – Select Location of Critical Failure Surface The location and shape of the theoretical critical failure surface is dependent on the type of reinforcement. Therefore, at this stage of the design, first the reinforcement type must be selected and categorized (i.e., extensible or inextensible) and then the potential critical failure surface should be determined accordingly. When inextensible reinforcements are used, the soil deforms more than the reinforcement. Therefore, the soil strength in this case is measured at low strain. The critical failure surface for this reinforcement type is determined by dividing the reinforced zone into active and resistant zones with a bilinear failure surface as shown in Figure 7-23a.


7-49


Figure 7-23. Location of Potential Failure Surface for Internal Stability Design of MSE Walls (a) Inextensible Reinforcements and (b) Extensible Reinforcements (after Elias et al., 2001). FHWA NHI-07-071 Earth Retaining Structures

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When extensible reinforcements are used, the reinforcement deforms more than the soil. Therefore, it is assumed that the shear strength of the reinforced fill is fully mobilized (residual strength) and active lateral earth pressures are developed. As a result, the critical failure surface for both horizontal and sloping backfill conditions are represented by the Rankine active earth pressure zone as shown in Figure 7-23b. 7.8.3.2 Step 10 – Calculated Factored Horizontal Stress The purpose of this design step is to calculate the maximum factored horizontal stress. It is specifically noted that load factors are typically applied to unfactored loads, not to an unfactored stress (as it is in Eq. 7-9) below. The AASHTO (2007) LRFD code, however, applies the “load” factor to the unfactored “stress” for this particular design calculation. The factored horizontal stress (σH) at each reinforcement level is based on AASHTO (2007) Equation 11.10.6.2.1-1:

σ H = γ P (σ V k r + ∆σ H )

(7-9)

where:

γP = maximum load factor for vertical earth pressure (EV), (1.35 per AASHTO (2007) Table 3.4.1-2); kr = lateral earth pressure coefficient; σV = pressure due to resultant of gravity forces from soil self weight within and immediately above the reinforced wall backfill, and any backfill present; and ∆σH = horizontal stress at reinforcement level resulting from a concentrated horizontal surcharge load (see AASHTO (2007) Section 11.10.10.1)). Research studies have indicated that the maximum tensile force is primarily related to the type of reinforcement in the MSE mass, which, in turn, is a function of the modulus extensibility, and density of reinforcement. Based on this research, a relationship between the type of the reinforcement and the overburden stress has been developed and is shown in (Figure 7-24). Figure 7-24 was prepared by back analysis of the lateral stress ratio from available field data where stresses in the reinforcements have been measured and normalized as a function of an active earth pressure coefficient. The lines shown on this figure correspond to usual values representative of the specific reinforcement systems that are known to give satisfactory FHWA NHI-07-071 Earth Retaining Structures

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20 ft

kr =

K Ka Ka

Figure 7-24. Variation of the Coefficient of Lateral Stress Ratio (K/Ka) with Depth in a Mechanically Stabilized Earth Wall (after AASHTO, 2007).

results, assuming that the vertical stress is equal to the weight of the overburden (gH). This provides a simplified evaluation method for all cohesionless reinforced fill walls. Future data may lead to modifications in Figure 7-24, including relationships for newly developed reinforcement types, effect of full-height panels, etc. For a vertical wall face (i.e., batters less than 8 degrees from vertical), the active earth pressure coefficient (Ka) is determined using Equation 7-1. For wall face batters equal to or greater than 8 degrees from the vertical, the following simplified form of Coulomb equation as presented in AASHTO (2007) Equation C11.10.6.2.1-2 and shown below can be used:


7-52


Ka =

sin 2 (θ + φ ' )  sin φ '  sin θ 1 +   sin θ 

(7-10) 2

3

where:

θ = inclination of the back of the facing as measured from the horizontal starting in front of the wall; and

φ′ = friction angle of retained fill. The value of Ka in the reinforced soil mass is assumed to be independent of all external loads, even sloping fills. If testing of the site-specific select backfill is not available, the value of φf used to compute the horizontal stress within the reinforced soil mass should not exceed 34o. Once the value of Ka is known, the lateral earth pressure coefficient (kr) (in Equation 7-9) that is used to compute sH at each reinforcement level is calculated as: kr = K/Ka (from Figure 7-24) x Ka (from Equation 7-10) This calculation is shown in Student Exercise 5 and the design example in Section 7.9. If present, surcharge load should be added into the estimation of σv. For sloping soil surfaces above the MSE wall section, the actual surcharge is replaced by a uniform surcharge equal to half of the height of the slope at the back of the reinforcements (Elias et al., 2001). For cases where concentrated vertical loads occur, refer to Figures 7-16 and 7-17 for computation of σv. 7.8.3.3 Step 11 – Calculate Maximum Factored Tensile Stress The maximum tension in each reinforcement layer per unit width of wall (Tmax) based on the reinforcement vertical spacing (Sv) is calculated as:

Tmax = σ H . S v

(7-11a)

where σH is the factored horizontal load calculated in step 10 using Equation 7-9.

Tmax may also be calculated at each level for discrete reinforcements (metal strips, bar mats, grids, etc.) per a defined unit width of reinforcement as: FHWA NHI-07-071 Earth Retaining Structures

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Tmax − R =

σ H . Sv Rc

(7-11b)

where, Rc = reinforcement coverage ratio = b/Sh (see Figure 7-25) (e.g., Rc = 1 for full coverage reinforcement); b = gross width of the reinforcing element; Sh = center-to-center horizontal spacing between reinforcements 7.8.3.4 Step 12 – Check Reinforcement Pullout Resistance The purpose of this design step is to check the pullout resistance of the reinforcements. The resistance develops after the stress transfer between the soil and the reinforcement takes place, which occurs through two mechanisms:

(1) friction along the soil-reinforcement interface (Figure 7-26a), and (2) passive soil resistance or lateral bearing capacity developed along the transverse sections of the reinforcement (Figure 7-26b). Stresses are transferred between soil and reinforcement by friction and/or passive resistance depending on reinforcement geometry. Friction develops at locations where there is a relative shear displacement and corresponding shear stress between soil and reinforcement surface. Reinforcing elements where friction is important should be aligned with the direction of soil reinforcement relative movement. Examples of such reinforcing elements are steel strips, longitudinal bars in grids, geotextile, and some geogrid layers. Passive resistance occurs through the development of bearing type stresses on "transverse" reinforcement surfaces normal to the direction of soil reinforcement relative movement. Passive resistance is generally considered to be the primary interaction for rigid geogrids, bar mat, and wire mesh reinforcements. The transverse ridges on "ribbed" strip reinforcement also provide some passive resistance.


7-54


Metal strips tc

Metal bars or wires

continuous reinforcement = Sh = b

(a)

Discontinuous geosynthetic sheets

Continuous geosynthetic sheets

(b) Figure 7-25. Definitions of b, Sh, and Sv for (a) metal and (b) geosynthetic reinforcements (after AASHTO, 2007). FHWA NHI-07-071 Earth Retaining Structures

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Normal Pressure

Pull Out Force

Frictional Force

Normal Pressure

(a)

Frictional Resistance Frictional Resistance

Passive Resistance

Pull Out Force

Pull Out Force Normal Pressure Passive Resistance

(b)

Figure 7-26. Mechanisms of Pullout Resistance. (a) By Friction, (b) By Passive Resistance (after Christopher et al., 1990 and Lawson, 1992).

The contribution of each transfer mechanism for a particular reinforcement will depend on the roughness of the surface (skin friction), normal effective stress, grid opening dimensions, thickness of the transverse members, and elongation characteristics of the reinforcement. FHWA NHI-07-071 Earth Retaining Structures

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Equally important for interaction development are the soil characteristics, including grain size and grain size distribution, particle shape, density, water content, cohesion, and stiffness. The primary function of reinforcements is to restrain soil deformations. In doing so, stresses are transferred from the soil to the reinforcement. These stresses are carried by the reinforcement in two ways: in tension or in shear and bending. The unfactored pullout resistance (Pr) of the reinforcement per unit width of reinforcement is estimated as: P r = F * α σv

(7-12)

Parameters in Equation 7-12 are defined in Equation 7-14. In this design step, the reinforcement pullout resistance is evaluated at each reinforcement level of the MSE wall. The required total length for reinforcement to generate appropriate pullout resistance for each level is calculated and then compared against the total reinforcement length initially estimated in design step 3 (Section 7.8.1.2). The initially estimated total reinforcement length may have to be adjusted based on the required length calculated in this step. The total length of reinforcement (L) required for internal stability is determined as: L = Le + La

(7-13)

where: Le = required length of reinforcement in resisting zone (i.e., beyond the potential failure surface) and La = remainder length of the reinforcement.

Estimating Le The length of reinforcement in the resisting zone (Le) is determined using the following equation (see AASHTO (2007) Equation 11.10.6.3.2-1): Le ≥


Tmax

ϕ F α σ v C Rc *

7-57

(7-14)


where: Tmax = maximum factored tensile load in the reinforcement (calculated in step 11);

ϕ F*

α σv C Rc

= resistance factor for reinforcement pullout = 0.9 (see AASHTO (2007) Table 11.5.6-1); = pullout friction factor (discussed below) Pullout = scale effect correction factor (discussed below) Resistance (Pr) = unfactored vertical stress at the reinforcement level in the resistant zone; = overall reinforcement surface area geometry factor; and (2 for strip, grid, and sheet-type reinforcement) = reinforcement coverage ratio as defined in Equation 7-11b.

Correction Factor (α) The correction factor (α) depends primarily upon the strain softening of the compacted granular backfill material, the extensibility, and the length of the reinforcement. Typical values of α based on reinforcement type are presented in Table 7-7. For inextensible reinforcement, α is approximately 1, but it can be substantially smaller than 1 for extensible reinforcements. The α factor can be obtained from pullout tests on reinforcements with different lengths or derived using analytical or numerical load transfer models which have been "calibrated" through numerical test simulations. In the absence of test data, α = 0.8 for geogrids and α = 0.6 for geotextiles (extensible sheets) is recommended (Elias et al., 2001). Table 7-7. Typical values for α (after AASHTO, 2007) Reinforcement type

α

All Steel Reinforcements

1.0

Geogrids

0.8

Geotextiles

0.6

Pullout Friction Factor (F*) The pullout friction factor can be obtained most accurately from laboratory or field pullout tests performed with the specific material to be used on the project (i.e., select backfill and reinforcement). Alternatively, F* can be derived from empirical or theoretical relationships developed for each soil-reinforcement interaction mechanism and provided by the FHWA NHI-07-071 Earth Retaining Structures

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reinforcement supplier. equation:

For any reinforcement, F* can be estimated using the general

F*

= Passive Resistance + Frictional Resistance

F*

= Fq . αβ + tan ρ

or (7-15)

where: Fq = the embedment (or surcharge) bearing capacity factor; αβ = a bearing factor for passive resistance which is based on the thickness per unit width of the bearing member; and ρ = the soil-reinforcement interaction friction angle. Equation 7-15 represents systems that have both the frictional and passive resistance components of the pullout resistance. In certain systems, however, one component is much smaller than the other and can be neglected for practical purposes. In absence of site-specific pullout testing data, it is reasonable to use these semi-empirical relationships in conjunction with the standard specifications for backfill to provide a conservative evaluation of pullout resistance. For steel ribbed reinforcement, F* is commonly estimated as: F* = tan ρ = 1.2 + log Cu at the top of the structure = 2.0 maximum

(7-16)

F* = tan φ at a depth of 20 ft. and below

(7-17)

where: ρ = interface friction angle mobilized along the reinforcement; f = wall fill peak friction angle; and Cu = uniformity coefficient of the backfill (D60/D10). If the specific Cu for the wall backfill is unknown during design, a Cu of 4 should be assumed (i.e., F* = 1.8 at the top of the wall), for backfills meeting the requirements of Section 7.4.1. For steel grid reinforcements with transverse spacing (St) > 6 in., F* is a function of a bearing or embedment factor (Fq), applied over the contributing bearing factor (αβ), as follows: FHWA NHI-07-071 Earth Retaining Structures

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F* = Fq αβ = 40 αβ = 40 (t/2St) = 20 (t/St) at the top of the structure

(7-18)

F* = Fq αβ = 20 αβ = 20 (t/2St) = 10 (t/St) at a depth of 20 ft and below

(7-19)

where t is the thickness of the transverse bar. St shall be uniform throughout the length of the reinforcement, rather than having transverse grid members concentrated only in the resistant zone. For geosynthetic (i.e., geogrid and geotextile) sheet reinforcement, the pullout resistance is based on a reduction in the available soil friction with the reduction factor often referred to as an interaction factor (Ci). In the absence of test data, the F* value for geosynthetic reinforcement should conservatively be estimated as: F* = 0.67 tan f

(7-20)

When used in the above relationships, f is the peak friction angle of the soil which, for MSE walls using select granular backfill, is taken as 34 degrees unless project specific test data substantiates higher values. The relationship between F* and depth below the top of wall for different reinforcement types is summarized in Figure 7-27.

Estimating La The La is obtained from Figure 7-23 for simple structures not supporting concentrated external loads such as bridge abutments. Based on this figure, the following relationships can be obtained for La:

•

For MSE walls with extensible reinforcement, vertical face, and horizontal backfill: La = (H - Z) tan (45 - φ́/2)

(7-21)

where Z is the depth to the reinforcement level.

•

For walls with inextensible reinforcement from the base up to H/2: La = 0.6 (H-Z)

•

(7-22)

For the upper half of a wall with inextensible reinforcements: La = 0.3H


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(7-23) 7 – MSE Walls June 2008

Figure 7-27. Typical Values for F* (after AASHTO, 2007). For ease of construction, based on the maximum total length required, a final uniform reinforcement length is commonly chosen. However, if internal stability controls the length, it could be varied from the base, increasing with the height of the wall to the maximum length requirement based on a combination of internal and maximum external stability requirements. See Chapter 5 of Elias et al. (2001) for additional guidance.

Pullout Resistance – Comparison to ASD In AASHTO (2007) Equation 11.10.6.3.2-1, the value for Tmax is based on calculated horizontal loads with a load factor of 1.35 for the vertical earth pressure (EV). It should be noted that increases in horizontal loads resulting from surcharges, i.e. ∆σH, are also factored by the load factor of 1.35 (see Equation 7-9). Therefore, with a pullout resistance factor of 0.9 (see AASHTO (2007) Table 11.5.6-1) for static loading, the equivalent ASD factor of safety would be 1.35/0.9 = 1.5. The ASD factor of safety used in AASHTO (2002) is 1.5. All other parameters in AASHTO (2007) Equation 11.10.6.3.2-1 are formulated exactly as for ASD. Therefore, for this specific case (e.g., no live load surcharges), pullout design evaluations in LRFD are equivalent to ASD. FHWA NHI-07-071 Earth Retaining Structures

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Pullout Resistance Check for Traffic Barrier Loading: For checking pullout resistance, the full length of the reinforcements is considered effective in resisting pullout due to the impact loading. The upper layers of soil reinforcement should have sufficient pullout resistance to resist a horizontal load of γCTPH = 10 kips distributed over a 20 ft base slab length. The rationale for assuming a greater length for pullout (as compared to 2 ft for tensile resistance; discussed in subsequent sections) is because the entire base slab must move laterally to initiate a pullout failure due to the relatively large deformations required. 7.8.3.5 Step 13 – Calculate Long-Term Reinforcement Design Strength In this design step, the maximum factored tensile stress in each reinforcement layer (Tmax, which is calculated in design step 11) is compared to the nominal long-term reinforcement design strength (see AASHTO (2007) Equation 11.10.6.4.1-1) as presented in Equation 7-24 below.

Tmax ≤ φRcTal

(7-24)

where:

φ = resistance factor for tensile rupture (see Table 7-8); Table 7-8. Resistance Factors for Tensile Resistance (after AASHTO, 2007). Metallic Strip Reinforcement Reinforcement and • Static Loading Connectors • Combined Static/Earthquake Loading Grid Reinforcement1 • Static Loading • Combined Static/Earthquake Loading Geosynthetic • Static Loading Reinforcement and • Combined Static/Earthquake Loading Connectors

0.75 1.00 0.65 0.85 0.90 1.20

Note: 1Applies to grid reinforcements connected to a rigid facing element, e.g., a concrete panel or block.

Rc = reinforcement coverage ratio as defined in Equation 7-11b; and Tal = nominal long-term reinforcement design strength. FHWA NHI-07-071 Earth Retaining Structures

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The nominal long-term reinforcement design strength (Tal) for LRFD is computed for inextensible and extensible reinforcements as presented below.

Tal for Inextensible Reinforcements – AASHTO (2007) Equation 11.10.6.4.3a-1: Tal =

Ac Fy

(7-25)

b

where: Fy = minimum yield strength of steel; b = unit width of sheet, grid, bar, or mat; and Ac = design cross sectional area corrected for corrosion loss (Ac for strips and bars are defined below) The lower resistance factor of 0.65 for grid reinforcement (as compared to a resistance factor of 0.75 for strip reinforcement) accounts for the greater potential for local overstress due to load nonuniformities for steel grids than for steel strips or bars. Ac for strips is determined as: Ac = b tc = b (tn – ts)

(7-26)

where: b = unit width of sheet, grid, bar, or mat; tc = thickness at end of design life (See Figure 7-28); tn = thickness at end of construction; and ts = sacrificial thickness of metal expected to be lost by uniform corrosion during the service life of the structure (see discussion on corrosion rates below).


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tc b Figure 7-28. Cross Section Area for Strips. When estimating ts, it may be assumed that equal loss occurs from the top and bottom of the strip. Ac for bars is determined as: Ac = Nb (πD*2/4)

(7-27)

where: Nb = No. of bars per unit width b; and D* = Bar diameter after corrosion loss (Figure 7-29). When estimating D*, it may be assumed that corrosion losses occur uniformly over the area of the bar.

D*

b

Figure 7-29. Cross Section Area for Bars


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Corrosion Rates The corrosion rates presented below are suitable for conservative design. These rates assume a mildly corrosive backfill material having the controlled electrochemical property limits that are discussed under electrochemical properties for reinforced fills. Corrosion Rates - mildly corrosive backfill For corrosion of galvanization on each side

•

0.58 mil./yr/side (first 2 years) and

•

0.16 mil./yr/side (thereafter)

For corrosion of residual carbon steel on each side

•

0.47 mil./yr/side (after zinc depletion)

Based on these rates, complete corrosion of galvanization with the minimum required thickness of 3.4 mil. (AASHTO M 111) is estimated to occur during the first 16 years and a carbon steel thickness or diameter loss of 0.055 in. to 0.08 in. would be anticipated over the remaining years of a 75 to 100 year design life, respectively. The designer of an MSE structure should also consider the potential for changes in the reinforced backfill environment during the structure's service life. In certain parts of the United States, it can be expected that deicing salts might cause such an environment change. For this problem, the depth of chloride infiltration and concentration are of concern. For permanent structures directly supporting roadways exposed to deicing salts, limited data indicate that the upper 8 ft of the reinforced backfill (as measured from the roadway surface) are affected by higher corrosion rates not presently defined. Under these conditions, it is recommended that a 30 mil (minimum) geomembrane be placed below the road base and tied into a drainage system to mitigate the penetration of the deicing salts in lieu of higher corrosion rates.

Tal for Extensible Reinforcements – AASHTO (2007) Equation 11.10.6.4.3b-1: Tal =

Tult RF

(7-28)

where:


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Tult = minimum average roll value ultimate tensile strength. Tult may be obtained from ultimate (or yield) tensile strength wide strip test (ASTM D4595) for geotextiles and wide strip (ASTM D4595) or single rib test (GRI:GG1) for geogrids based on minimum average roll value (MARV) for the product; and RF = combined strength reduction factor to account for potential long-term degradation due to installation damage, creep and chemical aging. Determined as RF = RFID x RFCR x RFD where: RFID = strength reduction factor to account for installation damage to reinforcement; RFCR = strength reduction factor to prevent long-term creep rupture of reinforcement; and RFD = strength reduction factor to prevent rupture of reinforcement due to chemical and biological degradation. According to AASHTO (2007) Section 11.10.6.4.2b, values for RFID, RFCR, and RFD shall be determined from product specific test results.

Reduction Factor RFID RFID can range from 1.05 to 3.0, depending on backfill gradation and product mass per unit weight. Even with product specific test results, the minimum reduction factor shall be 1.1 to account for testing uncertainties. The placement and compaction of the backfill material against the geosynthetic reinforcement may reduce its tensile strength. The level of damage for each geosynthetic reinforcement is variable and is a function of the weight and type of the construction equipment and the type of geosynthetic material. The installation damage is also influenced by the lift thickness and type of soil present on either side of the reinforcement. Where granular and angular soils are used for backfill, the damage is more severe than where softer, finer, soils are used. For a more detailed explanation on the RFID factor, see FHWA-NHI-00-044 “Corrosion/Degradation of Soil Reinforcements for Mechanically Stabilized Earth Walls and Reinforced Soil Slopes,” (Elias, 2000). To account for installation damage losses of strength where full-scale product-specific testing is not available, Table 7-9 may be used with consideration of the project specified backfill characteristics. In absence of project specific data the largest indicated reduction factor for each geosynthetic type should be used.


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Table 7-9. Installation Damage Reduction Factors (after Elias, 2000). Reduction Factor (RFID) No.

Geosynthetic

1 2 3 4 5 6

HDPE uniaxial geogrid PP biaxial geogrid PVC coated PET geogrid Acrylic coated PET geogrid Woven geotextiles (PP&PET)(1) Non woven geotextiles (PP&PET)(1) Slit film woven PP geotextile (1)

7

(1) Minimum weight 7.9 oz/yd

Type 1 Backfill Max. Size 4 in. D50 about 1.2 in.

Type 2 Backfill Max. Size 0.75 in. D50 about 0.03 in.

1.20 - 1.45 1.20 - 1.45 1.30 - 1.85 1.30 - 2.05 1.40 - 2.20 1.40 - 2.50

1.10 - 1.20 1.10 - 1.20 1.10 - 1.30 1.20 - 1.40 1.10 - 1.40 1.10 - 1.40

1.60 - 3.00

1.10 - 2.00

2

Reduction Factor RFCR RFCR is obtained from long-term laboratory creep testing as detailed in Elias et al. (2001). This reduction factor is required to limit the load in the reinforcement to a level known as the creep limit, that will preclude creep rupture over the life of the structure. Creep in itself does not degrade the strength of the polymer. Creep testing is essentially a constant load test on multiple product samples, loaded to various percentages of the ultimate product load, for periods of up to 10,000 hours. The creep reduction factor is the ratio of the ultimate load to the extrapolated maximum sustainable load (i.e., creep limit) within the design life of the structure (e.g., several years for temporary structures, 75 to 100 years for permanent structures). Typical reduction factors as a function of polymer type are indicated in Table 710.

Table 7-10. Creep Reduction Factors (RFCR)


Polymer Type

RFCR

Polyster Polypropylene Polyethylene

1.6 to 2.5 4.0 to 5.0 2.6 to 5.0

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Reduction Factor RFD RFD is dependent on the susceptibility of the geosynthetic to be attacked by microorganisms, chemicals, thermal oxidation, hydrolysis and stress cracking and can vary typically from 1.1 to 2.0. Even with product specific test results, the minimum reduction factor shall be 1.1. A protocol for testing to obtain this reduction factor have been described in Elias et al. (1997). Current research suggest the following information regarding RFD for polyster and polyolefin geosynthetics. •

Polyster (PET) geosynthetics are recommended for use in environments characterized by 3 < pH < 9, only. The following reduction factors for PET aging (RFD) are presently indicated for a 100-year design life in the absence of product specific testing as presented in Table 7-12.

•

Polyolefin geosynthetics have a unique and proprietary blend of antioxidants, product specific testing is required to determine the effective life span of protection at the inground oxygen content. Limited data suggests that certain antioxidants are effective for up to 100 years in maintaining strength for in-ground use. Table 7-11. Aging Reduction Factors (RFD) for PET.

No.

Product*

Reduction factor (RFD) 5≤pH≤8 3≤pH≤5 8≤pH≤9

Geotextiles Mn<20,000, 40 25,000, CEG<30 Mn = number average molecular weight 1

1.6

2.0

1.15

1.3 CEG = carboxyl end group

* Use of materials outside the indicated pH or molecular property range requires specific product testing.

Tensile Resistance – Comparison to ASD Use of resistance factors for tensile resistance provided in Table 7-8 for steel strips, steel grid, and geosynthetics will result in equivalent designs to those based on ASD for considerations of tensile rupture. For a load factor, EV=1.35, with a resistance factor of 0.75 for steel strips, 0.65 for steel grid, and 0.9 for geosynthetics, the equivalent ASD factor of FHWA NHI-07-071 Earth Retaining Structures

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safety would be 1.8 (=1.35/0.75) for steel strips, 2.08 (=1.35/0.65) for steel grids, and 1.5 (=1.35/0.9) for geosynthetics, respectively. Taking the inverse of the equivalent ASD factor of safety values for steel strips and steel grids results in 0.55 and 0.48, the allowable stress used for ASD design. The factor of safety of 1.5 for geosynthetics is the same as that used for ASD.

Tensile Resistance Check for Traffic Barrier Loading: As for CIP walls, effects of traffic barrier loading are considered at the Extreme Event II limit state for MSE walls. The upper layers of reinforcement (typically upper two rows) are checked to verify that they have sufficient tensile resistance to resist a concentrated factored horizontal load, γCTPH, of 10 kips distributed over a barrier length of 5 ft. The resulting unit load of 2 kips/ft of wall would be distributed as shown in Figure 3.11.6.3-2a of AASHTO (2007). If the traffic barrier base slab is cast integrally with a concrete pavement, the additional load may be neglected. For geosynthetic reinforcements, for this transient load case, the nominal long-term reinforcement design strength, Tal, does not to be reduced to account for creep (i.e., RFCR assumed equal to 1.0). Because traffic barrier loading is transient, when designing for rupture of a geosynthetic reinforcement, the reinforcement is designed for both static and transient components of the load (see Section 11.10.10.2 of AASHTO, 2007). 7.8.4

Service Limit States (Steps 14 and 15)

7.8.4.1 Step 14 – Check Overall Stability This design step is performed to check the overall stability of the wall. Overall stability is determined using rotational or wedge analyses, as appropriate, which can be performed using a classical slope stability analysis method. Computer programs that directly incorporate reinforcement elements (e.g., ReSSA) are available for these analyses. Generally, the reinforced soil wall is considered as a rigid body and only failure surfaces completely outside a reinforced mass (e.g., global failure planes) are considered. For simple structures with rectangular geometry, relatively uniform reinforcement spacing, and a near vertical face, compound failure planes (e.g., passing both through the unreinforced and reinforced zones) will not generally be critical. However, if complex conditions exist such as changes in reinforced soil types or reinforcement lengths, high surcharge loads, sloping faced structures, significant slopes at the toe or above the wall, or stacked structures, compound failures must be considered. The evaluation of overall stability of earth slopes with or without a foundation unit should be investigated at the Service 1 Load Combination and an appropriate resistance factor. FHWA NHI-07-071 Earth Retaining Structures

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Commonly used slope stability programs can be used to conduct this evaluation. In lieu of better information, the resistance factor (φ) is defined in AASHTO (2007) Section 11.6.2.3 as:

•

φ = 0.75; where the geotechnical parameters are well defined, and the slope does not support or contain a structural element; and

•

φ = 0.65; where the geotechnical parameters are based on limited information, or the slope contains or supports a structural element

The evaluation of overall stability should be performed with reasonable estimates of longterm water pressures acting on the wall. If the evaluation of overall stability does not indicate a satisfactory result then the reinforcement length may have to be increased or the foundation soil may have to be improved. The design must be revised according to these changes. It should be noted that wall designs that are performed by MSE wall suppliers typically will not include the overall stability check. 7.8.4.2 Step 15 – Estimate Wall Movements

Vertical Wall Movements MSE structures have significant tolerances, both longitudinally along the wall and perpendicular to its front face. Therefore, poor foundation conditions seldom preclude their use. However, conventional settlement analyses should be carried out to ensure that immediate, primary consolidation, and secondary consolidation settlements of the wall are less than the performance requirements of the project. Vertical wall movements can be evauluated at the Service I Limit State as described in AASHTO (2007) Section 10.6.2.2. Additionally, FHWA NHI-06-088 (Samtani and Nowatzki, 2006) describes various procedures to compute settlements. During the design, if significant differential settlement is anticipated (more than 1/100 for precast facings), sufficient joint width and/or slip joints must be provided to preclude facing distress. Limitations of system differential tolerance may influence the type and design of the facing panel selected. Square panels generally adapt to larger longitudinal differential settlements better than long rectangular panels of the same surface area. Guidance on minimum joint width and limiting differential settlements that can be tolerated is presented in Table 7-12. FHWA NHI-07-071 Earth Retaining Structures

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Table 7-12. Relationship between Joint Width and Limiting Differential Settlements for MSE Precast Panels (after Elias et al., 2001). Joint Width (in.)

Limiting Differential Settlement

0.75 0.50 0.25

1/100 1/200 1/300

Note: This table is applicable for panel size less than 30 square feet.

MSE walls constructed with full-height precast concrete panels should be limited to differential settlements of 1/500. Walls with drycast facing (MBW) should be limited to settlements of 1/200. For walls with welded wire facings, the limiting differential settlement should be 1/50. Where significant differential settlement perpendicular to the wall face is anticipated, the reinforcement connection may be overstressed. Where the back of the reinforced soil zone will settle more than the face, the reinforcement could be placed on a sloping fill surface, which is higher at the back end of the reinforcement to compensate for the greater vertical settlement. This may be the case where a steep surcharge slope is constructed. This latter construction technique, however, requires that surface drainage be carefully controlled after each day's construction. Alternatively, where significant differential settlements are anticipated, ground improvement techniques (see NHI course 132034 and its reference manual FHWA NHI-04-001 (Elias et al., 2004) may be warranted to limit the settlements. Although uniform settlement is of little consequence to the stability of the MSE structure, the total settlement may influence the serviceability of the wall, such as in the case of a bridge supported by a MSE abutment. The total settlement is particularly important when a MSE wall interacts with an adjacent structure that has different load-deformation characteristics, e.g., when an abutment is supported on piles and the approaches utilize MSE walls. In such cases, downdrag forces on the abutment piles due to settlement of the adjacent MSE structure need to be evaluated.

Lateral Wall Movements The evaluation of lateral wall movements in LRFD is the same as in ASD as the deformations are evaluated at the Service I limit state.


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In general, lateral deformations in the reinforced backfill of an MSE wall usually occur during construction. Post construction movements, however, may take place due to post construction surcharge loads or long-term settlement of the foundation soils. The lateral displacements depend on compaction effects, reinforcement extensibility, reinforcement length, reinforcement-to-facing connection details, and details of the wall facing. In general, increasing the length-to-height ratio of reinforcement, from its theoretical lower limit of 0.5H to 0.7H, decreases the deformation by 50 percent. Also, the anticipated deformation of MSE structures constructed with polymeric reinforcements (extensible) is approximately three times greater than if constructed with metallic reinforcements (inextensible). The rough estimate of probable lateral displacements of simple MSE walls that may occur during construction can be estimated based on empirical correlations (Elias et al., 2001). For critical structures requiring precise tolerances, such as bridge abutments, more accurate calculations using the finite element method may be warranted. 7.8.5

Design of Wall Drainage Systems (Step 16)

The discussions on subsurface and surface drainage of CIP walls (Chapter 5) are applicable to MSE walls. Additional considerations for MSE walls include location of drain(s) and potential infiltration of deicing salt runoff. A drain at the base of the wall immediately behind the wall facing, as shown in Figure 7-30, is normally used with an MSE wall. This drain primarily serves to collect surface water that has infiltrated immediately behind the facing and transported to an outlet. Outlet may be via weep holes, as shown in Figure 7-30, or may be piped to a downslope outlet or to a storm sewer. This drain may also serve to drain the wall fill if it is relatively free-draining and the wall is used in a fill situation. A drain behind the wall backfill, as illustrated in Figure 7-31, should be used when the zones of soils (in-situ or retained backfill to wall) are not free-draining relative to one another. A drain is recommended to collect and divert groundwater from the reinforced fill for side hill construction, where in-situ soils are excavated to accommodate the reinforced fill volume. Additionally, a drain at the backcut or at the wall fill and retained backfill interface (for fill situations) is recommended, unless soil permeabilities, filtration, and water flow are analyzed to ensure system is free draining. Soil filtration and permeability requirements must be met between the two adjacent zones of (different) soils to prevent impeded flow. Filtration criteria are summarized in Section 2.5.8.2.A geomembrane


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RETAINED BACKFILL REINFORCED BACKFILL (OR REINFORCED WALL FILL)

WEEP HOLE COLLECTION & DRAIN PIPE

Figure 7-30. Drain Immediately Behind Concrete Facing in MSE Wall.

RETAINED BACKFILL REINFORCED BACKFILL (OR REINFORCED WALL FILL)

CHIMNEY DRAIN BENCHED BACKCUT

WEEP HOLE COLLECTION & DRAIN PIPE

COLLECTION PIPE, OUTLET TO SLOPE, WEEPHOLE OR STORM SEWER

FOUNDATION SOIL

Figure 7-31. Drain Behind the Backfill in a Wall, in a Backcut, and Fill Situation. FHWA NHI-07-071 Earth Retaining Structures

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barrier beneath the pavement structure, sloping to a drain, should be used where significant de-icing salts are used. The purpose of the barrier is to prevent, or minimize, the infiltration of water immediately behind the wall facing. A common detail is shown in Figure 7-32. 30 mil (min) Impervious Membrane (HDPE) Wearing Surface Base Course

6” Perforated Pipe Cover in Geotextile (Drain to end of wall) Drainage Layer #57 Aggregagate

> 5%

Front Face of MSE wall

5 ft

1 ft

Specified Backfill

Figure 7-32. Impervious Geomembrane Details.

7.8.6 Additional Design Considerations

7.8.6.1 Connection Design Different types of connections are used in MSE walls depending basically on the type of reinforcement used and the wall facing details. The connections shall be designed to resist stresses resulting from the factored tensile load applied to the soil reinforcement connection at the wall face as well as from differential movements between the reinforced backfill and the wall facing elements (see Section 11.10.6.4.4 of AASHTO (2007) for a more detailed


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discussion on the design of connections involving steel and geosynthetic reinforcements). The resistance of the connection is usually tested according to standards outlined by AASHTO (2007). Connection details and their strength requirements are discussed by Elias et al. (2001) and AASHTO (2007). 7.8.6.2 Wall Face Design

Design of Concrete, Steel and Timber Facings Facing elements are designed to resist the horizontal forces developed internally within the wall. Reinforcement is provided to resist the average loading conditions at each depth in accordance with structural design requirements in Section 5, 6 and 8 of AASHTO (2007) for concrete, steel and timber facings, respectively. The embedment of the reinforcement to panel connector must be developed by test, to ensure that it can resist Tmax. As a minimum, temperature and shrinkage steel must be provided for segmental precast facing. Epoxy protection of panel reinforcement where salt spray is anticipated is recommended.

Design of Flexible Wall Facings Welded wire or similar facing panels shall be designed in a manner which prevents the occurrence of excessive bulging as backfill behind the facing elements compresses due to compaction stresses, self weight of the backfill or lack of section modulus. Bulging at the face between soil reinforcement elements in both the horizontal and vertical direction should be limited to 1 to 2 inches as measured from the theoretical wall line. This may be accomplished by requiring the placement of a nominal 2 ft wide zone of rockfill or cobbles directly behind the facing, decreasing the spacing between reinforcements – vertically and horizontally, increasing the section modulus of the facing material and by providing sufficient overlap between adjacent facing panels. In addition, the reinforcements must not be restrained and have the ability to slide vertically with respect to the facing material. Furthermore, the top of the flexible facing panel at the top of the wall shall be attached to a soil reinforcement layer to provide stability to the top facing panel. For modular concrete facing blocks sufficient inter-unit shear capacity must be available, and the maximum spacing between reinforcement layers should be limited to 24 in.. The maximum facing height above the uppermost reinforcement layer and the maximum depth of facing below the bottom reinforcement layer should be limited to the width, Wu, of the modular concrete facing unit used.


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For seismic performance categories "C" or higher (AASHTO Division 1A), facing connections in MBW shall not be fully dependent on frictional resistance between the backfill reinforcement and facing blocks. Shear resisting devices between the facing blocks and soil reinforcement such as shear keys, pins, etc. shall be used. For connections partially or fully dependent on friction between the facing blocks and the soil reinforcement, the long-term connection strength (Tac), should be reduced to 80 percent of its static value. Furthermore, for the blocks above the uppermost layer, soil reinforcement layer must be secured against toppling under all seismic events. Geosynthetic facing elements should not be left exposed to sunlight (specifically ultraviolet radiation) for permanent walls. If geosynthetic facing elements must be left exposed permanently to sunlight, the geosynthetic shall be stabilized to be resistant to ultraviolet radiation. Furthermore, product specific test data should be provided which can be extrapolated to the intended design life and which proves that the product will be capable of performing as intended in an exposed environment. Alternately a protective facing shall be constructed in addition (e.g., concrete, shotcrete, etc.). 7.8.6.3 Computer Assisted Design The repetitive nature of the computations required at each level of reinforcement lends itself to computer-assisted design. The computer program MSEW (ADAMA, 2000) developed under FHWA sponsorship analyzes and/or designs MSE walls using any type of metallic or geosynthetic reinforcement in conjunction with any type of facing (precast concrete, MBW, etc.). Version 1.0 has been designated exclusively for use by U.S. State Highway Agencies and by U.S. Federal agencies. Version 3.0 is available for purchase through ADAMA Engineering (www.MSEW.com) and includes LRFD-based computations.

7.9

MSE WALL EXAMPLE PROBLEM

7.9.1

Problem Statement

Design a 20 ft high retaining wall shown in Figure 7-33a for a 75 year design life. Total height of the wall, including embedment, will be 22 ft. The top of the wall will be loaded with traffic (i.e., live load surcharge). The wall will use Grade 60 inextensible ribbed steel strip reinforcements (as shown in Figure 7-28) that are 0.15 in. thick and 2 in. wide. The reinforcements have a 3.4 mil thick galvanized coating. FHWA NHI-07-071 Earth Retaining Structures

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The wall face will consist of precast concrete panels that are 5 ft x 5 ft. There will be 2 rows of reinforcements per panel and the horizontal (Sh) and vertical (Sv) spacing between reinforcements will be 2.5 ft as shown in Figure 7-34. The engineering properties of retained and reinforced backfills are as follows:

20 ft

φr’ and φf’ = 30o (clayey sand, medium dense) γr and γf = 120 pcf

22 ft

• •

PLSH PEH PEV

(a)

(b)

Figure 7-33. (a) Geometry of the Problem and (b) External Forces to be Considered in Analysis. 7.9.2

Solution of the Problem

Design steps for MSE walls are provided in Table 7-5. In this design example, it is assumed that the requirements for steps 1 and 2 are completed. Therefore, the solution presented below starts with design step 3. 7.9.2.1 Step 3: Initial Geometry

Design wall height (H) In this design, the wall height is selected as 22 ft including a 2 ft embedment depth. As presented in Table 7-6, the minimum required embedment depth for a 20 ft high wall is 1ft. For a complete design, the selected design wall height would also be evaluated based on overall stability, bearing resistance, and scour potential. FHWA NHI-07-071 Earth Retaining Structures

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5 ft Reinforcement Strip

2.5 ft

2.5 ft

5 ft

2.5 ft

2.5 ft

Figure 7-34. Wall Face Panels and Spacing between Reinforcements.

Minimum reinforcement length (L) As discussed in Section 7.8.1.2, the minimum reinforcement length should be greater of 0.7H or 6 ft. Therefore, as a preliminary estimate, L is selected to be: L = 0.7 x 22 ft = 15.4 ft (use 16.0 ft). 7.9.2.2 Step 4: Unfactored Loads

Vertical Loads Vertical Earth Pressure (EV) The weight of the reinforced soil backfill is: PEV = H L γr = (22 ft)(16.0 ft)(120 pcf) PEV = 42.24 kips/ft length of wall

Surcharge Load (LS) The vehicular surcharge load on the wall is calculated using an equivalent height of soil (heq). AASHTO (2007) provides a table summarizing heq based on the wall height (Table 7-13). FHWA NHI-07-071 Earth Retaining Structures

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Table 7-13. Equivalent Height of Soil for Vehicular Loading (after AASHTO, 2007). Wall Height (ft)

heq (ft)

5

5

10

3.5

≥20

2

Based on Table 7-13 above, heq for 22 ft high wall is 2 ft. Therefore, the vertical load due to vehicular loading can be estimated as: PLSV = (γr)(heq)(L) = 120 pcf x 2 ft x 16.0 ft = 3.84 kips/ft length of wall It should be noted that the traffic load surcharge over the reinforced zone is not considered for checks on sliding, eccentricity, or reinforcement pullout, but is considered in evaluation of bearing resistance, overall stability, and reinforcement tensile resistance.

Horizontal Loads The active earth pressure coefficient, Ka is:  φ'f K a = tan 2  45 − 2 

  

where: φ’f = 30o, therefore Ka = 0.33

Horizontal Earth Pressure (EH) PEH = 0.5 (γr)(H2)Ka = (0.5)(120 pcf)(22 ft)2(0.33) PEH = 9.58 kips/ft length of wall

Surcharge Load (LS) The horizontal load due to surcharge is computed based on the uniform increase in horizontal earth pressure due to traffic load surcharge (∆p) as: FHWA NHI-07-071 Earth Retaining Structures

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∆p = (ka)(γr)(heq) ∆p = (0.33)(120 pcf)(2 ft) = 0.0792 ksf For a 22 ft high wall, the resultant of the live load surcharge horizontal earth pressure (PLSH), acting on the reinforced soil mass becomes: PLSH = (∆p)(H) PLSH = 0.0792 ksf x 22 ft = 1.74 kips/ft length of wall Tables 7-14 and 7-15 summarize the unfactored vertical and horizontal loads, respectively. The moment arms about the toe of the wall for each of these loads are also summarized. Table 7-14. Summary of Unfactored Vertical Loads and Moment Arms for Design Example Load

V (kips/ft)

Moment Arm About Toe (ft)

PEV PLSV Total

42.24 3.84 46.08

8.0 8.0

Table 7-15. Summary of Unfactored Horizontal Loads and Moment Arms for Design Example Load

H (kip/ft)

Moment Arm About Toe (ft)

PEH = PaH PLSH Total

9.58 1.74 11.32

7.33 11.0

7.9.2.3 Step 5: Factored Loads and Moments The load combinations and load factors used in this example are summarized in Table 7-16. Table 7-16. Load Factors and Load Combinations GROUP Strength I Service I FHWA NHI-07-071 Earth Retaining Structures

γEV max min 1.35 1.00 1.00 1.00

γEH (Active)

γLS

1.50 1.00

1.75 1.00

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Factored vertical and horizontal loads can be determined based on multiplying unfactored loads (summarized in Tables 7-14 and 7-15) with load factors in Table 7-16. The factored loads are provided in Tables 7-17 and 7-18. Table 7-17. Factored Vertical Loads and Moments

GROUP Unfactored Strength I (γEV = 1.00) Strength I (γEV = 1.35) Service I

3.84

VTOT = PEV + PLSV (kips/ft) 46.08

Moment Arm About Toe (ft) 8.0

42.24

6.72

48.96

57.02

6.72

42.24

3.84

63.74 46.08

PEV (kips/ft)

PLSV (kips/ft)

42.24

MEV (kips-ft/ft)

MVTOT (kips-ft/ft)

337.92

368.64

8.0

337.92

391.68

8.0

456.19 337.92

509.92 368.64

8.0

Table 7-18. Factored Horizontal Loads and Moments

GROUP Unfactored Strength I (γEV = 1.00) Strength I (γEV = 1.35) Service I

Moment Moment Arm HTOT = PLSH PEH Arm About PEH + PLSH About Toe for (kips/ft) (kips/ft) Toe for PLSH (kips/ft) PEH (ft) (ft) 11.32 9.58 7.33 1.74 11.0

MHTOT (kips-ft/ft) 89.39

14.37

7.33

3.05

11.0

17.42

138.93

14.37

7.33

3.05

11.0

17.42

138.93

9.58

7.33

1.74

11.0

11.32

89.39

7.9.2.4 Step 6: Eccentricity The eccentricity check is summarized in Table 7-19. Table 7-19. Summary for Eccentricity Check

GROUP Strength I (γEV = 1.00) Strength I (γEV = 1.35) Service I

PEV (kips/ft)

MEV (kips-ft/ft)

MHTOT (kips-ft/ft)

Xo (ft)

eB (ft)

Eccentricity Check (eB < emax)

42.24

337.92

138.93

4.71

3.29

√

57.02

456.19 337.92

138.93

5.56 5.88

2.44 2.12

√

42.24


89.39 7-81

√ 7 – MSE Walls June 2008

where: Xo = Location of the resultant from toe of wall = (MEV - MHTOT)/PEV B = Base width = Length of reinforcement strips = 16.0 ft eB = Eccentricity = B/2 - Xo The location of the resultant must be in the middle half of the base. emax= B/4 = 16.0 ft/4 = 4.0 ft For all cases, eB < emax; therefore, the design is adequate with respect to eccentricity. 7.9.2.5 Step 7: Sliding Resistance The passive resistance of the foundation material is neglected when checking the sliding resistance of MSE walls as discussed in section 7.8.2.4. Therefore, the factored resistance against failure by sliding (RR) is computed with Equation 7-5 as:

RR = φτ Rτ where:

φt = 1.0 (see section 7.8.2.4) and Rt = V tan δ where V = PEV and tan δ = tan φf Therefore:

Rn = PEV tan φf PEV is obtained from Table 7-18 as the minimum vertical factored load for the strength limit state (i.e., using γEV = 1.00)

φf = 30o

(It is assumed in this example that the friction angle of reinforced fill is less than the friction angle of the foundation and the interface friction angle between reinforcement and soil. See section 7.8.2.4)

Rn = (42.24 kips/ft ) (tan30) = (42.24 kips/ft ) (0.58) Rn = 24.5 kips/ft length of wall FHWA NHI-07-071 Earth Retaining Structures

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Applying the resistance factor φ to Rn, the factored sliding resistance is:

RR = (1.0)(24.5 kips/ft) = 24.5 kips/ft length of wall

For the wall to have adequate sliding resistance:

RR must be > HTOT (maximum)

(HTOT is obtained from Table 7-19 as the maximum total horizontal factored load for the strength limit state (i.e., 17.42 kips/ft))

In this design example:

RR = 24.5 kips/ft >HTOT = 17.42 kips/ft, therefore, sliding resistance is adequate. 7.8.2.6 Step 8: Bearing Resistance The factored bearing resistance is computed with Equation 7-7 as:

qR = φ qn where:

φ = 0.65

(see Table 10.5.5.2.2-1 in AASHTO (2007)) and

qn = cNcm + γDfNqmCwq + 0.5γ BNγmCwγ c=0 ⇒

In this design example:

(see AASHTO (2007) Equation 10.6.3.1.2a-1)

cNcm = 0

Df = 0 ⇒ γDfNqmCwq = 0 (i.e., Df = 0 because embedment depth is neglected) qn = 0.5γ BNγmCwγ Nγm = Nγsγiγ

(see AASHTO (2007) Equation 10.6.3.1.2a-4)

Nγ = 22.4 (see AASHTO Table 10.6.3.1.2a-1) B  = 1.0  L

sγ = 1 − 0.4

(Assumed long wall)

iγ = 1.0 (see AASHTO Table 10.6.3.1.2a-8) Cwγ = 1.0 (Assumed that the groundwater depth is > 1.5B+Df)

qn = 0.5(120 pcf)(16 ft – 2 * (eB))(22.4)(1.0)(1.0)(1.0) = 1.344 (16 - 2eB (ft)) ksf FHWA NHI-07-071 Earth Retaining Structures

7-83


qR = φ qn = 0.874 (16 - 2eB (ft)) (ksf) To check whether the bearing resistance of the MSE wall is adequate, qR computed above must be compared against the following criteria: q R ≥ quniform

where:

quniform =

VTOT B − 2e B and

eB is computed with Equation 7-3 with Xo computed as (MVTOT – MHTOT)/VTOT. In this check the effects of live load surcharges are included because they increase the loading on the foundation. Table 7-20. Summary for Checking Bearing Resistance. GROUP Strength I (γEV = 1.00) Strength I (γEV = 1.35) Service I

VTOT (kips/ft)

MVTOT (kips-ft/ft)

MHTOT (kips-ft/ft)

Xo (ft)

eB (ft)

quniform (ksf)

qR (ksf)

48.96

391.68

138.93

5.16

2.84

4.74

9.0

63.74

509.92

138.93

5.82

2.18

5.48

10.2

46.08

368.64

89.39

6.06

1.84

3.80

10.8

Maximum value of quniform = (5.48 ksf). qR > quniform (10.2 ksf > 5.48 ksf), therefore bearing resistance is adequate. 7.9.2.7 Step 9: Critical Failure Surface The critical failure surface for the inextensible reinforcement is determined by diving the reinforced zone into active and resistant zones with a bilinear failure surface as shown in Figure 7-23a. For the top half of the wall, La = 0.3H. 7.9.2.8 Step 10: Factored Horizontal Stress The factored horizontal stress (σH) at each reinforcement level is determined with Equation 7-9 as: FHWA NHI-07-071 Earth Retaining Structures

7-84


σ H = γ P (σ V k r + ∆σ H ) where: γP = the maximum load factor for vertical earth pressure (= 1.35); kr = horizontal pressure coefficient; σV = unfactored vertical stress at the reinforcement level due to resultant of gravity forces (i.e., γr Hi) and any surcharge loads (i.e., γr heq);

γr = unit weight of reinforced fill; Hi = backfill thickness over the reinforcement within layer i; and heq = equivalent height of soil for vehicular loading. ∆σH = horizontal stress at reinforcement level resulting from a concentrated surcharge load. For this example, no concentrated surcharge loads are present. For this example, the horizontal stress will only be calculated for the top most layer (i.e., Layer 1). The reinforcement layout is based on a vertical spacing of 2.5 ft starting 1.25 ft from the top. Therefore the Layer 1 reinforcement will be overlain by 1.25 ft of backfill. In step 4, ka is computed as 0.33. For steel reinforcements at 1.25 ft depth, K/Ka from Figure 7-24 is approximately 1.67. Therefore, kr = 1.67 x 0.33 = 0.551. For Top Layer:

σv = γr Hi + γr heq = 120 pcf x 1.25 ft + 120 pcf x 2 ft = 390 psf σH = 1.35 (390 psf x 0.551) = 290.1 psf 7.9.2.9 Step 11: Maximum Tensile Force The maximum factored tensile force in the Top Layer (using Equation 7-11a) is computed as: Tmax = σ HSv Therefore, Tmax = 290.1 x 2.5 ft = 725.25 lb/ft


7-85


7.9.2.10 Step 12: Reinforcement Pullout Resistance The effective pullout length (Le) should be computed for each reinforcement layer. In this example, Le is computed for Layer 1 using Equation 7-14 as: Le ≥

Tmax φ F α σ vC R c *

where: Tmax = maximum tensile force determined in step 11;

φ = 0.9 based on AASHTO (2007) Table 11.5.6-1; F* = pullout friction factor determined from Figure 7-27 = 1.72; α = scale effect correction factor determined from Table 7-8 = 1; σv = unfactored vertical stress at the reinforcement level. For Top Layer, it is computed as 120 pcf (1.25 ft) = 150 psf; C = 2 for strip type reinforcement; and Rc = coverage ratio (b/Sh). b = 2 in. wide strip (given in problem statement) and Sh = 2.5 ft wide panel (Figure 7-34). Therefore, b/Sh = 0.066. For Layer 1, the required length of reinforcement in the resistance zone is calculated as follows: Le =

725.25 lb/ft = 23.6 ft 0.9 x 1.72 x 1 x 150 psf x 2 x 0.066

The total length of reinforcement (L) required for the Top Layer for internal stability is, therefore: L = La + Le For Top Layer, La is determined from step 9 as 0.3 H = 0.3 x 22 ft = 6.6 ft L = 6.6 + 23.6 = 30.2 ft Length of reinforcement estimated in step 3 was 16.0 ft. Therefore, the total length of reinforcement is unacceptable. An additional soil overburden over the top of the uppermost reinforcement would improve the pullout resistance of the uppermost reinforcement (and other reinforcement layers).


7-86


7.9.2.11 Step 13: Reinforcement Tensile Resistance The tensile resistance of reinforcement should be checked for each reinforcement layer. The tensile resistance of reinforcement for Layer 1 is checked by comparing the maximum factored tensile force (Tmax) computed in step 11 against the factored nominal long-term reinforcement design strength as presented in Equation 7-24 as: Tmax ≤ φ Rc Tal where:

φ = resistance factor for tensile rupture = 0.75 (see Table 7-9); Rc = 0.066 as computed in step 12; Tal = nominal long-term reinforcement design strength. For inextensible reinforcements, Tal is computed based on Equation 7-25 as: Tal =

A c Fy b

where: Fy = minimum yield strength of steel (60 ksi for grade 60 steel); b = unit width of steel sheet (i.e., 2 in.) Ac = design cross sectional area corrected for corrosion loss Ac for strips is determined based on Equation 7-26 as: Ac = b tc = b (tn – ts) where: b = unit width of steel strip sheet (i.e., 2 in); tc = thickness at end of design life; tn = thickness at end of construction (i.e., 0.15 in); and ts = sacrificial thickness of metal for service life of the structure FHWA NHI-07-071 Earth Retaining Structures

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The design life of the structure is 75 years. For this example, the soil is assumed to be nonaggressive and the steel strip is coated with a 3.4 mil thick galvanization. Corrosion rates for design are summarized below: Loss of galvanization = 0.58 mil/side/yr for first 2 years; and = 0.16 mil/side/yr for subsequent years. Loss of carbon steel = 0.47 mil/side/yr The time required for galvanization loss is calculated as: t a = 2 yrs +

(3.4 mil − 2 yrs(0.58 mil / yr ) ) 0.16 mil / yr

t a = 16 yrs Therefore, carbon steel corrosion is assumed to occur for 59 years. Corrosion losses of carbon steel are computed as: ts = (59 yrs) (0.47 mil/yr) = 0.0277 in./side Therefore: tc = tn − t s = 0.15 in. − 2(0.0277 in.) = 0.095 in. Ac = bt c = (2 in.)(0.095 in.) = 0.19 in.2

Tal = (0.19 in2 x 60 ksi) / 2 in = 5.7 kips/in = 68.4 kips/ft = 68,400 lb/ft The tensile resistance of reinforcement for Layer 1 is checked by: Tmax ≤ φ Rc Tal 725.25 lb/ft ≤ 0.75 x 0.066 x 68,400 lb/ft 725.25 lb/ft ≤ 3,385 lb/ft Therefore, the tensile resistance of the Layer 1 reinforcement is adequate.


7-88


CHAPTER 8 NONGRAVITY CANTILEVERED AND ANCHORED WALLS 8.1

INTRODUCTION

Externally supported structural walls rely primarily on the bending resistance of a vertical structural element to resist the applied lateral loads. The vertical wall elements may consist of discrete elements (e.g., soldier piles) spanned by a structural facing, or may be a continuous structure (e.g., sheet pile wall, tangent pile wall, slurry wall, jet-grouted wall, and deep mixing method (DMM) wall). The primary types of externally supported walls are illustrated in Figure 8-1. The wall configurations include nongravity cantilevered walls and structures with single or multiple levels of support. Wall support may be provided by a system of struts or rakers on the exposed side of the wall, or anchors installed through the wall (Figure 8-2). Externally supported structural walls can be used for both temporary and permanent wall applications. Typical applications for such walls are listed below: Walls for temporary excavation support include: • • • • •

cofferdams on land and in water for footing construction; excavation support for CIP wall construction; grade separation during staged roadway construction; excavation support for cut-and-cover roadway tunnels and culverts; and excavation support for utility trenches.

Permanent wall applications include: • • • • • •

roadway cuts; roadway fill containment; abutments; cut-and-cover tunnel walls; slide stabilization; and waterfront structures.


8-1

8 – Non-Gravity Cantilevered and Anchored Walls June 2008

Wood or precast Concrete lagging

(a)

(d)

5 to 10 ft

(e)

2 to 3 ft 1.5 to 3 ft

CIP concrete lagging or facing

(f)

(b)

1.5 to 6.5 ft

(g)

6 to 10 ft

(h) (c)

Figure 8-1. Primary Types of Externally Supported Structural Walls: (a) Soldier Pile and Lagging Wall; (b) Soldier Pile and Cast-In-Place Concrete Lagging Wall; (c) Master Pile Wall; (d) Sheet Pile Wall; (e) Slurry (Diaphragm) Wall; (f) Secant Pile Wall; (g) Tangent Pile Wall; and (h) Interlocking H-Pile Wall (after Dismuke, 1991) The primary advantages of externally supported structural walls (as compared to other types of walls) include: • •

• • •

top-down method of construction is used (i.e., wall elements are installed before the start of excavation) which generally results in reduced ground displacement; when used as permanent walls they require; o reduced quantity of excavation; o reduced quantity of backfill; o reduced work area; faster construction time; can be used to provide a seepage barrier (e.g., for a depressed roadway); and can effectively support large vertical loads as well as lateral loads.


8-2


Figure 8-2. Wall Support Systems: (a) Cantilever Wall; (b) Earth Berm Support; (c) Raker System; (d) Deadman Anchor; (e) Cross-lot Braced Wall; and (f) Anchored Wall (after NAVFAC, 1986). FHWA NHI-07-071 Earth Retaining Structures

8-3


Some of the limitations of externally supported structural walls (as compared to other types of walls) include: • • • •

typically require more specialized construction techniques; complicated soil-structure interaction may make analysis and design more difficult and costly; performance of the completed wall is more dependant on the construction method and quality of the work; and metallic components, especially anchors, if in contact with soil or rock, are more susceptible to corrosion.

8.2

SHEET PILE WALLS

8.2.1

General

A sheet pile wall consists of a series of interlocking sheet piles driven side by side into the ground, thus forming a continuous vertical wall. Sheet pile walls are commonly used for waterfront structures and for temporary earth support applications, but can also be used as permanent walls for highway structures. Sheet piling is also used for stabilizing ground slopes and for cellular cofferdam construction. Figure 8-3 show several applications for sheet pile walls. In certain applications sheet pile walls are termed as “bulkheads” or “cofferdams.” A bulkhead is generally a sheet pile retaining wall used for waterfront construction. A cofferdam is a reasonably water tight enclosure made of sheet piling, usually temporary, built around a working area for the purpose of supporting lateral soil and water loads and excluding water during construction. Sheet piles can be timber, reinforced concrete, steel, vinyl, or composite material. Timber sheet piling is used for short spans, light lateral loads, and for temporary support of shallow excavations. Concrete sheet piles are precast members, possibly prestressed, usually with a tongue and groove joint, designed to withstand the permanent service loads and handling stresses during construction. They are heavy and bulky, and require heavier equipment to drive and handle. Concrete sheet piles are generally used only for permanent wall applications. Steel sheet piles are the most commonly used type.


8-4


(c)

(a)

(b) (d)

(e)

Figure 8-3. (a) Sheet Pile Wall for Earth Support Behind a Cast-in-Place Wall; (b) Cofferdam for Construction of Foundations in Water; (c) Cofferdam for Footing Construction on Land; (d) Anchored Bulkhead; and (e) Bridge Abutment.


8-5


The advantages of steel sheeting include: • • • •

lightweight; easier to drive and extract; higher bending resistance; and sheet piles reusable several times for temporary walls.

Some of the limitations of steel sheeting are: • • •

cannot penetrate hard layers; limited sheeting length (about 95 ft); and wall elements are susceptible to corrosion.

Steel sheeting is fabricated either using hot rolling or a cold formed manufacturing process. Interlocking Z-shaped or U-shaped sections are typically used for retaining wall applications since these sections provide a higher bending resistance and corresponding greater moment of inertia. Cold formed sections are generally lighter (i.e., lower section modulus) than hotrolled sections and the connection formed by cold-rolled interlocks is looser than for hotrolled interlocks. Flat sheets are generally limited to use in cellular cofferdams. Some common steel sheet pile sections used for sheet pile walls are shown in Figure 8-4.

Figure 8-4. Steel Sheet Pile Sections Commonly Used for Retaining Walls and Cofferdams. (a) Z -Section, (b) U-Section, (c) Cold Formed Section.


8-6


Vinyl or composite sheet pile may be used for waterfront structures or to replace failing (or degrading) steel sheet pile walls by installation of vinyl or composite sheet piles in front of steel sheet piles. These materials are inert and thus may also be used where stay currents could potentially lead to accelerated corrosion of steel sheet piles. Example vinyl walls are shown in Figure 8-5.

Figure 8-5. Vinyl Sheet Pile Walls.

8.2.2

Wall Construction

The two general construction sequences for sheet pile walls include: (1) driving sheet piles into the ground and then backfilling behind the wall (i.e., “backfilled structure”); and (2) driving sheet piles into the ground and excavating the soil in front of the wall (i.e., “excavated structure”). The sequence of construction for a backfilled structure is illustrated in Figure 8-6 and generally proceeds as follows: Step 1: Excavate the in-situ soil in front and back of the proposed structure, if necessary. Step 2: Drive the sheet piles. Step 3: Backfill behind the wall to the level of the anchor and install the anchor system. Step 4: Backfill to the top of the wall. For cantilever type of backfilled sheet pile wall (i.e., without anchors), the construction sequence is the same except anchors are not installed. FHWA NHI-07-071 Earth Retaining Structures

8-7


Figure 8-6. Sequence of Construction for a Backfilled Sheet Pile Structure (after Das, 1990).

The sequence of construction for an excavated structure is illustrated in Figure 8-7 and generally proceeds as follows: Step 1: Step 2: Step 3: Step 4:

Drive the sheet piles. Backfill (or excavate) to the anchor level and install the anchor system. Backfill to the top of the wall. Excavate the front side of the wall.

For cantilever type of excavated sheet pile wall (i.e., without the anchors) the construction sequence is the same except anchors are not installed. There are three different sheetpile driving methods: (1) pitch and drive; (2) panel driving; and (3) staggered driving. The pitch and drive method is the simplest way of driving but can be used only for loose soils or short piles (Figure 8-8). This method involves driving each sheet pile to full depth before installing the next one. Panel driving is used for dense sands and stiff cohesive soils or for cases where there are possible obstructions in the ground (Figure 89). Panel driving enables greater control of verticality and alignment. Staggered driving is FHWA NHI-07-071 Earth Retaining Structures

8-8


Figure 8-7. Sequence of Construction for an Excavated Sheet Pile Structure (after Das, 1990).

Figure 8-8. Sheet Pile Pitch and Drive Method (after TESPA, 2001).


8-9


Figure 8-9. Sheet Pile Panel Driving (after TESPA, 2001).

used for difficult soil conditions and is combined with panel driving. The piles are installed between guide frames and then driven in short steps as piles 1, 3, and 5 first then piles 2 and 4 (Figure 8-10). If the soil is very dense, sheet piles 1, 3, and 5 are reinforced at the toe. In this case, these sheet piles are always driven first and piles 2 and 4 in the second stage. Depending on the hammer type sheet piles may be installed in variety of in-situ soils, however; it should be noted that sheet piles cannot penetrate very hard layers. The most favorable in-situ soil conditions for vibratory hammer driving include rounded sand and gravel and soft soils. The least favorable soil conditions are dense, angular soils. Also, dry FHWA NHI-07-071 Earth Retaining Structures

8-10


soils result in greater penetration resistance than those which are moist, submerged, or fully saturated. Granular soils have a tendency to compact as a result of the vibrations of the hammer, which can increase the penetration resistance thus reducing the efficiency of the installation. The most favorable in-situ soil conditions for impact hammer driving include soft soils such as silts and peats, loosely deposited medium and coarse sands and gravels without rock inclusions. The least favorable soil conditions are dense sands and gravels and hard clays. As is the case with vibratory hammer driving, it is harder to drive sheet piles in dry soils than moist or fully saturated soils.

Figure 8-10. Sheet Pile Staggered Driving (after TESPA, 2001). FHWA NHI-07-071 Earth Retaining Structures

8-11


8.2.3

Cost

For walls up to 15 ft, the typical cost range of cantilevered sheet pile walls (i.e., without anchors) is about $15 to $40/ft2 of exposed wall face with the lower end cost associated with temporary applications in which sheet piles are rented. Because of increased productivity, sheet pile walls installed using vibratory hammers may be less expensive than for walls installed with impact hammers. Vibratory hammers are also very effective in pulling sheet piles out of the ground. The wall facing is not included in the cost range given above. Costs for wall facings are described in Chapter 10. 8.2.4 Construction Inspection Inspector responsibilities for sheet pile wall construction are summarized in Table 8-1. Table 8-1. Inspector Responsibilities for a Typical Sheet Pile Wall Project. CONTRACTOR SET UP Review Plans and Specifications Review Contractor’s schedule Review in-situ and backfill property test results (i.e., grain size, Atterberg limits, unit weight, and shear strength) Discuss anticipated ground conditions and potential problems with Contractor Confirm that Contractor’s pile installation equipment and its size and type is consistent with Specifications Confirm that the layout and dimensions of sheet pile consistent with Plans Confirm Contractor stockpile area consistent with Plans Review corrosion protection requirements of metallic units (i.e., epoxy coating, paint) and confirm consistency with Specifications EXCAVATION, STEEL SHEETING AND REINFORCING ELEMENTS Confirm that excavation of slopes and/or structural excavation support is consistent with the Plans Confirm that disposal of excavated material is performed according to Specifications Confirm that during installation sheet pile with ball end driven first Confirm embedment depth of sheet pile Confirm the quality of interlock between adjacent sheet piles by visual inspection Confirm that sheet piles are driven according to the Specifications Visually inspect the top of sheet piles to asses any possible damage after installation Report to the Engineer cave-ins, excessive yielding of sheet pile, threatened flooding of excavation, unanticipated subsurface conditions Confirm the reinforcing elements used are consistent with Plans and Specifications(1) FHWA NHI-07-071 Earth Retaining Structures

8-12


Confirm that reinforcement elements are placed and connected to the sheet pile as shown in Plans(2) Confirm that reinforcing element is covered with backfill material to heights consistent with Plans(1) DRAINAGE SYSTEMS AND BACKFIILL Confirm backfill materials used are consistent with Specifications Confirm placement and compaction of backfill material are consistent with Specifications Confirm alignment and elevation of drainage systems POST INSTALLATION Verify pay quantities Note: (1)Anchored sheet pile structure and (2)Excavated structure.

8.3

SOLDIER PILE AND LAGGING WALLS

8.3.1

General

Soldier pile and lagging wall systems (also known as post and panel walls) are commonly used for temporary excavation support (Figure 8-11) in dense or stiff soils where sheet pile walls may not be suitable. They are also used frequently for permanent earth retaining structures (Figure 8-12). Soldier pile and lagging walls consist of soldier piles usually set at 5- to 10-ft spacing, and lagging which spans the distance between the soldier piles. The lagging is used to retain the soil face from sloughing and transmit the lateral earth pressure to the soldier piles. Included in this category of walls is the master pile (or king pile) wall system, shown in Figure 8-1(c), which consists of discrete vertical H-pile sections interlocked and alternating with steel sheet pile sections. The most common soldier piles are rolled steel sections, normally H-pile or wide flange sections. However, soldier piles can be almost any structural member such as pipe or channel sections, cast-in-place concrete, or precast concrete. Figure 8-13 shows several types of soldier piles. For temporary walls, lagging is most commonly wood, but may also consist of light steel sheeting or precast concrete. Cast-in-place concrete sheeting (or facing) is generally used for permanent wall applications, but such construction typically requires the use of temporary sheeting to support the soil face during excavation and concrete placement operations.


8-13


Figure 8-11. Soldier Pile and Lagging Wall and Bracing for Temporary Excavation Support.

(a)

(b)

Figure 8-12. Permanent Soldier Pile and Lagging Walls for (a) Roadway Embankment, and (b) Roadway Cut. FHWA NHI-07-071 Earth Retaining Structures

8-14


(a)

(b)

(c)

(d)

Figure 8-13. Types of Soldier Piles (a) Wide Flange Section, (b) Pipe Section without Anchor, (c) Double Channel Section, and (d) Pipe Section with Anchor (after Xanthakos, et al. 1994). In some cases, it may be more economical and rapid to attach the lagging boards to the outside of the steel beams (i.e., contact lagging) using small steel attachment devices. Once the attachment device is welded to the front of the beam, lagging board placement can be done rapidly (see Figure 8-13 b and Figure 8-14). Where soils are particularly stiff, this option may be considered. Lagging may be omitted in hard clays, soft shales and soils with natural cementation, provided that the soldier piles are installed at relatively close spacing and with adequate steps taken to protect against erosion and spalling of soil at the face. Soldier piles can be either driven or drilled in to the soil. Driven soldier piles can be installed in dense or stiff soils where sheet pile walls may not be suitable. Drilled in soldier piles can be installed in any soil type and even into bedrock. For a soldier pile and lagging wall to be viable, the soil in between soldier piles has to be free standing to allow for lagging installation. Soft clay and loose, cohesionless sands and silts may not be free standing. FHWA NHI-07-071 Earth Retaining Structures

8-15


Figure 8-14. Contact Lagging. The advantages of soldier pile and lagging walls include: • • • • •

availability of a wide range of pile sections to match loading requirements; ability to resist high bending moments; can be installed through hard soils and rock via predrilling; can support large vertical loads; and piles can be spliced.

Some of the limitations of soldier pile and lagging walls include: • • • •

they are generally free draining and not suitable for applications where it is necessary to maintain the groundwater level behind the wall; steel elements in direct contact with soil are more susceptible to corrosion; greater ground displacement in comparison to stiffer wall systems (i.e., slurry walls, tangent pile walls, etc.), especially if used with relatively soft ground; and excavation for placement of the lagging between soldier piles increase risk of ground loss.


8-16


8.3.2

Wall Construction

8.3.2.1 Construction Sequence The construction procedures presented below follow those described by Goldberg et al. (1976) for soldier pile and lagging walls which are braced (using struts or rakers). Anchored soldier pile and lagging walls are discussed in Section 8.8 and 8.10. The soldier piles are installed by driving, or by concreting them within pre-drilled holes (Figure 8-15). After installation of the piles, the excavation is performed from the top down. If braces are used then the excavation is performed to the first support level, placing lagging as the excavation proceeds. Brackets are then attached to the soldier pile to support the wale (if used), and the wale is then placed in position and connected to the soldier piles. The brace is cut slightly short to facilitate placement. This extra space is closed by plates and wedges when the final connection is made. The above sequence of excavation, and installation of lagging, walls and braces is continued until the required bottom of excavation elevation is reached. 8.3.2.2 Preloading Preloading of supports is conducted for braced soldier pile and lagging walls. Preloading is usually required for installation of bracing members since it results in more reliable load determination and load distribution within the structural support system, and it effectively reduces ground displacements adjacent to the excavation. Preloading is particularly important when the excavation is located near structures or other facilities which may be damaged by settlement or lateral ground movement. Cross-lot bracing members are commonly preloaded when they are installed. Rakers and corner braces are not generally preloaded due to the more complicated skew connections required, however, if necessary, these connections can be made allowing for preloading. Preloading of internal braces is accomplished by loading hydraulic jacks to the desired load followed by securing the member with steel blocking, steel wedges and welding. One procedure is to jack to the desired load, and then to drive steel wedges between the member and the wale until the jack load is essentially zero. A second procedure is to weld the connection tight while maintaining the jack load and then drop the pressure in the hydraulic jack, thus transferring the load through the connection to the wale. The second procedure may result in additional wall movement as the load is transferred, although the magnitude of movement is generally small.


8-17


(a)

(b)

Figure 8-15. Soldier Pile and Lagging Wall: (a) Drilling for Pile Installation, and (b) Excavation between Soldier Piles in Cohesive Soil for Installation of Lagging.

High preloads may cause over stressing of struts because of unforeseen job conditions or temperature effects. Accordingly, the general practice is to preload bracing members to about 25 to 50 percent of the bracing loads. This preload removes the slack from the support system and at the same time reduces the risk of over stressing. Larger preloads, up to as much as 80 percent of the bracing load, may be desirable to further reduce ground movements and protect adjacent structures from settlement. 8.3.2.3 Strut Removal and Rebracing Temporary support bracing, or struts, are generally removed during or after construction of the permanent structure elements (i.e., invert and/or roof slabs). Also, it may occasionally be necessary to remove and reinstall temporary bracing (rebracing) to maintain lateral support while installing the permanent structure elements. Strut removal (and rebracing) may be an additional source of wall and ground displacement. Factors controlling the amount of displacement are the wall stiffness, the properties of the retained soil, the span distance between remaining braces, and the quality and the compaction of backfill between the final structure and the excavation support wall. 8.3.2.4 Concrete Backfill Drilled-in soldier piles require the predrilled holes to be filled with concrete backfill. General design recommendations for concrete backfill include the use of structural concrete from the bottom of the hole to the excavation base and lean-mix concrete for the remainder of the hole. The design concept is to provide maximum strength and load transfer in the FHWA NHI-07-071 Earth Retaining Structures

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permanently embedded portion of the soldier pile while providing a weak concrete fill in the upper portion which can easily be removed and shaped to allow lagging installation. However, contractors often propose to use lean-mix concrete backfill for the full depth of the hole to avoid the delays associated with providing two types of concrete in relatively small quantities. 8.3.3

Cost

Cost for soldier pile and lagging walls up to 15 ft in height is typically about 15 to 20/ft2 of exposed wall face. One of the factors affecting the cost of soldier pile and lagging walls is the equipment that is used for soldier pile installation. As mentioned previously, soldier piles can either be driven or drilled in to the soil. Drilling is typically more expensive than driving the pile; however, wall alignment can be achieved easier with drilling than with driving. Compared to driving, the use of predrilling can add $10 to $20 per ft of soldier pile. 8.3.4 Construction Inspection The inspector responsibilities for a typical soldier pile and lagging wall are summarized in Table 8-2. Table 8-2. Inspector Responsibilities for a Typical Soldier Pile and Lagging Wall Project. CONTRACTOR SET UP Review Plans and Specifications Review Contractor’s schedule Review test results and certifications for materials (e.g., steel soldier piles, wood lagging, structural concrete, lean-mix concrete) Discuss anticipated ground conditions and potential problems with Contractor Check overall condition of Contractor’s pile driving or drilling equipment Confirm that the layouts and dimensions of the wall consistent with Plans Confirm that Contractor stockpile and staging area consistent with locations on Plans Review Contractor’s survey results against Plans SOLDIER PILE Review corrosion protection requirements of soldier piles and confirm that Contractor is following these requirements Confirm soldier pile driving operations or drilling consistent with Specifications Inspect driven soldier piles to assess damage at the top of the piles Inspect the integrity of soldier pile holes before they are filled with concrete backfill Confirm that loose material is removed from the bottom of the hole (for drilled-in) Confirm embedment depth of soldier piles consistent with Plans and can be achieved without damage to soldier pile Check that soldier pile is properly aligned before concrete placement and remains aligned during concrete placement FHWA NHI-07-071 Earth Retaining Structures

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LAGGING Confirm that lagging is placed at an appropriate time after excavation to ensure no local soil failure occurs Report observations of over cutting and/or significant soil sloughing to Engineer Confirm placement of lagging from top-down in sufficiently small lifts based on Specifications Confirm that gap is left between vertically adjacent lagging boards DRAINAGE SYSTEMS AND BACKFIILL Confirm backfill materials used are approved by Engineer Verify alignment and elevation of drainage systems Confirm that delivery, storage, and handling of the prefabricated drainage composite (i.e., geocomposite) is performed consistent with Plans and Specifications. Confirm that geocomposite is not damaged in any way, while it is being installed and is not exposed to excessive dust that could potentially clog the system Confirm that geocomposite strips are placed and secured tightly against the lagging with the fabric facing the lagging Confirm placement of drainage aggregate in front of the lagging in horizontal lifts Confirm that construction of drainage aggregate closely follow the construction of precast facing elements Confirm that perforated collector pipe is placed within the permeable material to the flow line elevations and at the location shown on Plans POST INSTALLATION Verify pay quantities

8.4

SLURRY WALLS

8.4.1

General

A slurry wall (or diaphragm wall) is a structural, cast-in-place concrete wall constructed by tremie placement of concrete in a pre-excavated, slurry-filled trench, as shown in Figure 816. The wall is constructed in a series of panels which interlock to form a stiff continuous structure. Commonly, trenches 2 to 3 ft wide are excavated in lengths of 10 to 20 ft. Slurry walls are used at sites where it is necessary to restrict ground displacements adjacent to the excavation. This is a particular concern when the excavation is in close proximity to a building or other structure which is founded above the bottom of excavation level. As a relatively rigid excavation support system, a slurry wall typically results in considerably smaller ground displacements than a sheet pile or soldier pile and lagging support system.


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Figure 8-16. Typical Construction Sequence for a Slurry Wall: (a) Excavation; (b) Insertion of Steel Tubing (End Stops); (c) Placement of Reinforcement Cage; and (d) Concrete Placement (Xanthakos, 1994). FHWA NHI-07-071 Earth Retaining Structures

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Slurry walls also provide a relatively water tight excavation support wall. A watertight excavation may be required when: (1) groundwater lowering outside the excavation limits may lead to potentially damaging settlement of nearby structures or other facilities; (2) it is not practical to dewater a site (e.g., adjacent to an open body of water); and (3) seepage gradients initiated by dewatering operations may risk migration of existing groundwater contamination plumes. With penetration into an underlying low permeability stratum or with sufficient penetration below the bottom of the excavation, a slurry wall provides an effective seepage barrier that can preserve groundwater conditions outside the excavation. It is noted, however, that other wall types such as DMM walls can provide a seepage barrier at a lower cost. Advantages of slurry walls include: • • • •

•

boulders, cobbles, and other obstructions are removed as part of the trenching operation; can be used solely for temporary support of excavations, or serve both as a temporary excavation support and as the permanent structural wall; eliminates the need for a costly cast-in-place concrete interior wall; allows a reduction in the width of the excavation; constructed as top-down construction which may be advantageous for projects that require a minimum duration for surface disturbance (i.e., construction of a depressed underpass at a busy intersection); and minimal vibrations during construction.

Limitations of slurry walls include: • • •

use of specialized equipment and construction methods; will generate significant spoils that needs to be disposed; and relatively high cost in comparison to other wall systems.

Although previously described as a general sequence of construction for a wall system, the term “top-down construction” may also be used to describe a method of constructing excavations (usually in urban environments) that allow the structure (e.g., tunnel) to be built within the excavation as the excavation proceeds rather than starting construction after the bottom elevation of the excavation is reached. Oftentimes, a slurry wall is used for this application (although other diaphragm wall types can be used). The general sequence for top-down construction includes:


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1. A retaining wall (typically a concrete diaphragm wall) is installed before excavation begins. 2. Soil is excavated to just below the elevation of the roof slab of the underground structure. Struts are installed to support the walls. 3. The roof slab is constructed with access openings on the slab to allow work to proceed downwards. 4. The next slab level is constructed and this process proceeds downwards until the base slab is completed. 5. Side walls of the underground structure are constructed upwards, followed by removal of the struts. Access openings on the roof slabs are sealed. 6. After the underground structure is completed, soil backfill is placed to the top strut level and then this strut is removed. Backfilling to the ground surface is then completed. The two primary types of slurry walls are the conventional reinforced concrete wall and the Soldier-Pile-Tremie-Concrete (SPTC) wall. Other wall types include the Precast-ConcretePanel wall and the Post-Tensioned Concrete wall. These walls are briefly described below. 8.4.2 Slurry Wall Types 8.4.2.1 Conventional Reinforced Concrete Wall A conventional reinforced concrete slurry wall includes steel reinforcement cages that are placed in the slurry trench before the concrete tremie pour. The reinforcing bars are sized to resist bending principally in the vertical direction between bracing levels. Supplemental horizontal bars may be designed to serve as “internal walers” or to distribute forces around inserts or openings. External wales are then eliminated provided that lateral bracing supports are provided at panel joints. This method of reinforcement is currently most popular and is used extensively throughout the U.S., (Figure 8-17). Panel end joints are formed by stop end pipes, shown in Figure 8-16, or other suitable forming devices. 8.4.2.2 Soldier-Pile-Tremie-Concrete (SPTC) Wall SPTC walls are becoming increasingly popular, particularly in deep excavations requiring high bending resistance. These walls are also referred to as “Soldier Pile and Concrete Lagging Walls” owing to the similarity with soldier pile and lagging walls. SPTC walls commonly include vertical wide flange sections set in a slurry-stabilized trench. In some cases, a reinforcing bar cage is placed between the soldier piles to transfer earth and water loads laterally to the soldier piles. Alternatively, the concrete may be designed to span (arch) FHWA NHI-07-071 Earth Retaining Structures

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3 ft

2 ft

3 in

6.5 to 26 ft

Figure 8-17. Conventional Reinforced Concrete Wall (Tamaro, 1990).

between the soldier piles which eliminates the need for reinforcing bar cages between soldier piles. In this case, soldier piles should be spaced at least 4.5 ft or more on center to facilitate concrete placement. This slurry wall type is relatively watertight, has significant strength in the vertical direction, and allows for relatively easy connection for temporary cross lot bracing and wales (Figure 8-18). 8.4.2.3 Other Slurry Wall Types A variation of the above wall types is the Precast-Concrete-Panel wall in which precast concrete wall elements are placed in a slurry-stabilized excavated trench. This method produces the best quality finished wall but is limited in use by cost, transportation and handling length limitations, and other specific site constraints. Wall panels are best cast on site adjacent to the work and installed directly into an oversized trench excavation. This type of wall requires the use of cement-bentonite slurry which will eventually harden in the void between the wall and soil. Rubber water stops are usually required between panels since the cement-bentonite joints are subject to drying and shrinkage cracking (Figure 8-19). The precast panels are temporarily suspended within the excavation until the cement-bentonite hardens. Wall installation usually proceeds in a linear fashion with precast panel installation following closely behind trench excavation. The use of a Precast-Concrete-Panel wall becomes less practical for deeper excavations and at locations where utility crossings or obstructions are expected. FHWA NHI-07-071 Earth Retaining Structures

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3 ft

2 ft

6.5 to 16 ft

3 to 6 in 2.5 ft

1.5 ft

Figure 8-18. Soldier-Pile-Tremie-Concrete (SPTC) Wall (Tamaro, 1990).

5 to 10 ft

Figure 8-19. Precast-Concrete-Panel Wall (Tamaro, 1990).

Another wall type used successfully in Europe but which has had limited use in the U.S. is the Post-Tensioned-Concrete wall (Figure 8-20). For this type of wall, post-tensioning provides increased bending resistance. However, major disadvantages of this type of wall are its higher cost and the need for specialized construction techniques.


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3 ft

2 ft

6.5 to 26 ft

Figure 8-20. Post-Tensioned-Concrete Wall (Tamaro, 1990).

8.4.3

Wall Construction

8.4.3.1 Slurry The slurry used to stabilize the excavated trench consists of bentonite and water. When mixed with water, the bentonite (which comes as a powder, chips or pellets) forms a colloidal suspension (or slurry). To date, the use of additives in the bentonite slurry has been limited. Polymer slurries are generally not used for slurry wall construction since they are not as effective as mineral slurries for supporting large size excavations in coarse granular soils. To maintain a fluid slurry until concrete is completed, the slurry must be circulated and agitated. Desanding devices are typically used to remove a sufficient amount of suspended soil so that the slurry can be used two or more times prior to disposal. The contractor should verify that a heavily contaminated slurry suspension, which could impair the free flow of concrete, has not accumulated in the bottom of the trench. Tamaro and Poletto (1992) recommend the following specifications for fresh bentonite slurry (at the beginning of excavation): • • •

Minimum specific gravity of 1.03; Minimum viscosity of 32 seconds measured by the Marsh Cone Funnel; and pH between 7 and 11.


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Prior to placement of concrete, the bentonite in the excavated trench should meet the following requirements: • • •

Sand content not more than 5 percent, measured at about 5 ft above the bottom of the trench; Specific gravity not more than 1.10; and Viscosity not more than 50 seconds.

Additional information on material requirements and the use of slurry is provided in O’Neill and Reese (1999). 8.4.3.2 Concrete The concrete must be a free-flowing mix capable of displacing the bentonite slurry and bonding to the reinforcement. Tamaro and Poletto (1992) recommend the following for concrete: • • •

concrete strength of 3,000 to 5,000 psi; well graded aggregate less than 0.75 in. in maximum dimension; and eight (8) in. slump.

In addition, Goldberg, et al. (1976) recommend a water/cement ratio less than 0.6, a sand content of 35 to 40 percent of the total weight of aggregate, and a cement content of at least 25 pcf of the tremie concrete. Premature stiffening of the cement may negatively affect the tremie operation. Retarders are sometimes added to the mix to keep the concrete workable during the entire pour. Some of the retarders, however, may reduce the strength. The retarders most commonly used are discussed in Xanthakos (1994). 8.4.3.3 Reinforcement The slurry wall reinforcing can be in the form of a rebar cage, or a combination of a cage and vertical wide flange sections. The rebar cage can be prefabricated and assembled either in a shop or at the site. The minimum bar spacing is generally 6 in. for vertical bars and 12 in. for horizontal bars (6 in. for horizontal spacing with internal wales). Selection of minimum bar spacing should also consider reduced spacing caused by splices, vertical picking bars used to lift the cage, and other items (e.g., inclinometer, pipes for non-destructive testing) tied to the cage. There must be sufficient space between the bars to enable the free flow of tremie concrete and FHWA NHI-07-071 Earth Retaining Structures

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permit scouring of the reinforcing bars to remove slurry and provide good bond between concrete and the reinforcing bars. The rebar cages, as well as inserts (e.g., sleeves for soil or rock anchors, casing for instrumentation, etc.), should be secured with tie wire. Spacer devices should be used on the outside of the rebar cage to provide a minimum concrete cover of 3 in. 8.4.3.4 Equipment A wide variety of trench-excavating equipment is used in slurry wall construction, including backhoes, draglines, clamshells, bucket scrapers and others. Buckets are raised and lowered by cable or kelly bar, and are opened or closed hydraulically or by cable (Figure 8-21a). To penetrate hard layers, percussive tools can be used either to assist in clamshell excavation or as independent excavating tools. The cable-operated bucket, hung from a crane, is the excavation equipment most commonly used by U.S. contractors. Another type of slurry wall excavation equipment is the “hydromill” which was developed by French and Italian equipment manufacturers (Figure 8-21b). This excavator is basically a grinding device consisting of two milling heads rotating in opposite directions about axes perpendicular to the trench. The rotating heads excavate soil and soft rock from the bottom of the trench, and the excavated material is lifted by suction to the surface where the soil is removed from the bentonite by sand separators. For excavation in rock, heavy drop chisels, chisel drills, or large diameter roller bits can be used. Pre-drilling can also be used to facilitate advancement of the clamshell through hard layers. 8.4.3.5 Procedures Construction of a slurry wall includes five primary elements: placement of guide walls, trench excavation, placement of reinforcement, concreting, and the formation of joints. Guide Walls The construction of a slurry wall usually begins with the installation of guide walls. The purpose of the guide walls is to: (1) prevent caving of the trench wall in the uppermost part of the excavation; (2) align the trench; (3) contain the slurry; and (4) support suspended precast elements in Precast-Concrete-Panel walls, or reinforcing steel in Cast-In-Place walls. Figure 8-22 shows a cross-section of a guide wall.


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(a)

(b) Figure 8-21. Slurry Trench Excavation Equipment: (a) Hydraulically Operated Clamshell Bucket, and (b) Hydromill. FHWA NHI-07-071 Earth Retaining Structures

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4 – 5± ft

4 – 5 ± ft

Figure 8-22. Cross-Section of a Guide Wall (a) Compact Cohesive Soil, (b) Loose Cohesionless Soil (Goldberg, et al., 1976).

The guide walls are usually 6 to 12 in. thick and 3 to 6 ft deep. To provide the necessary support for suspended rebar cages or pre-cast panels, the guide walls should be cast on a stable subgrade. FHWA NHI-07-071 Earth Retaining Structures

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Cohesive soil is used for backfill behind the guide walls. Cohesionless soils may be mixed with cement to prevent undermining and to increase stability when used as guide wall backfill. Trench Excavation In conventional bucket excavation, the bucket brings the material to the surface, discharges its load, and then is lowered back into the trench. With direct or reverse circulation equipment, the material is broken up into smaller particles so that it may be suspended in the bentonite slurry, which is circulated to the surface, screened and desanded. The cuttings are brought to the surface by suction and/or air lift through suction pipes, or the excavation tool itself. The stability of the trench is maintained by the slurry pressure on the trench wall, and soil arching. Also, local penetration of the bentonite into pervious soils will provide some cohesion that helps to prevent spalling. The slurry level in the trench is maintained at an elevation at least 4 ft higher than that of the groundwater table. A membrane or a “mudcake” is formed against the walls of the trench by a combination of hydrostatic pressure, osmotic pressure, and electrolytic properties of the colloid. This mudcake will maintain the pressure against the trench walls and prevent fluid losses through pervious materials. The verticality of the wall is checked as the excavation advances. If the excavation is found to be out of vertical tolerance, the trench can be backfilled with lean concrete and the excavation operation repeated. This technique can also be used to fill cavities formed in the side wall of the trench due to caving. Reinforcement Prior to placement of the rebar cage in the slurry filled trench, the trench bottom should be sounded by a weighted tape, or other means, to verify the depth of the trench and assess the cleanliness of the trench bottom. Also, the slurry in the trench should be sampled and tested to assure that it meets specification requirements. If necessary, slurry circulation and trench bottom cleaning operations should be continued to meet the specified requirements. The rebar cage and end stops should be placed in the trench as soon as possible after the trench is cleaned and inspected (Figure 8-23). Figure 8-24 shows a reinforced panel in castin-place slurry wall.


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(a)

(b)

Figure 8-23. Slurry Wall Reinforcement Cage (a) On Fabrication Bed, Showing Styrofoam Knock-Out Panel, and (b) During Lifting for Installation into Slurry Filled Trench.

Figure 8-24. Reinforced Panel in Cast-In-Place Slurry Wall (Tamaro and Poletto, 1992). FHWA NHI-07-071 Earth Retaining Structures

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In the Soldier-Pile-Tremie-Concrete wall, the soldier piles, together with the reinforcing cage, are set within the excavated trench prior to concreting. An alternative approach is to set the soldier piles in pre-augered holes and then excavate and place tremie concrete between consecutive piles. Concreting The concrete should be placed as soon as practical after installation of the reinforcement cage, and concrete placement should proceed continuously until completion of the slurry wall panel. Concrete placement is performed through one or more tremie pipes lowered to the bottom of each panel (Figure 8-25). Two or more tremie pipes may be used for long panel lengths and for SPTC walls. The tremie pipe must remain embedded in fresh concrete a minimum of 6 ft and a maximum of 15 ft. The tremie concrete displaces the bentonite slurry progressively as it rises uniformly to the surface. The concrete should be sampled and tested at the start of placement and at defined intervals during placement, to verify that the delivered concrete meets the specified requirements.

Figure 8-25. Slurry Filled Trench with Tremie Pipes Just Prior to Concrete Placement. FHWA NHI-07-071 Earth Retaining Structures

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End Joints A construction joint is provided between two adjacent panels. This joint should allow excavation of the new panel without significant disturbance to the previously poured panel. It should be watertight and capable of transferring shear and compressive stresses. The joint can be formed using different configurations and details, but generally consists of a steel tube, steel plate, or steel beam (Xanthakos, 1979). Unless permanent steel piles are used, the end stops are removed after the initial set of the concrete. For deep walls, it may be necessary to partially extract the end stops while concrete is placed in the upper part of the wall panel. The end stops should be removed in a smooth and continuous manner. 8.4.3.6 Types of Supports Slurry walls can be supported using either internal bracing or soil or rock anchors. Similar to soldier pile and lagging walls, the support elements are generally preloaded when they are installed to reduce ground displacements behind the wall, and to obtain more reliable and more uniform loading in the support elements. 8.4.4

Cost

The cost of a slurry wall is typically between $60 and $100/ft2 of exposed wall face. The use of braces or ground anchors would add to the cost of the wall. Typical mobilization costs for slurry walls are around $50,000. Slurry walls may be cost competitive compared to other wall systems when at least two of the following conditions can be met (Godfrey, 1987): • • • • •

slurry wall can be used as the permanent wall; slurry wall eliminates the need for underpinning; slurry wall eliminates the need for dewatering; natural or man made obstructions restrict the use of conventional sheeting systems; or slurry wall can be used to support vertical loads.

8.4.5 Construction Inspection Slurry (diaphragm) wall construction requires relatively detailed construction inspection due to the use of specialized equipment and materials. These walls are usually constructed by specialty contractors. FHWA NHI-07-071 Earth Retaining Structures

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The construction of a wall panel for a slurry (diaphragm) wall is performed in a narrow trench. For this reason, construction tolerances related to panel width, depth of wall, and alignment of the wall are critical. Proper construction of the guidewalls is essential to ensure that accurate alignment is maintained. Guide walls should be cast against compact subgrade. On-site quality control of the slurry and concrete is also necessary. The inspection of slurry should include: (1) documentation of bentonite source and quality; (2) records indicating age of slurry (i.e., time for bentonite hydration); and (3) results of on-site tests for slurry viscosity, specific gravity, pH, and sand content. During placement of concrete, plots of concrete volume placed versus the rise of the concrete within the excavated trench should be recorded. This information can be used to evaluate whether a cave-in has occurred. After the wall has been exposed due to excavation, the wall should be checked to verify that excessive seepage is not occurring through vertical panel joints or through any openings in the wall face. Inspection responsibilities for slurry walls are summarized in Table 8-3.

Table 8-3. Inspector Responsibilities for a Typical Slurry Wall Project CONTRACTOR SET UP Review Plans and Specifications Review Contractor schedule Discuss anticipated ground conditions and potential problems with Contractor Check overall condition of Contractor equipment Verify that the layouts and dimensions of the wall consistent with Plans Review Contractor stockpile area GUIDE WALL Verify that no caving occurs in the uppermost part of the trench excavation Confirm alignment of the guide wall trench Confirm consistency of Contractor’s guide wall details, width, and height with Plans Observe placement of concrete to form the Guide Wall Confirm that inside face of the guide wall does not have any ridges or abrupt changes and the guide wall face and top does not vary horizontally or vertically from a straight line or specified profile in the Plans Confirm that the distance between inside faces of the guide wall is within the width of slurry wall plus the distance specified in Specifications Confirm consistency of placement and compaction of backfill soil behind the guide wall with Specifications Confirm that guide wall is effectively containing the slurry without losses Confirm that guide walls are removed at the end of construction


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TRENCH EXCAVATION Confirm that no cavities exist in the side of the trench Inspect the bottom of the trench immediately after excavating to the final depth and confirm the conditions against the Specifications Confirm circulation, screening, and desanding of trench material Inspect bottom elevation of the trench immediately after each cleaning and desanding Confirm that each excavated panel is filled and maintained at all times with stable suspension of slurry Test slurry in the bottom and top of the trench for density, viscosity, pH, and sand content with the frequency specified in Specifications. Report test results to the Engineer Check the depth and verticality of the wall to determine compliance with Plans Perform fluid loss measurements in the trench at the beginning or end of every shift as directed by the Engineer. Report measurement results to the Engineer REINFORCEMENT Confirm that bottom of trench excavation has been reached prior to reinforcement cage placement Confirm the condition of reinforcement cage prior to installation Confirm installation of casings and sleeves as indicated on the Plans Confirm construction of joints between two adjacent reinforcement panels are consistent with Specifications CONCRETING Inspect the bottom elevation of trench immediately before placement of concrete Test concrete in the trench to assure it meets specification requirements Confirm that concrete is placed in slurry excavations within specified time frame after placement of reinforcing and proceeds uninterrupted until completion Confirm the placement of concrete occurs right after installation of reinforcement cage Confirm that tremie pipes remain embedded in fresh concrete during concreting Confirm appropriate handling of slurry as concrete is placed in as specified in Specifications POST INSTALLATION Verify pay quantities Note: Submit daily reports to the Engineer documenting: (1) trench excavation; (2) verticality verification of each section of panel; (3) slurry testing and modifications to the mix or process; (4) reinforcement placement; and (5) concrete placement.


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8.5

TANGENT/SECANT PILE WALLS

8.5.1

General

A tangent or secant pile wall consists of a line of drilled shafts (also referred to as bored piles) (Figure 8-26). If the bored piles are contiguous, or tangent, to each other, the wall is called a “tangent pile” wall. In an alternate case, referred to as a “secant pile” wall, the pile elements overlap so as to form an interlocking wall. Another variation of this wall type is called an “intermittent wall” in which the piles are installed at a spacing exceeding the pile diameter; this type of wall can be considered only if the ground is stable or secondary elements, such as shotcrete or a cast-in-place facing, is used to provide a continuous wall. Various configurations of bored pile walls are shown in Figure 8-27.

(a)

(b)

Figure 8-26. Tangent Pile Wall a) With Structural Steel Section as Reinforcement, and b) Face of Completed Wall. Tangent pile walls and secant pile walls are stiff continuous walls that are constructed by the top-down method. Similar to slurry (diaphragm) walls, tangent pile and secant pile walls can be used when it is necessary to minimize groundwater lowering outside the excavation or to reduce ground displacements. Also, similar to slurry walls, tangent pile and secant pile walls can be used for either temporary or permanent ground support. Advantages of tangent / secant pile walls include: •

they can be constructed using conventional drilled shaft excavation equipment and procedures; and


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(a)

(b)

(c)

(d)

(e) Figure 8-27. Various Configurations of Bored Pile Walls: (a) Tangent Pile Wall; (b) Staggered Tangent Pile Wall; (c) Secant Pile Wall; (d) Intermittent Pile Wall with Grouted Openings; and (e) Intermittent Pile Wall with Lagging.

•

tangent / secant pile walls may be more suitable than slurry walls at work sites with limited space since less area is needed for slurry containment and treatment, and for fabrication and handling of rebar cages.

Limitations include: • • •

increased seepage through vertical joints; more difficult connections for bracing members, ground anchors and attached slabs; and rough and irregular exposed face.


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8.5.2

Wall Construction

Bored piles for a tangent or secant pile walls are usually installed using auger type drill rigs. Use of temporary casing may be necessary when penetrating into unstable soil strata, which may be susceptible to squeezing or caving. A faster and more economical construction method, however, involves rotary drilling and the use of bentonite (or polymer) slurry to keep the drilled-hole stable and remove the excavated material to the surface by reverse circulation (FHWA, 1988). Figure 8-28 illustrates the typical construction sequence for tangent pile and secant pile walls. The piles should be installed in a staggered pattern to avoid disturbing the concrete in an adjacent pile that has not completely cured.

Figure 8-28. Construction Sequence (a) Tangent Pile and (b) Secant Pile Walls (after Xanthakos, 1994).

For a tangent pile wall, shown in Figure 8-28a, the direction of pile installation is from the edges of a section towards its center. This sequence prevents interference between adjacent piles during concreting, and allows all the piles in the section to be installed along the same alignment except possibly the center pile which may have to be displaced slightly to fit in the remaining space and still be tangent with the two adjacent piles. In secant pile walls (Figure 8-28b), alternate piles (numbered 1, 3, 5, etc.) are drilled and concreted first with or without reinforcement. Reinforced piles (numbered 2, 4, 6, etc.) are cut into these piles about one day later after the concrete in the first group has achieved its initial set but before it becomes too hard. Sometimes, reinforcement is provided in every FHWA NHI-07-071 Earth Retaining Structures

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pile; however, this is generally practical only when the piles are reinforced with steel sections. Due to the difficulty in cutting reinforcing bars, it is usually necessary to place rebar cages only in alternate piles. Favorable soil conditions for tangent and secant piles include soft clay and granular silty and sandy soils. Maintaining vertical tolerances for a wall constructed in hard soils is difficult. The wall facing may consist of reinforced shotcrete, pre-cast concrete panels, or cast-in-place concrete. A drainage gallery may be provided behind the concrete facing to intercept and channel any seepage that may penetrate the wall. Grouting is sometimes performed behind the joints between piles to reduce water seepage. 8.5.3

Cost

Tangent pile walls are typically less expensive than secant pile walls simply because less shafts are required for a given wall length. The cost of tangent pile walls up to 30 ft in height is about $25 to $40/ft2 of wall face and the cost of secant pile walls up to 30 ft is about $30 to $45/ft2 of exposed wall face. One of the factors affecting the cost of tangent/secant pile walls is the drilling. Equipment used for drilling should have adequate capacity to excavate a hole through soft and hard soils, as well as, obstructions and in secant pile walls, through previously installed concrete piles. When hard soils, or other material encountered cannot be drilled using conventional earth augers, more costly alternative equipment has to be used. For very competent hard soils, piles can be installed apart from each other (i.e., intermittent pile wall). This wall type requires fewer piles for a given length than tangent or secant piles, resulting in a smaller cost as compared to tangent and secant pile walls. The cost of an intermittent wall (up to 30 ft in height) is about $20 to $35/ft2 of exposed wall face. The cost of ground anchors and a permanent wall face are additional costs that should be included. These costs are all described in Chapter 10. 8.5.4 Construction Inspection Construction inspection for tangent pile walls and secant pile walls is similar to that for drilled shafts. Inspection includes verifying that drilling techniques are consistent with the soil type and ground-water conditions at the site and that all construction tolerances are maintained. For tangent pile walls, it is important that vertical tolerances are maintained if a relatively watertight wall is required. The inspection responsibilities of tangent pile and secant pile walls are summarized in Table 8-4. FHWA NHI-07-071 Earth Retaining Structures

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Table 8-4. Inspector Responsibilities for a Typical Tangent/Secant Pile Wall Project CONTRACTOR SET UP Review Plans and Specifications Review Contractor schedule Discuss anticipated ground conditions and potential problems with Contractor Check overall condition of Contractor’s equipment Review test results and certificates for materials (e.g., steel reinforcing, concrete) Confirm consistency of location and positioning of each pile with Plans PILE EXCAVATION Observe Contractor’s test pile installation and confirm that drilling technique is consistent with ground conditions Confirm that drilling fluid is tested for density, fluid loss, viscosity, shear strength, sand content, and pH according to Specifications When penetrating unstable soil strata, confirm that Contractor is using temporary casing (if required). Confirm that upon completion of pile excavation that all loose soil and sediment is removed to expose a firm base Inspect pile excavation to its full length where practical and confirm no squeezing or caving exist on the sides Confirm that all drill holes are protected from surface water REINFORCING Confirm that reinforcing is free from rust and mud Confirm that reinforcement is maintained in its correct position during concreting For secant pile walls, confirm that overcutting of previously installed piles occurs approximately one day after installation. Note, this will be site-specific and depends on time required for concrete to reach minimum strengths and type of overcutting drilling equipment used. CONCRETING Confirm that before placing concrete, no accumulation of drilling fluid or other deleterious material exist at the base of the drill hole Confirm that concrete for each pile is from the same source Confirm that method of concrete pouring (e.g., free fall, tremie pipe, or concrete pumps) is consistent with Specifications Confirm that vertical tolerances are maintained as concrete is poured To check for possible necking or loss of ground, verify the actual volume of concrete pored for each pile consistent with calculated volume required If required in the Specifications, observe the pile integrity test on each pile to locate potential defects, necking, soil inclusions and to verify pile length. Report the results to Engineer Confirm that Engineer-rejected piles are replaced by the Contractor WALL FACING Confirm that wall facing is consistent with Plans and Specifications If grouting is required, confirm consistency of grout and grouting operations behind the joints between piles with Specifications and Plans. FHWA NHI-07-071 Earth Retaining Structures

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POST INSTALLATION Verify pay quantities

8.6

JET GROUTED WALLS

8.6.1

General

Jet-grouted walls commonly use cement for grout. A variation of these walls is the lime column wall wherein lime or lime-cement mix is used in a dry or liquid form to stabilize the soil. Jet grouting consists of injecting high pressure fluids into the ground through horizontal nozzles to segregate the soil and mix it with a cementing agent. Segregation of the soil is achieved by the high energy of the jet fluid(s) which may consist of the cementing agent or another cutting fluid. Jet grouting is basically an erosion/replacement process which removes a portion of the soil particles and replaces them with a mixture of soil and grout that has high strength and low permeability when hardened. The jet grouted elements, which make up the wall, are typically either overlapping cylindrical columns or, in some applications, panel elements. Cylindrical column elements are the most common and are formed by rotation of the high pressure fluid jets. Panel elements are formed by allowing the jet to remain stationary, thus cutting and mixing planar jet grouted elements. Figure 8-29 shows various grout column layouts used in construction of jet-grouted walls. The column layout is governed by the intended use of the wall as well as stability considerations. For example, in cases where water tightness is the main concern, as in the case of cutoff walls, then multiple rows of overlapping columns should be used. Multiple rows are also required if the wall is to be designed as a gravity structure. Figure 8-30 illustrates jet-grouted wall applications for excavation, underpinning, settlement control, and water control. The advantages of jet-grouting include: •

jet grouting can be used with a wider range of soil types than any other grouting technique, however, it is not suitable for highly plastic clays (see Figure 8-31).


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Figure 8-29. Grout Columns Layouts on Jet-grouted Walls.

•

since large diameter columns can be created from relatively small boreholes, local obstructions such as timber piles or large boulders can be bypassed or encapsulated into jet-grouted soil mass;

•

jet grouting can be conducted from any suitable access point, and can be terminated at any elevation, providing treatment only in the target zones; and

•

jet-grouting can be performed vertically or (sub)horizontally, above or below the water table.

The limitations of jet-grouting include: •

it is an operator-sensitive construction process, which requires specialized skills and an experienced labor force;

•

it requires specialized mechanical equipment subject to high maintenance costs;

•

due to the high pressure used, there is a possibility of ground heave or lateral movements which may damage adjacent utilities and underground structures;


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Figure 8-30. Jet-grouted Wall Applications for Excavation Support, Underpinning, Settlement Control, and Water Control.

•

spoil handling and removal may be particularly difficult if the jet-grouted soil is contaminated;

•

a cost-effective method for measuring the dimensions of a large number of columns in a routine production project does not currently exist;

•

jet-grouted soil strengths tend to be much more variable than concrete strength as they are strongly influenced by the silt and clay content of the native soils and it is difficult to predict the final strength of the jet-grouted soil during the design stage; furthermore, if groundwater flow velocities are high, the fluid soil-grout mass may experience local removal of the cement (bleeding) prior to its stiffening, and hence variability in quality may be observed;


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Note: 1 mm = 0.04 in

Figure 8-31. Range of Soil Types Treatable by Chemical and Jet-Grouting (after Welsh et al., 1986). •

it requires a higher level of quality control than for other wall systems; and

•

there is no commonly accepted method available to engineers to design jet grouted structures.

8.6.2

Wall Construction

8.6.2.1 Jet Grouting Procedure The basic procedure for jet grouting is shown in Figure 8-32 and summarized below: •

Drill and stabilize a hole to the required depth in the soil to be treated.

•

If separate drilling and grouting equipment are used, withdraw the drilling bit and rods and insert the jet grouting monitor and special grouting rods to the bottom of the predrilled hole. In many jet grouting systems and ground conditions, however, the same equipment is used to drill the hole and then perform the jet grouting. In these cases, after the hole is drilled to the required depth, a steel ball (check ball) is inserted into the drill rods to redirect the fluid jets to the horizontal nozzles and allow jet grouting to begin (Figure 8-33).


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•

Inject jet grouting fluids while slowly rotating and withdrawing the grout monitor and rods. The grouting fluids cut the in situ soil and mix it with the grout to form a jet grouted column. Excess soil-grout spoil is expelled out of the top of the drilled hole by the circulating fluids.

8.6.2.2 Jet Grouting Systems There are basically three jet grouting systems in general use: the single, double, and triple fluid systems. A schematic representation of each system is shown in Figure 8-34. The principal characteristics of each system are described below (Kauschinger and Welsh, 1989 and Bruce, 1994). •

It is the simplest of the three systems.

•

It uses a grout jet to simultaneously erode the soil and mix it with the cement grout.

Figure 8-32. Jet Grouting Procedure (after Pacchiosi, 1985).


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Figure 8-33. Details of Jet Grouting Monitors (a) Single Fluid System and (b) Triple Fluid System. Single Fluid System •

It involves only partial replacement of the soil and therefore results in the lowest volume of spoils as compared to other systems.

•

It produces the smallest column diameter.

•

It can achieve significant compaction outside of the perimeter of the column up to a distance of one-half a column diameter.

•

It is the only method used for horizontal column formation.

Double Fluid System •

It uses a grout jet engulfed in compressed air. The compressed air enhances the cutting ability of the grout jet.

•

The equipment is more complex and susceptible to clogging. The pathway for the air between the inner rod (carrying the grout) and the outer rod must be kept open, or the process will revert to a single fluid system.


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•

A higher degree of soil replacement and a larger column diameter than that of the single fluid system (almost twice the size) can be created. This enhancement is due to the following factors: o The air acts as a buffer between the jet stream and any groundwater present, thus permitting deeper penetration by the jet; o The soil cut by the jet is prevented from falling back onto the jet, thus reducing the energy lost through the turbulent action of the cut soil; and o The cut soil is more efficiently removed from the region of jetting by the bubbling action of the compressed air.

•

A potential drawback of this system is that the soil-grout mix may have a higher air content, and therefore may have a lower strength than those of the other systems.

Triple Fluid System •

Fluids are emitted from two levels. An upper water jet engulfed in compressed air is used to excavate the soil which is then mixed with, or replaced by, a grout jet emitted from a lower port.

•

It permits virtually full replacement of the jetted soil, and provides the largest column diameter. This also results in the largest amount of spoils/cuttings.

•

In this system, unlike the double fluid system, the grout is not injected with air. Hence, there is no problem with high air content in the final jet-grouted soil mass.

•

Fluid circulation is more efficient because each medium is carried in a separate passageway within the rod (the grout in the central core, the water in the middle annulus, and the air in the outer annulus). Hence, this system is the least likely of the three systems to exert high pressures on the surrounding soils.


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Figure 8-34. Schematic of Jet Grouting Systems (a) Single Fluid System, (b) Double Fluid System, and (c) Triple Fluid System. 8.6.2.3 Hydrofracture Potential In general, a jet grouting system conducts incompressible fluids and its success is contingent on maintaining free and efficient fluid circulation. In particular, the cuttings must flow freely from the point of injection up to the ground surface. Otherwise, pressures, up to the jetting pressures may build up in the soil. These pressures could hydrofracture the soil and cause severe lateral soil movement and ground heave (Kauschinger and Welsh, 1989). 8.6.2.4 Jet Grouting Equipment A typical jet grouting set up consists of: (1) a drilling rig; (2) an automated grout mixing plant; and (3) a grout injection plant, which consists of automatic batchers and high-pressure pumps. In multiple fluid systems, additional pumps and an air compressor are used (Figure 8-35). Figures 8-36 through 8-39 show the jet grouting equipment in operation. The drill rig automatically regulates rotation and withdrawal rates of a string of special drill rods and a jet grouting monitor which is mounted at the end of the drill string. The jet grouting monitor is a special tool through which the jet grouting fluids are passed and FHWA NHI-07-071 Earth Retaining Structures

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directed out of the nozzles and into the ground. Figure 8-40 shows the triple rods and the monitor for a triple fluid jet grouting system. Rotary drilling is most commonly used. In coarse-grained soils, which may include cobbles and boulders, rotary percussion is more suitable. The mast length of the drilling rig is an important consideration on construction sites with overhead obstructions. Controlled jet grouting creates a spoil material during the erosion and mixing process. The volume of soil-cement spoil can be predicted from the injected volumes. The spoil usually contains significant cement content and gains strength over time. Within 12 hours, it can typically be handled as a firm to stiff clay and can be used as a construction material or be carried away from the site for disposal. If jet grouting is used in contaminated ground, special handling procedures may be required. 8.6.2.5 Jet Grouting Materials Neat, rapid setting, cement grout is typically used for jet grouting, although chemical grouts can also be used. Grout viscosity and rigidity should be low to allow maximum penetration. Portland Cement Types I, II or III are used. Type I is the most economical and is, therefore, used when possible. Water/cement ratios of 1:1 to 2:1 are commonly used. Where high strength is required, ratios as low as 0.6:1 may be used. Potable water is normally used. Although not recommended, salt water is sometimes allowed, provided no steel reinforcement is used. Fly ash is sometimes added to the cement grout in ratios of cement:fly ash between 1:1 and 1:10 by weight. Where low permeability is needed, bentonite additives are often used. The addition of 2 percent of bentonite to the grout mix can reduce shrinkage during curing. No aggregates are added to the grout mix.


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Figure 8-35. Jet Grout Set Up.

Figure 8-36. Grout Mixing and Injection Plant.


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Figure 8-37. Drilling and Grouting Rig.

Figure 8-39. Jet Grouting Monitor.

Figure 8-38. Forming a Jet-Grouted Column.


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Figure 8-40. Rods for a Triple Fluid Grouting System. 8.6.2.6 Field Trial A field trial is considered an essential element of any jet grouting project. Its basic steps are as follows: •

Select a test site with ground conditions similar to those of the constructed project.

•

Design a field trial to model actual construction. In general, it is necessary to: (1) perform the field trial with the same equipment to be used in the actual construction; and (2) provide adequate instrumentation to study pore pressures, heave and lateral deformation of the surrounding ground.

•

Construct a number of jet-grouted test columns to determine optimum combinations of the various operational parameters and determine the column diameter and jetgrouted soil strength that can be achieved in the field.

•

Evaluate the quality of the jet grouted test section. This is usually done by: (1) excavating and exposing columns to study the wall continuity; (2) drilling inclined cores to estimate column diameter; (3) testing the strength of the cores obtained along the centerline and near the column perimeters; and (4) performing field permeability tests on completed columns or a pumping test on a completed cut-off enclosure.


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8.6.3

Cost

The costs of jet grouted walls typically range from $60 to $90/ft2 of exposed wall face. Additional costs will be incurred if ground anchors are used. The mobilization/demobilization cost for jet grouted walls is similar for slurry walls, which is typically around $50,000. The cost of cement has a significant impact on the overall cost of the jet grouted walls. Typically, cement is 40 to 45 percent of the entire jet grouted wall cost. Also, jet grouted walls require a specialty contractor with special equipment. This requirement makes jet grouted walls expensive compared to sheet pile, soldier pile, or tangent/secant pile walls. 8.6.4 Construction Inspection 8.6.4.1 General To ensure the quality of the completed wall, it is critical to exercise strict control and monitoring during the construction stage. Such control and monitoring will also provide data to verify the design criteria and to modify, if necessary, the jet grouting parameters. Depending on the scale of the project, the following controls can be exercised (Bruce, 1994): •

Alignment survey of the drilled holes;

•

Assessment of the column diameter and verticality by drilling, inclinometers, or a special verticality-measuring device introduced into the drilling rod;

•

Monitoring the vertical movements on the ground surface or adjacent structures;

•

Monitoring the horizontal displacements (e.g., by inclinometers) of the surrounding ground; and

•

Quantitative and qualitative evaluation of the ejected materials. This is particularly important for estimating the quantity of cement refuse and checking the actual composition of the treated ground.

In addition to the above controls during construction, the following options may be considered where applicable (Gallavresi, 1992): •

Coring to provide laboratory specimens to verify physical properties such as strength and permeability. This is highly recommended.


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•

Direct examination by exposing completed columns.

•

Penetration tests (such as standard penetration test or cone penetration test) and pressuremeter tests to evaluate the bearing capacities of intercolumn soils.

•

Sonic testing to evaluate the improvement of the mechanical properties, uniformity of the treatment and the possible overlapping of adjacent columns. In practice, vertical pipes are placed at preestablished spacings through which the transmitter and receiver are lowered.

8.6.4.2 Electronic Data Control An electronic data control system enables the site engineers to continuously monitor the various operational parameters both in graphical and numerical form and immediately identify and rectify any malfunctions such as drop in pressure or flow rate, or clogging of the nozzles. The site engineer can identify any change in the soil condition and can accordingly modify the jet grouting process. Furthermore, the owner can monitor the execution of the work, confirming that it has been carried out correctly and within design limits. Finally, such a system permits the use of the data stored magnetically for preparation of graphical outputs showing the variation of the measured quantities with depth. 8.6.4.3 Responsibility of Field Personnel Since there are no standard design procedures for jet-grouted works, quality control testing during the field trial and production grouting is critical to the success of the project. One of the key elements of quality control is to have experienced, qualified individuals from both the contractor and owner on site. In addition, the equipment and procedures should be suitable for the site and the requirements of the project. The first step in quality control involves the education of the field personnel in the details of the project, the jet grouting process to be used, the jet grouting parameters and the design requirements. All field staff should understand exactly what jet grouting can do, how the system works, and what data must be recorded. They should also be familiar with the project’s specifications and approved shop drawings. The next phase is the actual implementation of the jet grouting work. At this phase, critical information and observations must be made by the field personnel to ensure that the jet grouting procedures and parameters are correctly applied for the project. Also, data collected by the quality control staff can be used to identify potential problems or deficiencies in the jet-grouted wall, which may require remediation prior to any excavation. Typical items FHWA NHI-07-071 Earth Retaining Structures

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which should be observed and monitored are usually included in a check list prepared specifically for the project. Typical responsibilities of the inspector for jet grouted walls are summarized in Table 8-5. An example Guide Specification for Jet Grouting is provided in Elias et al. (2006) including additional information on QA/QC for jet grouting projects.

Table 8-5. Inspector Responsibilities for a Typical Jet-Grouted Wall Project. CONTRACTOR SET UP Review Plans and Specifications Review Contractor’s schedule Discuss anticipated ground conditions and potential problems with Contractor Review quality assurance/quality control program for the project Review Contractor’s survey for the location of the jet grout holes and confirm their consistency against Plans TEST SECTION Confirm that location of the test section is consistent with Plans Confirm consistency of Contractor’s drilling, batching, and jetting technique with Specifications and report it to the Engineer Retrieve wet-grab and core samples from the test section and confirm that satisfactory unconfined compressive strengths is achieved DRILLING Confirm that angle and depth of drilling consistent with Plans Confirm that technique used for drilling consistent with approved method after test section results Confirm drilling parameters (i.e., lift, speed, and rotation rate) consistent with what is approved during test section operations Review the soil obtained from drilled holes and report to the Engineer for his comparison against the assumed soil type for design BATCHING Confirm the preparation of grout slurry for consistency in material content Obtain sample from grout slurry and confirm its physical and chemical properties with Specifications JETTING Confirm that volume of spoil created from jetting is consistent with what is being predicted in Design Test the water-cement ratio and confirm the results with Specifications In the case spoil is considered as construction fill, confirm that it is stockpiled and left in place according to Specifications If monitoring system is required to be used, confirm that real-time electronic data is obtained continuously during jet grouting Monitor cement-grout injection pressures and rates, rotational speeds, penetration and withdrawal rates, horizontal and vertical alignments during jet grouting FHWA NHI-07-071 Earth Retaining Structures

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Write daily reports explaining the production and submit it to the Engineer Within 60 minutes of the withdrawal of jet grouting equipment, obtain wet grab soilcement samples at pre-determined depths according to the Specifications Ensure that one wet grab soil-cement cylinder from each sampling depth, selected by the Engineer, is tested to determine 7 and 28 days unconfined compressive strength. The remaining samples should be tested at 56 days As directed by the Engineer, obtain core samples to evaluate compressive strength, unit weight, and composition of the soil-cement. Ensure that coring is not performed until wet grab samples are tested Ensure that the Contractor obtains vertical alignment profiles over the length of one soilcement element per day as directed by the Engineer. The Contractor shall advise the Engineer within one hour after measuring the vertical alignment of any non-compliance with tolerance requirements POST INSTALLATION Verify pay quantities

8.7

DEEP MIXING METHOD (DMM) WALLS

8.7.1

General

The DMM wall is a system of overlapped soil-cement columns formed by mixing the in-situ soil with cement grout at depth. The overall process is termed “deep mixing method” (DMM). DMM differs from jet grouting in two ways. First, the equipment does not use high pressure grout. Second, the DMM columns are built to defined dimensions, while the jet-grouted columns are formed, under high pressure, to uncertain dimensions. DMM columns can be arranged in a variety of patterns as shown in Figure 8-41. Wall applications include containment/cutoff walls and structural retaining walls, as illustrated in Figure 8-42. The advantages of DMM include: •

can be accomplished to depths up to 100 ft for land operations, and up to approximately 230 ft for marine work;

•

the DMM wall can be built in a broad range of soils including soft to very stiff and low to highly plastic clay and silt, loose to dense sand, gravel and cobble, and soft rock;


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Figure 8-41. Plan View of Typical DMM Wall Layouts: (a) Cut-off Wall, (b) and (c) Excavation-Support Wall, and (d) Lattice Pattern for Liquefaction Control.

Figure 8-42. Various DMM Wall Applications: (a) as containment/cutoff wall and (b) as structural retaining wall.


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•

the process generates a relatively small volume of spoils because most of the in-situ soil is used as construction material and remains underground;

•

a relatively short construction time and high production rate can be anticipated (as compared to slurry walls, for example) because the method involves in situ materials and multi-axis augers;

•

the wall configuration is well defined, and the continuity between the columns is maintained; and

•

construction does not create significant noise and vibration, and the impact on adjacent facilities is minimal.

The limitations of DMM include: •

since the wall relies on use of in-situ soil as a construction material, all obstructions such as rubble, pieces of concrete, abandoned pipes and boulders must be completely removed and replaced with suitable soil, otherwise, pre-boring may have to be used to partially loosen and/or break up these obstructions prior to installation of the wall;

•

the equipment and procedures used may not be easily amenable to variation in column geometry with depth (e.g., the column cannot be wrapped around utilities), therefore, relocation of utilities may be required;

•

rigs require substantial headroom;

•

until the soil-cement is hardened, the constructed column may constitute a weak spot that may trigger movement of adjacent structures or utilities, thus making underpinning or temporary protection works necessary; and

•

although the quantity of waste is generally less than for other methods, soil mixing may produce a spoil volume equivalent to between 30 and 100 percent of the in-situ soil volume, depending on in-situ soil moisture content thus making disposal of the waste a significant cost item.


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8.7.2

Wall Construction

8.7.2.1 Basic Construction Methodology Multiple hollow-stem augers equipped with mixing paddles penetrate the ground to the required depth and back. During penetration and withdrawal, cement grout is pumped through the auger stem. The auger flights and mixing paddles mix the soil and the grout in place to form continuous overlapping soil-cement columns that constitute the wall (Figure 843). If needed, steel reinforcement is inserted in the DMM column (Figure 8-41). Additional details on deep mixing methods are provided in Elias et al. (2006). 8.7.2.2 DMM Equipment Figures 8-44 and 8-45 show typical DMM equipment. A typical DMM system consists of a mixing plant and a drilling/mixing unit. The mixing plant consists of a grout mixer, a grout agitator, automatic batching scales to control grout composition, and a computer for mixing and grout flow control. The drilling/mixing unit consists of multiple axis augers guided by a vertical steel lead on a track-mounted base machine as shown in Figure 8-43. Two to five augers can be used, with typical auger diameters varying from 22 to 42 in. The base machine, together with the lead, are supported at three points during operation for maintaining accurate vertical alignment which is critical for eliminating unmixed zones between columns and maintaining wall continuity. The typical auger has an auger head, discontinuous auger flights and mixing paddles. The auger flights are positioned so that they overlap with each other to form overlapped soilcement columns. The discontinuous auger flight is designed to provide some vertical displacement of the soil for mixing, but also to prevent transporting the soil to the surface. Thus, the auger mixes the grout with the soil at its original depth, uniformly and continuously. In recently developed equipment, jetting or spreadable mixing tools are used at the tip of the auger to enhance the grout penetration and increase the column diameter at specific depths. The interested reader is referred to Table 3 of Chapter 6 of Elias et al. (2006) for detailed operational information for installation equipment currently available.


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Figure 8-43. DMM Equipment (a) Mixing Shaft for General Use, (b) Mixing Shaft for Soil with Boulders, and (c) Mixing Shaft for Cohesive Soil (after Taki and Yang, 1991). 8.7.2.3 DMM Installation Procedures DMM walls are installed by constructing a series of sets or elements. Typically, a set consists of three overlapping columns as shown in Figure 8-46. The stepped installation procedure, which is commonly used, drills two primary column sets followed by drilling a secondary column set using two boundary columns of the primary column sets as guide holes to construct a continuous wall as shown in Figure 8-41. The redrilling in this procedure increases the uniformity of the soil-grout mix. The redrilling ratio can be adjusted according to the required level of soil-mixing. During the drilling process, grout is injected into the soil through the tip of the hollow-stem auger. A separate positive displacement pump supplies the grout to each of the injecting augers for accurate control of grout flow. The auger flights penetrate and loosen the soil, and lift it to the mixing paddles which blend the grout with the soil. As the auger continues to advance, the soil and the grout are remixed by additional paddles attached to the shaft.


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Figure 8-44. Drilling and Mixing Unit.

Figure 8-45. Mixing Plant (courtesy of MnDOT). FHWA NHI-07-071 Earth Retaining Structures

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Figure 8-46. DMM Installation Procedure (after Taki and Yang, 1991).

About 60 to 80 percent of the slurry is injected as the augers penetrate downward and the remainder is injected as they are withdrawn so that the mixing process is repeated on the way up. If reinforcing elements are to be used, they are inserted into the soil-cement column immediately after the auger is withdrawn. 8.7.2.4 DMM Operational Parameters There are three principal DMM operational parameters as follows: 1.

Drilling Speed: The drilling speed is governed by the properties of the soil to be treated, and the soil mixing effort required to obtain the design soil-cement properties.

2.

Mix Design: The DMM wall usually comprises three basic materials: the soil, the grout and the reinforcement, if any. The most suitable soil and grout mixes are similar to those discussed under jet grouting. Portland cement is commonly used. A small amount of bentonite is sometimes added to increase the workability of the soilmixing work. Cement-based additives such as silicate, slag and gypsum are used for gaining strength in saline or organic soils, or for stabilizing contaminated soils.


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The mix design is governed by the required engineering properties of the DMM wall such as strength and permeability. The type of the soil essentially determines the extent of improvement in engineering properties. Laboratory strength and permeability tests are normally performed to identify the cement proportion that could provide the required properties. The final selection of a mix design is influenced by the selection of equipment and installation procedures used, efficiency and economy. In the field, an automated batching system measures the water, cement and other additives by weight to produce a more reliable grout than that produced by a volumetric batch system. The desired weight of each grout component can be entered in the computer at a control panel, and changes to the mix design can be made by simply adjusting the component weights at the control panel. 3.

Grout flow rates: The grout flow rate is usually adjusted constantly to accommodate varying drill speeds in different soil strata, so that the design volume of grout per unit volume of in situ soil is maintained. The grout flow is electronically controlled.

8.7.3

Cost

Given the large cranes required to support the multiple mixing augers, coupled with the specially designed on site batcher, the current minimum cost for mobilization/demobilization for DMM is in the vicinity of $100,000. In addition, the cost per unit volume of mixed soil may vary depending on the normal project variables, e.g. labor, size of the project, depth and type of in-situ soil being treated. A typical cost for DMM walls is $40 to $55/ft2 of exposed wall face. 8.7.4 Construction Inspection During construction of the wall, field sampling of the soil-cement is performed regularly and unconfined compressive strength and total unit weight is performed to confirm the design parameters. Wet soil-cement samples should be obtained routinely and cured in the laboratory for testing. After construction of a soil-cement column, core samples should be obtained from the exposed wall according to a testing schedule to be determined based on the site conditions observed during excavation. Usually, the core samples are tested after 7, 28 and 56 days to establish the strength increase with time. Based on the results of unconfined compression testing for several projects as part of the Central Artery Project in Boston, average strengths from wet soil-cement samples were approximately 50 percent greater than for core samples from hardened columns. This was attributed to wet soil-cement sampling


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methods in that less clay material is retrieved, resulting in a higher percentage of cement (O’Rourke and McGinn, 2004). It is recommended that for the finished DMM wall, direct shear tests and/or triaxial compression tests be conducted for strength assessment, in addition to the unconfined compression tests. For quality control during construction, however, the unconfined compression test is adequate, and the results may be used as standard values. Monitoring of equipment and operational procedures is a key element of the QC/QA program. The number of revolutions of the mixing paddles per unit volume of in situ soil is usually used in monitoring the soil mixing effort. The drilling time spent in installation of each DMM panel is recorded and used as part of the QC/QA program. The grout injection rate is also monitored and adjusted constantly during production to accommodate varying drilling speeds in different soil strata, so that the design volume of grout per unit volume of column is maintained. Vertical tolerance may also be monitored (via inclinometer) when the augers are at maximum depth. The inspection responsibilities for DMM walls are summarized in Table 8-6. Table 8-6. Inspector Responsibilities for Typical DMM Walls CONTRACTOR SET UP Review Plans and Specifications Review Contractor’s schedule Discuss anticipated ground conditions and potential problems with Contractor Review test results of in-situ soils for consistency with design assumptions Review Contractor’s surface water control methods and confirm its consistency with Specifications Review Contractor’s survey results to confirm column locations are consistent with Plans TEST COLUMN Confirm that location of the test column is consistent with Plans Confirm that test column is constructed full scale as specified in Specifications Review the test results for consistency with Specifications and report to the Engineer DRILLING Confirm that location of drilling consistent with Plans Confirm that angle and depth of drilling consistent with Plans Confirm that drilling speed consistent with Specifications Review the soil from drilled holes and report to the Engineer for his comparison against the assumed soil type for design If needed, confirm that redrilling is performed according to Specifications


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MIXING Confirm freshness of stabilizers by inspecting the expiration dates Confirm consistency of jetting and soil mixing parameters (i.e., lift, speed, rotation and injection rates) with Specifications Confirm that grout flow rates adjusted when needed so that design volume of grout per unit volume of in situ soil is maintained Confirm that during the course of all soil-cement mixing, no water, debris or spoil material is dumped or otherwise allowed to enter the soil-cement element Obtain in-situ soil-cement samples in accordance with the locations and frequencies required in Specifications Confirm that in-situ strength tests are performed as outlined in Specifications. An example of in-situ testing program is given in Table 8-7 Confirm that any soil-cement column that exhibits partial or total instability is backfilled with weak cement grout and remixed full depth Confirm that Contractor is taking all precautions to keep the site cleared of all debris and water at completion of soil-cement mixing operations according to Specifications Prepare a daily report summarizing all field activities and submit it to the Engineer Ensure that Contractor submits to the Engineer any deviations of the center coordinates and vertical alignment of soil-mix columns POST INSTALLATION Verify pay quantities

Table 8-7. Example of QA/QC Testing Program for DMM Walls Subject

Test Items

Instrument

Frequency

Quantity of cement

Total weight

Delivery record

Daily

Cement grout

Specific gravity

Mud balance

Each batch

Quantity of grout

Volume

Flow meter

Each column

Drilling speed

Record

Each column

Rotation speed

Record

Each column

Wet sampling

Sampling tool

1 to 2 times/day

Column length

Drill stem

Each column

Tool Diameter

Daily

QC Mix condition

Length Diameter QA

Continuity

Core Drilling

Visual

Random Sampling

Uniformity

Core Drilling

Visual

Random Sampling

Strength

Core Drilling

Unconfined Compression Test

Random Sampling


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8.8

GROUND ANCHORS

8.8.1

General

Ground anchors, sometimes referred to as tiebacks, are structural tension elements that receive their support in soil or rock. The basic anchor components include: (1) anchorage; (2) unbonded length; and (3) bond length. These and other components of a ground anchor are shown schematically in Figure 8-47. The anchorage is the combined system of anchor head, bearing plate, and trumpet that is capable of transmitting the prestressing force from the prestressing steel (bar or strand) to the ground surface or the supported structure. The unbonded length is that portion of the prestressing steel that is free to elongate elastically and transfer the resisting force from the anchor bond length to the structure. A bondbreaker is a smooth plastic sleeve that is placed over the tendon in the unbonded length to prevent the prestressing steel from bonding to the surrounding grout. It enables the prestressing steel in the unbonded length to elongate without obstruction during testing and stressing and leaves the prestressing steel unbonded after lock-off. The tendon bond length is that length of the prestressing steel that is bonded to the grout and is capable of transmitting the applied tensile load into the ground. The anchor bond length should be located behind the critical failure surface.

Un bo nd ed Trumpet Le ng th

Anchor Head

Sheath

Bearing Plate

An ch or Te Bo nd nd on Le Bo ng nd th Le ng th

Wall Unbonded Tendon Anchor Grout Bonded Tendon

An Di cho am r ete r

Figure 8-47. Components of a Ground Anchor (after Sabatini et al., 1999). FHWA NHI-07-071 Earth Retaining Structures

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The sheath is a smooth or corrugated pipe or tube that protects the prestressing steel in the unbonded length from corrosion. Centralizers position the tendon in the drill hole such that the specified minimum grout cover is achieved around the tendon. For multiple element tendons, spacers are used to separate the strands or bars of the tendons so that each element is adequately bonded to the anchor grout. The grout is a Portland cement based mixture that provides load transfer from the tendon to the ground and provides corrosion protection for the tendon. Anchors may be used for either temporary or permanent support applications. The design and construction of both are similar, except that permanent anchors generally are designed for lower stresses in the tendon, and are provided with added protection against corrosion. The advantages of ground anchors are listed below: • • • • • • •

eliminates internal bracing that may impair construction operations; generally easy installation; unlike internally braced excavations, an unobstructed working space can be achieved on the excavation side of the wall; relatively large horizontal earth pressures can be resisted by an anchored wall; reliable load capacity, since each production anchor is load tested; wall system is suitable for temporary applications; and more uniform load distribution.

The limitations of ground anchors are listed below: • • • • • • 8.8.2

may not be suitable for organic and some cohesive soils; requires penetration through the retaining wall, creating potential path of groundwater seepage; the vertical load component of the anchor is an added vertical load on the wall; potential interference with utilities, basements and other underground facilities; corrosion potential of the steel elements; and construction of wall system requires skilled labor and specialized equipment. Wall Construction

The method of anchor construction is greatly influenced by the type of ground at the site. Ground anchors can be used essentially in all cohesionless soil deposits. Good performance has also been observed for anchors in stiff to hard cohesive soils with N-value greater than 9 (Weatherby, 1982). Soil deposits not generally suitable for the anchor bond length (Cheney, 1988) include: FHWA NHI-07-071 Earth Retaining Structures

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• • •

organic soils; cohesive soils with an average liquidity index greater than 0.2; and cohesive soils with an average liquid limit greater than 50.

In addition to these criteria, caution should be exercised with soils that have a plasticity index greater than 20 percent as they may exhibit excessive creep. There are three main ground anchor types that are currently used in U.S. practice: (1) straight shaft gravity-grouted ground anchors (Type A); (2) straight shaft pressure-grouted ground anchors (Type B); and (3) post-grouted ground anchors (Type C). Although not commonly used today in U.S. practice, another type of anchor is the underreamed anchor (Type D). These ground anchor types are illustrated schematically in Figure 8-48. Straight shaft gravity-grouted ground anchors (Type A) are typically installed in rock and very stiff to hard cohesive soil deposits using either rotary drilling or hollow-stem auger methods. Tremie (gravity displacement) methods are used to grout the anchor in a straight shaft borehole. The borehole may be cased or uncased depending on the stability of the borehole. Anchor resistance to pullout of the grouted anchor depends on the shear resistance that is mobilized at the grout/ground interface. Straight shaft pressure-grouted ground anchors (Type B) are most suitable for coarse granular soils and weak fissured rock. This anchor type is also used in fine-grained cohesionless soils. With this type of anchor, grout is injected into the bond zone under pressures greater than 50 psi. The borehole is typically drilled using a hollow stem auger or using rotary techniques with drill casings. As the auger or casing is withdrawn, the grout is injected into the hole under pressure until the entire anchor bond length is grouted. This grouting procedure increases resistance to pullout relative to tremie grouting methods by: (1) increasing the normal stress (i.e., confining pressure) on the grout bulb resulting from compaction of the surrounding material locally around the grout bulb; and (2) increasing the effective diameter of the grout bulb. Post-grouted ground anchors (Type C) use delayed multiple grout injections to enlarge the grout body of straight-shafted gravity grouted ground anchors. Each injection is separated by one or two days. Postgrouting is accomplished through a sealed grout tube installed with the tendon. The tube is equipped with check valves in the bond zone. The check valves allow additional grout to be injected under high pressure into the initial grout which has set. The high-pressure grout fractures the initial grout and wedges it outward into the soil enlarging the grout body. Two fundamental types of post-grouted anchors are used. One system uses a packer to isolate each valve. The other system pumps the grout down the post-grout tube without controlling which valves are opened. FHWA NHI-07-071 Earth Retaining Structures

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Figure 8-48. Main Types of Grouted Ground Anchors (after Littlejohn, 1990). Underreamed anchors (Type D) consist of tremie-grouted boreholes that include a series of enlargement bells or underreams. This type of anchor may be used in firm to hard cohesive deposits. In addition to resistance through side shear, as is the principal load transfer mechanism for other anchors, resistance may also be mobilized through end bearing. Care must be taken to form and clean the underreams. 8.8.2.1 Materials Cement Grout The cement grout consists of sand, cement, and water. Under normal conditions, most specialty contractors prefer not to use chemical additives in the grout (Munfakh, et al., 1987). Portland cement with low sulfate content is normally used. The grout mix should attain a minimum cube strength (AASHTO T 106) of 3,500 psi at 7 days (Cheney, 1988). The water/cement ratio of tendon bonding grouts is usually in the range of 0.35 to 0.60. FHWA NHI-07-071 Earth Retaining Structures

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Steel Bar and Strand Tendons Both bar and strand tendons are commonly used for soil and rock anchors for highway applications in the U.S. Material specifications for bar and strand tendons are codified in American Society for Testing and Materials (ASTM) A722 and ASTM A416, respectively. Indented strand is codified in ASTM A886. Bar tendons are commonly available in 1 in., 1.25 in., 1.375 in., 1.75 in., and 2.5 in. diameters in uncoupled lengths up to approximately 60 ft. For lengths greater than 60 ft and where space constraints limit bar tendon lengths, couplers may be used to extend the tendon length. As compared to strand tendons, bars are easier to stress and their load can be adjusted after lock-off. Strand tendons comprise multiple seven-wire strands. The common strand in U.S. practice is 0.6 in. in diameter. Anchors using multiple strands have no practical load or anchor length limitations. Tendon steels have sufficiently low relaxation properties to minimize long-term anchor load losses. Couplers are available for individual seven-wire strands but are rarely used since strand tendons can be manufactured in any length. Strand couplers are not recommended for routine anchor projects as the diameter of the coupler is much larger than the strand diameter, but strand couplers may be used to repair damaged tendons. Where couplers are used, corrosion protection of the tendon at the location of the coupler must be verified. Spacers and Centralizers Spacer/centralizer units are placed at regular intervals (e.g., typically 10 ft) along the anchor bond zone and within 3 ft of either end of the anchor tendon. For strand tendons, spacers usually provide a minimum interstrand spacing of 0.25 to 0.50 in. and a minimum outer grout cover of 0.5 in. Both spacers and centralizers should be made of non-corrosive materials and be designed to permit free flow of grout. Figures 8-49 and 8-50 show a cut away section of a bar and a strand tendon, respectively. Epoxy-Coated Bar and Epoxy-Coated Strand Epoxy-coated bar (AASHTO M284) and epoxy-coated strand (supplement to ASTM A882), are becoming more widely used. The epoxy coating provides an additional layer of corrosion protection in the unbonded and bond length as compared to bare prestressing steel. Additional details on epoxy-coated bar and epoxy-coated strand can be found in PTI (2004).


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SMOOTH PLASTIC SHEATHING

CORRUGATED PLASTIC SHEATHING

CEMENT GROUT SPACER

BAR

Figure 8-49. Cut away view of bar tendon (after Sabatini et al., 1999).

CENTRALIZER STRAND SHEATH

GROUT TUBE

SPACER

Figure 8-50. Cut away view of strand tendon (after Sabatini et al., 1999). FHWA NHI-07-071 Earth Retaining Structures

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8.8.2.2 Construction Sequence Ground anchors can be used with both flexible and stiff nongravity cantilevered walls, however; soldier pile and lagging walls are the most commonly used type of anchored wall system in U.S. The construction sequence for a permanent soldier pile and lagging wall is illustrated in Figure 8-51 and the components of ground anchor construction is described below. Drilling A variety of drilling equipment is available for installation of ground anchors, including hollow-stem augers, rotary drills with casing, and down-the-hole hammers with casing (Figure 8-52). Irrespective of the drilling method used, care should be taken not to use high pressures with any flushing medium to minimize the risk of hydrofracturing the surrounding ground, particularly in built-up areas. For this reason, an open return within the borehole may be beneficial to limit the pressure and give the driller an indication of the type of ground through soil cuttings so that the fluid pressure may be adjusted accordingly. After drilling is completed and the hole is thoroughly flushed out, the drilled hole should be probed to verify that no collapse of material has occurred before installation of the prestressing element. This installation and the subsequent grouting should be carried out on the same day as drilling to avoid potential ground deterioration. Tendon Installation Steel tendons should be stored indoors, if possible. If left outdoors, however, they should be stacked off the ground under a waterproof cover that allows air circulation. Neither bare nor coated tendons should be dragged across abrasive surfaces. The tendon is lowered in the borehole manually or using mechanical equipment (Figure 853). Immediately before its installation, the tendon should be carefully inspected for damage and corrosion. Grouting The grouting operation involves injecting cement grout at the lowest point of the borehole so the hole will fill evenly without air voids. A grout pipe is often tied to the tendon before inserting it in the borehole.


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Figure 8-51. Construction Sequence for Permanent Soldier Pile and Lagging Wall (after Sabatini et al., 1999). The cement grout is prepared by batching the dry materials by mass and mixing them mechanically, with water added, for at least two minutes in order to obtain a homogeneous mix. High-speed colloidal mixers (1000 rpm minimum) or paddle mixers (150 rpm minimum) are used for grout preparation. The colloidal mixers are preferred when grouting water-bearing ground since grout dilution is minimal. After mixing, the grout is kept in continuous motion until it is pumped to its final position. Grout should not be used after a period equivalent to its initial setting time. Prior to grouting, all air in the pump and the line should be removed. An airtight system should be maintained at all times during grouting. FHWA NHI-07-071 Earth Retaining Structures

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(a)

(b)

Figure 8-52. Drilling Equipment for Installation of Ground Anchors: (a) Hollow Stem Auger and (b) Down-the-Hole Hammer.

Figure 8-53. Installation of Anchor Tendon.


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Prestressing The stressing operation involves fitting the jack assembly on the anchor head, loading and/or unloading of the anchors, locking off the load by anchor nuts or wedges, then removal of the assembly from the anchor head. The equipment used for stressing operations need to be calibrated. Stressing should not begin until the grout strength has reached a crushing strength of approximately 3,500 to 4,500 psi. Cheney (1988) recommends a waiting period of 7 days before stressing can take place. 8.8.3

Load Testing of Ground Anchors

The load testing of ground anchors should be considered an integral part of the design. The typical types of load tests include: • •

• •

Preproduction Tests: These tests are used to verify the available resistance in the anchor bond length and establish anchor loads for designs. Performance Tests: These tests are conducted at the beginning of construction and periodically during construction to verify short and long-term performance of the anchor under the design load. Proof Tests: These tests are used to determine the behavior of each production anchor after installation. Lift-Off Tests: These tests are used to confirm the load in the tendon after completion of installation and lock-off of the applied load.

Figure 8-54 shows a typical set-up for strand and bar anchor testing. 8.8.3.1 Preproduction Tests Preproduction tests may be specified when unusual conditions are identified during the design stage. Situations prompting such tests include: (1) soil deposits in which no previous experience exists; (2) very long anchors; (3) difficult drilling conditions; or (4) creep susceptible soils in the bond length. The preproduction tests can be performed under a separate test contract awarded during design stage, or at the beginning of the construction contract. The anchors used preproduction testing are usually not incorporated in the final structure because of potential damage that may be induced by the high test loads. Also, the bond length preproduction test anchors is typically shorter than that of the production anchors.

the for the for

Cheney (1988) provides recommended testing procedures for preproduction tests. FHWA NHI-07-071 Earth Retaining Structures

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STRESSING ANCHORAGE DIAL GUAGE

JACK FIXED BASE

(a)

(b) Figure 8-54. Typical Equipment For Load Testing Of (A) Strand Ground Anchor And (B) Bar Ground Anchor. FHWA NHI-07-071 Earth Retaining Structures

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8.8.3.2 Performance Tests Performance tests provide the necessary information to verify that the production anchors will be able to hold the design load without excessive movement or creep. Cheney (1988) recommends performance tests be conducted on the first two anchors installed, then on 2 percent of the remaining anchors in rock or cohesionless soils and 10 percent in cohesive soils. The performance test is a cyclic test made by incrementally loading and unloading the anchor until the maximum test load is reached. The typical loading and unloading sequence (PTI, 2004) is as follows: AL, 0.25P, AL, 0.25P, 0.50P, AL, 0.25P, 0.50P, 0.75P, AL, 0.25P, 0.50P, 0.75P, 1.00P, AL, 0.25 P, 0.50 P, 0.75 P, 1.00P, 1.20 P, AL, 0.25 P, 0.50 P, 0.75P, 1.00P, 1.20P, 1.33P, 1.00 P, LL where AL is the alignment load, P is the design load (i.e., unfactored anchor load), and LL is the lock-off load. Each load or unload increment is held constant just long enough to obtain the movement reading but no longer than one minute. The maximum load should generally be held for one hour in cohesive soils to determine the long-term creep potential. Coarse granular soils and rock do not generally exhibit creep; creep tests in such deposits may be terminated if negligible creep, i.e., less than 0.04 in. movement, is observed between the 1 minute and 10 minute readings of the test. The deflection measurements at the maximum load level (i.e., 1.33 P) are taken at the following intervals: 1, 2, 3, 4, 5, 6, 10, 20, 30, 40, 50 and 60 minutes. At each load increment, the total movement of the pulling head should be recorded to the nearest 0.001 in. 8.8.3.3 Proof Tests Every production anchor (that is not performance tested) is proof tested. The proof test is a single cycle test in which the test load is applied in increments until the maximum load. For granular soils and rock, the maximum test load should be 1.33 times the anchor design (i.e., unfactored) load; for cohesive soils, the maximum test load should be 1.50 times the anchor design (i.e., unfactored) load. The maximum load is held constant for at least one minute or until the measured deflection is negligible. The typical loading and unloading sequence is as follows: AL, 0.25 P, 0.50 P, 0.75 P, 1.00 P, 1.20 P, 1.33 P, AL (optional), LL FHWA NHI-07-071 Earth Retaining Structures

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The maximum test load (i.e., 1.33 P) is maintained constant for 10 minutes, and total movement readings are recorded at 1, 2, 3, 4, 5, 6 and 10 minutes. If the total creep movement between 1 and 10 minutes exceeds 0.04 in., the maximum test load should be maintained for an additional 50 minutes, and the movement readings are recorded at 20, 30, 40, 50 and 60 minutes. 8.8.3.4 Supplementary Extended Creep Tests An extended creep test is a long duration test (e.g., approximately 8 hours) that is used to evaluate creep deformations of anchors. These tests are required for anchors installed in cohesive soil having a plasticity index (PI) greater than 20 or liquid limit (LL) greater than 50. For these ground conditions, a minimum of two ground anchors should be subjected to extended creep testing. Where performance or proof tests require extended load holds, extended creep tests should be performed on several production anchors. The test arrangement for an extended creep test is similar to that used for performance or proof tests. The increments of load for an extended creep test are the same as those for a performance test. At each load cycle, the load is held for a specific period of time and the movement is recorded. During this observation period, the load should be held constant. The load is assumed to remain reasonably constant if the deviation from the test pressure does not exceed 50 psi. The loading schedule and observation periods for each load cycle in an extended creep test for a permanent anchor are provided in Table 8-8.

Table 8-8. Load Schedule And Observation Periods For Extended Creep Test For Permanent Anchor. Loading Cycle 1 2 3 4 5 6

Maximum Cycle Load

Total Observation Period (min)

0.25P 0.50P 0.75P 1.00P 1.20P 1.33P

10 30 30 45 60 300


Movements measured at following times (min)

1, 2, 3, 4, 5, 6, 10 1, 2, 3, 4, 5, 6, 10, 15, 20, 25, 30 1, 2, 3, 4, 5, 6, 10, 15, 20, 25, 30 1, 2, 3, 4, 5, 6, 10, 15, 20, 25, 30, 45 1, 2, 3, 4, 5, 6, 10, 15, 20, 25, 30, 45, 60 1, 2, 3, 4, 5, 6, 10, 15, 20, 25, 30, 45, 60, 300

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If the creep rate exceeds 0.08 in. per logarithmic cycle, the observation period may be extended in an attempt to determine if the creep rate will diminish to the 0.08 in. per logarithmic cycle of time. Extended creep tests are not normally performed on rock anchors since they do not exhibit time dependent movements. However, anchors installed in very decomposed or argillaceous rocks may exhibit significant creep behavior. 8.8.3.5 Acceptance Criteria for Load Testing Three criteria are commonly used to determine the acceptability of anchor tests (PTI, 1996). These are as follows: • • •

Creep; Movement; and Lock-off Load.

Creep The creep amount shall not exceed 0.04 in. at the maximum test load during the period of 1 to 10 minutes. If this value is exceeded, then the total creep movement within the period of 6 to 60 minutes shall not exceed 0.08 in. The creep behavior of epoxy coated strand itself is significant and the measured anchor creep movements must be adjusted to reflect the behavior of the material. At a maximum test load of 80 percent of the guaranteed ultimate tensile strength of the tendon, creep movements of epoxy coated strand can be estimated to be 0.017 percent of the apparent free stressing length during the 6 to 60 minute log cycle, but may be significantly higher than this value (PTI, 2004). PTI (2004) provides an approach to evaluate creep test data and acceptability based on creep test results for epoxy-coated strand. For epoxy coated bars, these considerations do not apply. Movement a) Residual Movement Residual movement is defined as the non-elastic (non-recoverable) movement of an anchor obtained during load testing. The residual movement (δr) is the difference between the total movement (δt) measured during the load testing and the elastic (recoverable) movement (δe), as illustrated in Figure 8-55. FHWA NHI-07-071 Earth Retaining Structures

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Figure 8-55. Typical Plots of Tendon Movement for (a) Performance Test, and (b) Proof Test (after PTI, 1996).


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There is no absolute criterion for the amount of residual movement which is acceptable. Measurement of this residual movement is, however, essential to determine the elastic movement. From the computed elastic movement, the apparent free length of the anchor can be calculated for which the acceptance criteria are described below. b) Minimum Apparent Free Length The minimum apparent free length is calculated to verify that the anchor load is being transferred beyond any potential failure or slip plane in accordance with the overall stability requirements of the anchor-structure system. The minimum apparent free length at the maximum test load, as calculated on the basis of elastic movement, should be not less than 80 percent of the designed free tendon length plus the jack length. If this criterion is not met, the anchor should be reloaded up to two times more from the alignment load to the maximum test load, and the calculation repeated on these cycles. If the criterion is still not met, then (i) the cause of this inefficiency in load transfer should be investigated; and (ii) the anchor may be rejected or assigned a lower resistance (or capacity). A limit higher than 80 percent of the designed free length should be set in cases where future additional movements occurring as a result of redistribution of the free length friction would cause unacceptable structural movement. c)

Maximum Apparent Free Length

The acceptance criterion based on maximum apparent free length was used in the past when load transfer along the bond length was assumed to propagate at a uniform rate as the applied load was increased. For that assumption, the maximum value of apparent free length was restricted to elastic movements of 100 percent of the free length plus 50 percent of the bond length plus the jack length. However, the concept of uniform distribution of bond is not valid for soil anchors and only approximates the behavior of most rock anchors. The primary use of this criterion is as an alternate acceptance criterion for proof tests in sound rock where creep tests are waived. Anchors that do not pass this preliminary criterion are subsequently creep tested to determine acceptability before a decision is made to reject the anchor. d) Acceptability Based on Total Movements The criteria for the minimum and maximum apparent free length, as described above, are not strictly relevant if only total movement data are available. However, it is conventional to apply these criteria also to total movement data when, from past experience or previous tests in the same conditions, the magnitude of the residual movements is well known, and elastic movements can, therefore, be estimated. In such cases, the criteria listed above should be FHWA NHI-07-071 Earth Retaining Structures

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applied. Otherwise, only the criterion for the minimum free length should be used as a basis for acceptance, even though it is based on total movements. Lock-Off Load After the factored anchor load of a production anchor has been verified by testing, the anchor load is immediately transferred to the structure. The magnitude of this initial transfer load must be determined based on structural design assumptions and the mechanical losses associated with the tendon type selected. The mechanical losses associated with transferring load to various anchorage systems and long-term relaxation or creep should also be considered in determining the final transfer load. Seating losses may vary from 0.08 in. for bars to 0.25 in. for strand; however, these values will be dependent on specific details of the anchorage system. The magnitude of the transfer load is generally specified, and should not exceed 70 percent of the guaranteed ultimate tensile strength of the tendon. In practice, transfer loads of 80 percent of the design load are commonly specified. Higher transfer loads are sometimes used when it is required to minimize long-term wall deflection. In selecting the transfer load, an evaluation should be made to verify that the transfer load, particularly at the top anchor level, will not exceed the ultimate passive resistance of the soil behind the wall since this could cause ground displacement and heave which might damage existing facilities. Lift-off tests are performed either during construction to check the magnitude of seating and transfer losses or after construction to determine if long-term load losses are taking place. The test is performed by applying load gradually until the tendon begins to elongate. When a sudden deflection is observed on the dial gauge, the jack extension should be immediately terminated, and the load required for lift-off recorded. This load should be approximately equal to the design load plus an allowance for long-term losses. If the lift-off load varies more than 5 percent from this value, the tendon load is adjusted and the lift-off test is repeated. When the load in a strand tendon is more than 5 percent above the desired lock-off load, and where no shims have been prepositioned under the wedge plate for later extraction, then it is preferable to accept this load and so avoid the danger of having wedge marks below the wedge plate as a result of strand/wedge regripping. Failure during Testing If an anchor does not reach the maximum test load as a consequence of bond failure, subsequent actions depend on whether the anchor can be postgrouted or not. Regroutable anchors should be postgrouted and then subjected to all the original acceptance criteria. FHWA NHI-07-071 Earth Retaining Structures

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Anchors without a postgrouting system should be rejected (and replaced) or locked off at not more than 50 percent of the maximum load attained. In this event, no further acceptance criteria are applied. If an anchor fails the creep test at the maximum test load, then the anchor should be postgrouted and subjected to an enhanced creep criterion, assuming the other acceptance criteria are met. This enhanced criterion requires a creep movement of not more than 0.04 in. between 1 and 60 minutes at the maximum test load. Anchors, which cannot be postgrouted may be rejected or should be locked off at 50 percent of the maximum test load. In this event, no further acceptance criteria are applied. 8.8.4

Cost

The cost of ground anchors ranges from approximately $45 to $60 per ft of anchor. Higher costs are associated with longer anchors, rock drilling, and the need for larger diameter holes. Typically, strand tendons are less expensive than steel bars, however strand tendons are not suitable for ground anchors less than 40 ft long. Ground anchors require typically require 0.6 H (where H is the height of the wall) distance behind the wall face for installation. As with most wall systems, unit costs will increase where access is limited and where right-of-way costs are significant. 8.8.5 Construction Inspection Ground anchors are contracted using either method approach, performance approach, or design/build approach. Method approach typically includes the development of a detailed set of plans and material and construction specifications for the bidding documents. However, the selection and installation of the anchors should be the responsibility of the contractor. The contract documents should only establish minimum dimensions for drill hole diameter, unbonded length, and bond length. The contractor should select the necessary anchor installation dimensions and techniques to successfully pass the acceptance tests. In no case should the owner specify the installation details for the anchors. For anchored walls, the spacing, inclination, method of drilling, drill hole size, and length of each anchor must be consistent with the design assumptions. Inspection personnel must pay close attention to these issues because the ground anchors are the principal load-carrying components of the wall system. Significant deviations in any of these items should be reported to the design engineer. The inspection responsibilities of ground anchors are summarized in Table 8-9. FHWA NHI-07-071 Earth Retaining Structures

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Table 8-9. Inspector Responsibilities for Ground Anchors and Anchored Walls CONTRACTOR SET UP Review project Plans and Specifications Review Contractor’s schedule and work plan describing ground anchor number, anchor design load, type and size of tendon, minimum anchor length, minimum bond length, minimum tendon length, and minimum unbonded length for consistency with Specifications Review the Plans submitted by Contractor for the ground anchor tendon and corrosion protection system and check their consistency with Specifications Review Contractor’s grout mix design and procedures for placement of the grout for consistency with Specifications Discuss anticipated ground conditions and potential problems with Contractor Check overall condition of Contractor’s equipment Confirm that the layouts and dimensions of the ground anchors consistent with Plans Review Contractor’s stockpile area and confirm its consistency with Plans VERTICAL WALL ELEMENT Follow inspector responsibilities for the selected vertical wall element ANCHOR HOLE DRILLING Review Contractor’s survey results and confirm location of anchor holes based on Plans Confirm consistency of length, orientation, and diameter of each hole with Plans and Specifications Confirm consistency of drilling method used with Specifications Identify soil and rock types from samples obtained from each hole and report to the Engineer If used, confirm consistency of drilling muds and/or foams used during drilling holes with Specifications Confirm placement of centralizers according to Plans TENDON INSERTION Inspect cleanliness of drilled holes Inspect all metallic units prior to installation to evaluate corrosion protection Inspect all tendons for damage prior to insertion Confirm the consistency of the dimensions of each tendon with Specifications prior to insertion Confirm consistency of tendon insertion with Specifications and be aware for any indications of drill hole collapse Confirm that tendon can be completely inserted without difficulty ANCHOR GROUTING Confirm that grouting equipment produces a grout, free of lumps and undispersed cement Confirm consistency of grout injection pressures with Specifications and record the pressures Confirm that grout is injected at the lowest point in the drill hole. Record grout volume placed in each hole Confirm that withdrawal rate is less than grout placement rate FHWA NHI-07-071 Earth Retaining Structures

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Confirm that after grouting, tendon is not loaded for a minimum of 5 days ANCHORAGE INSTALLATION Confirm consistency of alignment of anchorage and tendon with Specifications and Plans Confirm that installation of trumpet does not result in damage to sheath or tendon Confirm that bearing plate is installed without bending or kinking the prestressing steel elements If used, confirm that wedge holes and wedges are clean and free of rust LOAD TESTING Inspect calibration certifications for all jacks, gauges, and load cells according to the requirements specified in Specifications Inspect all deformation gauges Confirm that the load test is performed in accordance with Specifications and test results are consistent with load test acceptance criteria as outlined in Specifications If the ground anchor fails, report to the Engineer and do not allow any retesting until the Contractor modifies the installation procedures Upon successful completion of load testing, confirm that the ground anchor load is reduced to the required lock-off load After load testing and load lock-off, confirm that trumpet is completely filled with corrosion inhibiting grout or grease as specified in the Specifications POST INSTALLATION Verify pay quantities

8.9

DESIGN OF FLEXIBLE NONGRAVITY CANTILEVERED WALLS

8.9.1

Sheet Pile and Soldier Pile and Lagging Walls

Cantilever sheet pile walls derive their support from the structural stiffness of the wall sheeting and the passive soil resistance developed below the exposed base of the wall. This type of wall is generally limited to heights up to about 15 ft and is generally suitable only in granular soils or stiff clays. For cantilever walls greater than about 15 ft in height the maximum bending moment in the wall may exceed the structural resistance of commonly available sheeting, or ground displacement behind the wall may become excessive. Typically, the required penetration of the sheeting will be 1.0 to 1.5 times the exposed height of the wall, depending on soil and groundwater conditions, ground slope, and surcharge loads. Soldier pile and lagging walls derive their lateral resistance and bending moment capacity through the embedment of vertical elements (soldier piles). The soil behind the wall is retained by lagging. The spacing of the lagging varies from 6.5 to 10 ft with a common spacing of 8 ft. Soldier pile and lagging walls are considered flexible walls with discrete vertical wall elements. A portion of the load from the retained soil is transferred to the FHWA NHI-07-071 Earth Retaining Structures

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vertical elements through arching (i.e., load is redistributed away from the lagging to the much stiffer soldier piles). The purpose of the lagging is to prevent the retained soil from eroding, which would destroy the arching effect. The lateral displacement of a cantilever wall penetrating a sand layer is shown in Figure 856. The wall rotates about point O, resulting in the reversal of active and passive earth pressures in the three different zones shown. The corresponding net earth pressure distribution on the wall is shown in Figure 8-56(b) and 8-56(c). While the net earth pressure diagram is instructive in understanding the link between deformations and earth pressures, the net pressure diagram cannot be used directly in LRFD calculations because active earth pressures (and resultant forces) are treated using factored loads whereas passive earth pressures are considered to be a resistance. For example, active earth pressures at the Strength I limit state may use a load factor of 1.5 whereas the passive resistance may use a resistance factor of 0.75, thus the active side and passive side of the wall must be “factored” individually. 8.9.2

Design Steps for Flexible Nongravity Cantilevered Walls

Table 8-10 summarizes the major design steps for a flexible nongravity cantilevered wall. 8.9.2.1 Steps 1, 2, and 3 Herein, it is assumed that Step 1 has been completed and a nongravity cantilevered wall has been deemed appropriate for the project and Steps 2 and 3 have been completed to establish soil and/or rock parameters for design. 8.9.2.2 Step 4: Select Earth Pressure Diagram The earth pressure distribution that develops on nongravity cantilevered walls (e.g., sheetpile or soldier pile and lagging walls which are not anchored) can be expected to undergo lateral deformations sufficiently large to induce active earth pressures for the entire wall height. For design of these systems, theoretical active earth pressure diagrams using Coulomb analysis methods to evaluate the active earth pressure coefficient are used. AASHTO Figures 3.11.5.6-1 through 3.11.5.6-3 may be used to calculate the unfactored lateral earth pressure distributions for permanent nongravity cantilevered walls. AASHTO Figures 3.11.5.6-4 through 3.11.5.6-7 may be used to calculate the unfactored lateral earth pressure distributions for temporary walls supporting or supported on cohesive soils with the following restrictions: FHWA NHI-07-071 Earth Retaining Structures

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Active pressure

Zone A

Active pressure Zone B

Passive pressure Active o pressure

Passive Zone C pressure (b)

(a)

(c)

Figure 8-56. Typical Deformation Conditions and Pressure Distribution for Cantilever Walls: (a) Yielding Pattern of Cantilever Wall Penetrating a Sand Layer; (b) Net Actual Earth Pressure Distribution; (c) Simplified Net Earth Pressure Distribution (after Das, 1990). Table 8-10. Design Steps for Flexible Nongravity Cantilevered Walls. Step 1.

Establish project requirements including all geometry, external loading conditions (temporary and/or permanent, seismic, etc.), performance criteria, and construction constraints.

Step 2.

Evaluate site subsurface conditions and profile, water profile, and relevant properties of in situ soil and rock parameters.

Step 3.

Evaluate soil and rock parameters for design and establish resistance factors.

Step 4.

Select lateral earth pressure distribution acting on back of wall for the final wall height. Evaluate water, surcharge, and seismic pressures.

Step 5.

Evaluate factored total lateral pressure diagram for all appropriate limit states.

Step 6.

Evaluate embedment depth of vertical wall element and factored bending moment in the wall.

Step 7.

Check flexural resistance of vertical wall elements. Check combined flexural and axial resistance (if necessary).

Step 8.

Select temporary lagging (for soldier pile and lagging wall).

Step 9.

Design permanent facing (if required).

Step 10.

Estimate maximum lateral wall movements and ground surface settlements at the Service Limit State. Revise design if necessary.


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•

The ratio of overburden pressure (computed at the elevation of the base of the excavation) to undrained shear strength of the soil below should be less than 3 at the elevation of the base of the excavation.

•

The active earth pressure shall not be less than 0.25 times the effective overburden pressure at any depth or 0.035 ksf/ft of wall height, whichever is greater.

For nongravity cantilevered wall designs involving clayey soils, the strength of the clay changes with time and consequently the earth pressure changes with time also. Immediately after the sheet piling is installed, the lateral earth pressures can be calculated based on the total stress method of analysis, i.e., using undrained shear strength parameters. For permanent wall design, both the short and long-term conditions need to be considered. For the long term condition, sheet piling in clay should be analyzed using effective strength parameters, c΄ and φ΄, obtained from triaxial shear strength tests. Assuming that the effective cohesion value is small, the value of c΄ can conservatively be taken as zero. The long term condition can then be analyzed based solely on the effective friction angle, φ΄, of the clay. Information on the maximum wall friction (δ) for design is provided in Section 3.6. Where the wall may settle relative to the retained ground, assume δ=0. Other sources of lateral pressure including lateral pressures due to surcharges, seismic pressures and water should be added to these diagrams using corresponding load factors for each applicable limit state (as described in Section 8.9.2.3). On the passive side of the excavation, for nongravity cantilevered walls with discrete elements in competent ground, it is assumed that loading is resisted over a width equal to 3 times the width of the section, i.e., 3b, where the width of the section is the width of the flange or diameter of the element for driven sections and the diameter of the concrete-filled hole for sections encased in structural concrete. The maximum width of 3b can be used when material in which the vertical element is embedded does not contain discontinuities that would affect the failure geometry. The width should be reduced if planes or zones of weakness would prevent mobilization of resistance through this entire width, or if the passive resistance zones of adjacent elements overlap. For drilled-in soldier piles, the effective width b is assumed to be equal to the diameter of the drill hole provided the embedded portion of the drill hole is backfilled with structural concrete (as indicated previously) or lean mix concrete with a compressive strength of no less than 50 psi. Above the bottom of the excavation, discrete elements (e.g., soldier piles) are designed for the total lateral force in the span distance between adjacent soldier piles. Below the bottom of the excavation, the active earth pressure is assumed to act only on the soldier pile width, b. FHWA NHI-07-071 Earth Retaining Structures

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8.9.2.3 Step 5: Evaluate Factored Loads For this step, the active earth pressures, surcharge pressures, seismic pressures, and other sources of lateral pressure are factored using the load factors in AASHTO Tables 3.4.1-1 and 3.4.1-2 for all appropriate limit states. Passive resistance developed on the excavation side of the wall uses a resistance factor of 0.75 (see Table 11.5.6-1 in AASHTO 2007). 8.9.2.4 Step 6: Evaluate Embedment Depth and Factored (Maximum) Bending Moment With the factored total lateral pressure diagrams evaluated for the active and passive side of the wall, the following calculation steps are used to design the vertical wall element: 1. 2.

3.

The required embedment depth corresponds to the depth where factored moments due to active side loading are equivalent to factored moments resulting from passive side resistance. Determine the depth at which the shear force in the wall is zero. This corresponds to the depth where the resultant force from the active side of the wall is equivalent to the resultant force from the passive side of the wall (based on factored loads). Calculate the factored bending moment (maximum bending moment) at the point of zero shear.

These steps are most easily performed using a spreadsheet where forces and moments are calculated at various assumed embedment depths. Figure 8-57 shows an example for a nongravity cantilevered sheet-pile wall with a height of 10 ft retaining a granular soil. The retained ground has a friction angle of 30 degrees which results in a calculated KA = 0.33 and Kp =3.0 (assuming δ = 0 and using equations 3-2 and 3- 3 for KA and KP, respectively). In this example, it is assumed that only lateral pressures resulting from active earth pressures are present with a load factor of 1.5 (i.e., maximum load factor for horizontal earth pressure for active conditions based on AASHTO Table 3.4.1-2). In current practice, sheet pile wall designs may be performed using computer programs such as CWALSHT (1995), Pile Buck (2004), and others. These programs, however, are not LRFD based. AASHTO Figures 3.11.5.6-3, -6, and -7 address nongravity cantilevered wall designs for continuous elements, such as sheet piles. These figures indicate that the length required to obtain moment equilibrium (based on the procedure described above using factored loads and resistances) should be increased by 20 percent. The design method used to evaluate the required wall embedment is based on achieving moment equilibrium. This method is referred to as the simplified method in that it is assumed that the difference between the passive resistance at the back of the wall and the active pressure at the front acts as a concentrated force, F, at the toe. By taking moments about the toe (thereby eliminating the force F from the equation), the depth of embedment, do, is found. Because of this FHWA NHI-07-071 Earth Retaining Structures

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Cantilevered Sheet Pile Wall design

LRFD

γ= Ka =

10'

125 pcf 0.33

γp =

1.5

Pa

L La

Pp Lp A Kp =

3

ϕp =

0.75

Factored Pa = Factored Pp =

2

γp * 0.5 * (L+10) * Ka * γ 2 ϕp * 0.5 * L * Kp * γ

Summing Moments at pile tip (point A) Factored Pa (lbs)

L (ft) 0 2 4 6 8 8.5 8.9 14 15 15.2 20 22 24 26

-3093.8 -4455.0 -6063.8 -7920.0 -10023.8 -10588.4 -11051.2 -17820.0 -19335.9 -19646.6 -27843.8 -31680.0 -35763.8 -40095.0

La (ft)

Ma (ft-lbs)

= (L+10)/3 3.33 4.00 4.67 5.33 6.00 6.17 6.30 8.00 8.33 8.40 10.00 10.67 11.33 12.00

=Pa*La -10312.5 -17820.0 -28297.5 -42240.0 -60142.5 -65294.9 -69622.5 -142560.0 -161132.8 -165031.0 -278437.5 -337920.0 -405322.5 -481140.0

Factored Pp (lbs) 0.0 562.5 2250.0 5062.5 9000.0 10160.2 11138.9 27562.5 31640.6 32490.0 56250.0 68062.5 81000.0 95062.5

Perform. Ratio

Lp (ft)

Mp (ft-lbs)

=L/3 0.00 0.67 1.33 2.00 2.67 2.83 2.97 4.67 5.00 5.07 6.67 7.33 8.00 8.67

=|Mp/Ma| =Ma+Mp =Pp*Lp 0.0 0.00 -10312.5 375.0 0.02 -17445.0 3000.0 0.11 -25297.5 10125.0 0.24 -32115.0 24000.0 0.40 -36142.5 28787.1 0.44 -36507.8 33045.4 0.47 -36577.0 128625.0 0.90 -13935.0 158203.1 0.98 -2929.7 164616.0 1.00 -415.0 375000.0 1.35 96562.5 499125.0 1.48 161205.0 648000.0 1.60 242677.5 823875.0 1.71 342735.0

Based on above: Required embedment for (factored resist = factored load) = Maximum factored moment occurs at Maximum factored moment = Required section modulus for A-36 steel Resistance factor for flexure Factored resistance ϕ * fy = Z = M/(ϕ * fy) =

M (ft-lbs)

15.2 ft 8.9 ft -36577.0 ft-lbs

0.9 32.4 ksi 3 13.55 in

Figure 8-57. Design Analysis for Nongravity Cantilevered (Sheet Pile) Wall. FHWA NHI-07-071 Earth Retaining Structures

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simplification, the value of do is slightly less than d obtained from the full method (i.e., fixed earth method). To account for this, typical practice is to increase do by a small amount (up to 20 percent). However, for analyses of discrete elements (see AASHTO Figures 3.11.5.6-1, 2, -4, and -5), this increase of 20 percent is specifically not required. It is noted that for ASD design, the minimum factor of safety with respect to passive resistance is 1.5. For LRFD (for both continuous and discrete element walls), the resistance factor used for passive resistance evaluations is 0.75 (see Table 11.5.6-1 in AASHTO 2007). Assuming only active earth pressure loadings on the active side with a load factor of 1.5 indicates that the LRFD approach provides a corresponding ASD FS value of approximately 2.0 (i.e., 1.5/0.75). It is noted that in the 2002 Interim AASHTO LRFD Specifications, the resistance factor for passive resistance was 1.00. This information indicates that embedment depths computed for nongravity cantilevered walls with continuous wall elements using LRFD will likely be greater than those computed using ASD. 8.9.2.5 Step 7: Check Flexural Resistance of Wall In general, nongravity cantilevered walls comprise either a structural steel section (e.g., steel sheet-pile or soldier beam) or reinforced concrete (e.g., secant shafts). To date, LRFD-based formulations for the structural evaluation of vinyl or composite sheet piles are not available. For structural steel vertical wall elements (which are not subject to significant axial loads), the design check for flexural resistance can be written as:

M max ≤ ϕM n = ϕ Fy Z

(8-1)

where: Mmax = maximum factored bending moment in the wall element Mn = nominal flexural resistance Fy= Steel yield stress Z = Plastic section modulus φ =0.90 for flexure (see Table 11.5.6-1 in AASHTO, 2007)


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The plastic section modulus for sheet piles can be obtained from the sheet pile producer. For resistance factors for combined axial and flexural resistance of structural steel elements, refer to AASHTO (2007) Section 6.5.4.2. For reinforced concrete walls, the check for flexural resistance follows the methods presented in AASHTO (2007) Section 5.7.3. 8.9.2.6 Step 8: Select Temporary Lagging for Soldier Pile and Lagging Wall Timber lagging is only used for support of temporary loads applied during excavation, however, pressure-treated timber lagging has been used to support permanent loads. The contribution of the temporary lagging is not included in the structural designs of the final wall face. Temporary lagging is not designed by traditional methods, rather lagging is sized from charts developed based on previous project experience which accounts for soil arching between adjacent soldier piles. Table 8-11 presents experience-based recommendations for selecting wood lagging thickness based on soil type, depth of excavation and clear distance between soldier piles. As noted, permanent timber lagging has been used in lieu of a concrete face to carry permanent wall loads. For permanent applications, the timber grade and dimensions should be designed according to structural guidelines. Several problems may exist for permanent timber lagging including: (1) need to provide fire protection for the lagging; and (2) limited service life for timber. 8.9.2.7 Step 9: Design Permanent Facing For nongravity anchored walls with discrete vertical elements, permanent facing should be designed to resist factored earth pressures, surcharges, water pressures, and seismic pressures for all appropriate limit states. The factored bending moments in the permanent facings can be estimated using Table 8-12. Permanent facings that are cast-in-place (CIP) are typically 8 to 12 in. thick. This thickness will typically ensure that the wall is structurally sound and allow for some deviations in soldier beam placement. Significant deviations, however, in soldier beam alignment may require that additional concrete in excess of that required for the nominal thickness of the wall be used so that the finished wall face is properly aligned. Precast concrete facing may be cost-effective if there is a local fabricator and if there is adequate on-site storage. Precast panels are designed as simple spans between the soldier beams.


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1) 2) 3)


Notes:

CL, CH ML, SM-SL

Clay, heavily overconsolidated, fissured

Cohesionless silt or fine sand and silt below water table CL, CH

SC

Clayey sands (medium dense to dense) below water table

Soft clays γH >5 Su

SW, SP, SM

Sand and silty sand (loose)

CL, CH

CL, CH

GW, GP, GM, GS, SW, SP, SM

Sands and gravels (medium dense to dense)

Clays (stiff to very stiff); non-fissured Clays, medium consistency and γH <5 Su

ML, SM-ML

Silt or fine sand and silt above water table

0 - 16

26 - 60

0 - 26

26 - 60

0 - 26

Depth (ft)

3 in

3 in

3 in

3 in

2 in

5 ft

3 in

3 in

3 in

3 in

3 in

6 ft

4 in

4 in

3 in

3 in

3 in

7 ft

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-----------------

---------

---------

5 in

4 in

4 in

4 in

9 ft

6 in

5 in

4 in

4 in

4 in

3 in

8 ft

Recommended thickness of lagging (roughcut) for clear spans of:

Slightly plastic silts below ML 16 - 26 3 in 4 in 5 in water table Clayey Sands (loose), below SC 26-36 4 in 5 in 6 in water table In the category of "potentially dangerous soils", use of soldier pile and lagging wall systems is questionable. The values shown are based on construction grade lumber. Local experience may take precedence over recommended values in this table.

POTENTIALLY DANGEROUS SOILS

DIFFICULT SOILS

COMPETENT SOILS

Unified Soil Classification

Soil Description

Table 8-11. Recommended Thickness of Wood Lagging (after Goldberg et al., 1976)

---------

--------

--------

5 in

5 in

5 in

4 in

10 ft

Table 8-12. Bending Moments for Facing Design (AASHTO, 2007). Factored moment on a unit width or height of facing

Support and soil condition Simple span No soil arching (e.g., soft cohesive soils; rigid concrete facing placed tightly against soil)

p l 2/8

Simple span Soil arching (e.g., granular soil or stiff cohesive soil with flexible facing; rigid facing where space is available to allow in place soil to arch)

p l 2/12

Continuous facing No soil arching (e.g., soft cohesive soils; rigid concrete facing placed tightly against soil)

p l 2/10

Continuous facing Soil arching (e.g., granular soil or stiff cohesive soil with flexible facing; rigid facing where space is available to allow in place soil to arch)

p l 2/12

Note: p = average factored lateral pressure, including earth, surcharge, and water pressure acting on the section of facing being considered l = spacing between vertical elements or other facing supports

8.9.2.8 Step 10: Evaluate Lateral Displacements at the Service Limit State Lateral wall displacements may be estimated using beam on elastic foundation analyses or other soil-structure interaction analysis methods. These methods are discussed in Section 8.12.2.

8.10

ANCHORED BULK HEAD (DEADMAN ANCHOR) WALLS

8.10.1 Overview Anchored bulk head walls derive their lateral support from the passive resistance on the front of the embedded portion of the wall, and the pull out resistance of the anchors. Anchors reduce the required depth of penetration of vertical wall face elements and also permit wall heights to be increased to about 35 ft, depending on the soil and groundwater conditions. For higher walls the use of high- strength steel sheet piling, reinforced sheet piling, or additional levels of anchors may be necessary. The methods of analysis presented in this section apply for walls with a single anchorage level and which are built from the bottom-up. LRFD for these wall types is not addressed in AASHTO (2007) and therefore all design information presented herein is in ASD. FHWA NHI-07-071 Earth Retaining Structures

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Anchored walls with one level of anchors and which are built from the bottom-up are designed using triangular active and passive earth pressure distributions. Using these applied pressures, there are two basic methods for design of anchored sheet pile walls: (1) the free earth support method; and (2) the fixed earth support method. In selecting the method to be used for wall design, consideration must be given to a number of factors such as the relative stiffness of the sheet piles, the depth of pile penetration, the relative compressibility and strength of the soil, and the amount of anchor deflection. Figure 8-58 shows the assumed deflection and bending moment distribution for the two methods.

Figure 8-58. Variation of Deflection and Bending Moment (a) Free Earth Support Method and (b) Fixed Earth Support Method (after Das, 1990).

Typically, the fixed earth method is used in granular soils and stiff cohesive soils and for walls of relatively low stiffness; the free earth method is used in cohesive soils, for walls of relatively higher stiffness, and for situations where the sheeting penetration may be restricted because of obstructions, a shallow rock surface, or sheeting length limitations. The fixed earth method generally results in smaller anchor loads and smaller wall bending moments, but a greater depth of sheeting penetration than the free earth method. 8.10.2 Free Earth Support This method is based on the following assumptions (Teng, 1962 and USS, 1975):

•

The soil into which the piling is placed is incapable of producing effective restraint from passive resistance to the extent necessary to induce negative bending moments;


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• • • •

The piling is placed just deep enough to assure stability, assuming that the passive resistance is fully mobilized; No pivot point exists along the embedded portion of the wall, i.e., no passive resistance develops on the back side of the piling (in the active zone); The piling is free to rotate but cannot deflect laterally at the anchor level; and The piling is perfectly rigid as compared to the surrounding soils.

With the above assumptions, the problem can be solved by considering static equilibrium. Figure 8-59 presents analysis by the free earth support method for two common cases. The free earth support method assumes a rigid piling and therefore uses lateral pressures according to a triangular distribution as shown in Figure 8-59. However, in reality, many wall systems are quite flexible causing the earth pressure to redistribute and thus differ from this triangular distribution. With increasing flexibility, the embedded portion of the wall rotates about its lower edge (Figure 8-58a) causing the center of the passive resistance to move closer to the exposed wall base. This in turn decreases the maximum bending moment. Figure 8-60 presents a procedure to reduce the maximum design moment obtained from the free earth support method based on wall flexibility considerations. The moment reduction factor (Md/Mmax in Figure 8-60a) should be used where a factor of safety was applied to the soil shear strength parameters prior to computing the penetration depth. Also, it is recommended that the reduced bending moment not be lower than that computed by the fixed earth method. 8.10.3 Fixed Earth Support The fixed earth support method is based on the assumption that the toe of the wall is restrained from rotation as shown in Figure 8-58b. The deflected shape reverses its curvature at the point of contraflexure, c, and becomes vertical at the toe of the wall. The problem may be solved using the theory of beam on elastic foundation, but this procedure is laborious. In practice, a procedure known as the equivalent beam method proposed by Blum is used. This method utilizes a theoretical relationship between the angle of internal friction (φ) and the distance below the exposed wall base to the point of contraflexure. Figure 8-61 presents the equivalent beam method, which is limited in its use to granular soils. For cohesive soils, use methods based on the theory of beam on elastic foundation (Hetenyi, 1946).


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8-60a

Figure 8-59a. Analysis by Free Earth Support Method for Sheet Piling in Granular Soils (after Teng, 1962 and USS, 1975). FHWA NHI-07-071 Earth Retaining Structures

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7.

Select pile section for the maximum moment or use the moment reduction theory, Figure 860b.

8.

Add 20 to 40 percent D’ to provide for safety margin, or divide qu by a factor of safety of 1.5 to 2.0 in above equations.

Figure 8-59b. Analysis by Free Earth Support Method for Sheet Piling Backfilled in Granular Soil and Embedded in Cohesive Soil (after Teng, 1962 and USS, 1975). FHWA NHI-07-071 Earth Retaining Structures

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Plot of log ρ against Md/Mmax for sheet piles penetrating sand. Md is design or reduced moment; Mmax is theoretical maximum moment computed as per Figure 8-59a; Md < Mmax. 1.

Choose a sheet pile section.

2.

Find the section modulus, S, of the selected section per unit length of the wall.

3.

Determine the moment of inertia, I, of the section per unit length of wall.

4.

Obtain the total height of the sheet pile, H+D, and calculate the relative flexibility of the wall, ρ. as follows:

ρ = 10.91 x 10-7 [(H+D)4/(EI)} where (H+D) is in meters, E is the modulus of elasticity of the pile material in MPa and I, the moment of inertia, is in m4/m of wall. 5.

Find log ρ.

6.

Find the moment capacity of the pile section chosen in Step 1 as Md = σallS where σall is the allowable stress in pile material.

7.

Determine Md/Mmax.

8.

Plot log ρ (Step 5) and Md/Mmax in the above Figure.

9.

Verify that the plotted point for the selected pile section falls above the curve (loose sand or dense sand, as appropriate). If the plotted point falls below the curve, select another sheet pile section and repeat the above Steps 2 through 9.

Figure 8-60a. Moment Reduction for Anchored Sheet Pile Wall Analyzed by Free Earth Support Method for Granular Soils (after Rowe, 1952; Rowe, 1957; and Das, 1990). FHWA NHI-07-071 Earth Retaining Structures

8 – Non-Gravity Cantilevered and Anchored Walls 8-100 June 2008

1. Compute the Stability Number, Sn, as follows: Sn = (cu / γeqH)(1 + ca / cu)0.5 where cu(=qu/2) is the undrained cohesion below the exposed base of wall, ca is wall adhesion (See Table 3-2). For design purposes, Sn =1.25(cu/γeqH) may be used. Any sheet piling driven into cohesive soils should have a minimum Sn of 0.3 times a desired factor of safety; otherwise failure may occur. 2. Compute α as α = H / (H+D). 3. Compute ρ as per Equation in Figure 8-60a. 4. For the magnitudes of α and Sn obtained above, determine the Md/Mmax for various values of log ρ from above charts and plot Md/Mmax against log ρ. Values of log ρ can be interpolated. 5. Verify that the plotted point for the selected pile section falls above the curve developed in Step 4. If the plotted point falls below the curve, select another sheet pile section and repeat the above Steps 1 through 5. Figure 8-60b. Moment Reduction for Anchored Sheet Pile Wall Analyzed by Free Earth Support Method for Cohesive Soils (after Rowe, 1952; Rowe, 1957; and Das, 1990). FHWA NHI-07-071 Earth Retaining Structures


8.10.4 Deadman Anchor Systems A deadman anchor, if used, must be located outside the potential active failure zone developed behind the sheet pile wall as shown in Figure 8-62. If the anchor system must be located closer to the wall (see Figure 8-62), anchorage resistance is decreased and an additional passive reaction is required for stability of the wall base. The tie rods must be protected to resist corrosion. Typically, the rods are coated with one or more applications of tar-base paint, and spirally wrapped with a durable fabric or fiber glass tape after the application of the first coat of paint. The tie rods may also be enclosed in a rigid casing or supported vertically to eliminate sag for cases where the backfill will settle significantly or unevenly.

8.11

ANCHORED WALL DESIGN

The design methods presented in this section have been developed primarily for anchored walls with relatively flexible vertical wall elements (i.e., sheet piles or soldier piles) (see GEC No. 4 (Sabatini et al., 1999). A key distinction in the design of flexible anchored walls (as compared to stiff anchored walls) is in the earth pressure diagram used for design analyses. For flexible wall systems constructed from the “top-down”, the wall deformation pattern is complex and may not be consistent with the development of a theoretical Rankine or Coulomb (i.e., triangular) earth pressure distribution over the full height of the wall. For example, higher than active earth pressures develop at the upper anchor location since the upper anchor restrains the wall from moving outward sufficiently to locally cause a reduction of earth pressures to the active state. For this reason, flexible anchored walls may be designed using apparent earth pressure diagrams. The use of apparent earth pressure diagrams has resulted in reasonable estimates of ground anchor loads and conservative estimates of wall bending moments between anchors for flexible walls constructed in competent soils. In general, this approach may also be used for the design of stiff anchored wall systems; however, experience indicates that stiff anchored wall designs based on apparent earth pressure diagrams may be overly conservative and therefore expensive. Apparent pressure diagram analyses permit relatively simple “hand-calculations” of ground anchor loads and wall bending moments. They represent an envelope that can be used to develop an adequate anchored system for the entire history of the excavation, but they do not provide actual loads that might exist on the wall at any time. Where an assessment of the FHWA NHI-07-071 Earth Retaining Structures


Figure 8-61. Analysis by Equivalent Beam Method (after USS, 1975 and Teng, 1962).



Figure 8-62. Effect of Anchor Location Relative to the Wall (after NAVFAC, 1986).



actual loads on the wall is required, staged construction analyses such as soil-structure interaction analyses (e.g., beam on elastic foundation) may be used. Staged construction analyses may also be required where: (1) the wall is influenced by loadings from nearby foundations; (2) large surcharge loadings need to be resisted by the wall; or (3) the ground is relatively soft or weak or and/or there are preexisting instabilities or planes of weakness in the retained soil. The use of soil structure interaction analyses for the design of anchored wall systems is discussed in Section 8.12.6. The major design steps for an anchored wall are outlined in Table 8-13 and Steps 1 and 2 have been discussed in previous portions of this manual. Soil and rock parameter evaluation (as part of Step 3) has been discussed in Section 2. Table 8-13. Design Steps for Anchored Walls. Step 1.

Establish project requirements including all geometry, external loading conditions (temporary and/or permanent, seismic, etc.), performance criteria, and construction constraints.

Step 2.

Evaluate site subsurface conditions and relevant properties of in situ soil and rock parameters.

Step 3.

Evaluate soil and rock parameters for design, establish resistance factors, and select level of corrosion protection for the anchor.

Step 4.

Select lateral earth pressure distribution acting on back of wall for the final wall height. Evaluate appropriate water, surcharge, and seismic pressures.

Step 5.

Evaluate factored total loads for all appropriate limit states.

Step 6.

Calculate horizontal ground anchor loads and subgrade reaction force. Resolve each horizontal anchor load into a vertical force component and a force along the anchor. Evaluate horizontal spacing of anchors based on wall type and calculate individual factored anchor loads.

Step 7.

Evaluate required anchor inclination based on right-of-way limitations, location of appropriate anchoring strata, and location of underground structures.

Step 8.

Select tendon type and check tensile resistance.

Step 9.

Evaluate anchor bond length.

Step 10.

Evaluate factored bending moments and flexural resistance of wall.

Step 11.

Evaluate bearing resistance of wall below excavation subgrade. Revise wall section if necessary.



Step 12.

Evaluate overall stability of anchored system at the service limit state. Revise ground anchor geometry if necessary.

Step 13.

Estimate maximum lateral wall movements and ground surface settlements at the service limit state. Revise design if necessary.

Step 14.

Select lagging. Design walers, facing drainage systems, and connection devices.

8.11.1 Step 3: Select Corrosion Protection of Ground Anchors 8.11.1.1Overview The corrosion protection system for a ground anchor consists of components that combine to provide an unbroken barrier for each part of the tendon and the transitions between them. Steel components of the anchor include the anchor head, bearing plate, trumpet, prestressing steel, and couplers (where used). Components of the corrosion protection system include: (1) for the anchorage, a cover or concrete embedment, a trumpet, and corrosion inhibiting compounds or grout; (2) for the unbonded (free) length, grout and a sheath filled with a corrosion inhibiting compound or grout; and (3) for the bond length, grout and encapsulations with centralizers and/or epoxy coatings. The components of corrosion protection for bar and strand tendons are shown in Figures 8-63 to 8-65 and brief descriptions are provided below. 8.11.1.2

Protection of Free Length

The free length is protected by a variety of means since the protection is not required to transfer stresses from the tendon to the ground. The protection must: a. b. c. d.

Accommodate elastic movements during testing and stressing; Withstand potential damage during storage, handling, and construction; Resist attack in aggressive environments; and Allow elastic movements after lock-off.

The most common methods of protecting the free length include the use of smooth sheaths filled with anti-corrosion grease, heat shrink sleeves, and secondary grouting after stressing. Except for secondary grouting, the protection is usually in place prior to inserting the tendons in the hole. FHWA NHI-07-071 Earth Retaining Structures


Figure 8-63. Examples of Corrosion Protection for Anchorages (a) Strand Tendon and (b) Bar Tendon (after Sabatini et al., 1999).



(a)

(b) Figure 8-64. Simple Corrosion Protected Tendons (a) Strand Tendon and (b) Bar Tendon (after Sabatini et al., 1999). FHWA NHI-07-071 Earth Retaining Structures


(a)

(b)

Figure 8-65. Encapsulated Double Corrosion Protection (a) Strand Tendon and (b) Bar Tendon (after Sabatini et al., 1999).



Secondary grouting typically involves filling the annular space between the ground and the smooth sheath in the free length. Secondary grouting should be done by gravity or low pressure methods to prevent damage to the sheath and load transfer to the ground in the free length. The grout should terminate before the bearing plate area to prevent load transfer up the grout column to the structure. Such secondary grouting, when done properly, provides an extra measure of protection and is recommended for most installations. Secondary grout, in the true sense, refers to grout placement after stressing and lock-off around a tendon with an unsheathed free length. After set, the grout surrounding the free length becomes bonded to the ground and the tendon over the initially unbonded length. These anchors are only normally recommended for semi-permanent, low-risk applications. When secondary grouting is used for these anchors, extreme care must be taken to insure no voids in the grout exist beneath the anchor bearing plate. Such areas are the most critical to protect against corrosion and require complete encapsulation by cement grout. 8.11.1.3

Protection of Bond Length

Figures 8-64 and 8-65 show several schemes for corrosion protection of the tendon bond length. These are briefly described below: a. Simple (Class II) Protection: The use of simple protection (Figure 8-64) relies on Portland cement grout to protect the tendon, bar or strand, in the bond length. Steel will not corrode in a high pH environment such as Portland cement which possesses pH values up to 12.6. When Portland cement is used for protection, it is assumed that the pH will not be lowered with time. When the simple protection is used, care should be taken that the tendon has at least 0.5 in. of grout cover and the anchor grout extends 24 in. over the bottom of the free length sheathing. b. Double (Class I) Protection: Complete encapsulation of the anchor tendon is accomplished by uniformly corrugated plastic or steel tube (Figure 8-65). The tube must be capable of withstanding the deformations associated with transportation, installation, stressing and testing of the anchor, and transferring the load applied to the tendon. Regardless of the encapsulating medium, the annular space between the corrugated tube and tendon is usually filled with neat cement grout containing admixtures to control bleed of water from the grout. Shorter tendons are grouted before insertion in the hole. 8.11.1.4

Protection of the Anchor Head

The most critical area to protect from corrosion is in the vicinity of the anchor head connection. Below the bearing plate, the corrosion protection over the free length is usually FHWA NHI-07-071 Earth Retaining Structures


terminated to expose the bare tendon. Above the bearing plate, the bare tendon is gripped by either wedges, nuts, or deformed in the case of wires. Regardless of the type of tendon, the gripping mechanism creates stress concentrations at the connection. In addition, a very aggressive corrosion environment may exist at the anchor head since oxygen is readily available. The vulnerability of this area is demonstrated by the fact that most anchor failures that have been reported have occurred within a short distance of the anchor head. Typical protection system for the anchor head includes the use of a trumpet. A “trumpet,” usually of steel or strong durable plastic, is used to overlap with the free length corrosion protection, and to protect the short exposed length of the tendon below the anchor plate. One end of the trumpet is fastened and sealed to the bearing plate and the other fitted with a deformable seal that fits tightly around the protective tube, but allows free tendon movement within the trumpet. The annular space between the trumpet and the tendon is filled with anticorrosion grease. The anchor head, including exposed tendon and friction grips or locking nuts above the bearing plate, is protected by a cap filled with anti-corrosive grease or by embedment in concrete. Covering the anchor head with a grease-filled cap allows future lift-off tests and/or load adjustment. When filling the trumpet and cap, care is required to ensure that the grease or grout fills the entire space. Protection of the bearing plate is accomplished by painting it with bituminous or other corrosion-resistant materials. Care should be taken to ensure that the applied coating is compatible with the materials selected for both inner head and outer head protection (Littlejohn, 1990). 8.11.1.5

Selection of Corrosion Protection Level

All permanent ground anchors must be designed against corrosion. Depending on anticipated corrosion rates and the design life of a temporary earth retaining structure, corrosion protection measures may also be required for temporary anchors. A selection flowchart for ground anchor corrosion protection is shown in Figure 8-66. As shown in the flowchart the minimum level of corrosion protection for ground anchors should be selected considering the service life of the anchored system, the aggressivitiy of the ground environment, the consequences of failure of the anchored system, and the cost of providing a higher level of corrosion protection.



TEMPORARY SUPPORT OF EXCAVATION

PERMANENT

SERVICE LIFE

AGGRESSIVITY

NOT KNOWN OR AGGRESSIVE

NON-AGGRESSIVE

CLASS II PROTECTION

NONE

AGGRESSIVITY

NOT KNOWN OR AGGRESSIVE

NON-AGGRESSIVE

CONSEQUENCES OF FAILURE OF ANCHORED SYSTEM

CLASS I PROTECTION

SERIOUS

NOT SERIOUS

COST FOR INCREASING PROTECTION LEVEL

CLASS I PROTECTION

SMALL

SIGNIFICANT

CLASS I PROTECTION

CLASS II PROTECTION

Note: Class I also means Simple Protection and Class II also means Double Protection.

Figure 8-66. Decision Tree for Selection of Corrosion Protection Level (after PTI, 1996).

Service Life of the Anchored Structure Service life is used to distinguish between a temporary support system and a permanent anchor application. The service life of a temporary earth support system is based on the time required to support the ground until a permanent earth retention system is constructed. The time period for temporary systems is commonly stated be 18 to 36 months but may be shorter or longer based on actual project conditions. If the service life of a temporary support of excavation anchor is likely to be extended, an evaluation should be made to determine whether or not to provide additional corrosion protection for the tendon, particularly in aggressive ground conditions.

Aggresivity of the Ground Environment Ground anchors in environments classified as aggressive or of unknown aggressivity will require the highest class of corrosion protection listed for each service life classification, FHWA NHI-07-071 Earth Retaining Structures


Class II corrosion protection (or Simple Protection) for temporary support of excavation anchors and Class I corrosion protection (or Double Protection) for permanent anchors. Tests and/or field observations are used to classify the aggressivity of the ground environment. In general, ground environments may be classified as aggressive if any one of the following conditions are present in the ground or may be present during the service life of the ground anchor (PTI, 1996) (Table 8-14). Table 8-14. Criteria for Electrochemical Properties of Soils for Ground Anchor Applications (after Cheney, 1988). Property

Criteria

Test Method

Resistivity pH Chlorides Sulfates

< 2000 ohm-cm < 4.5 >100 ppm > 3000 ppm

AASHTO T-288 AASHTO T-289 AASHTO T-291 AASHTO T-290

The presence of stray currents or buried concrete structures adjacent to the anchored system project which have experienced corrosion or direct chemical (acid) attack are also indication of aggressiveness of the ground. Tests from a nearby site can be used to evaluate the aggressivity of the site if the designer can establish that the ground conditions are similar. Otherwise, if site-specific aggressivity tests are not performed, then the ground should be assumed to be aggressive. The following ground environments are always considered aggressive: (1) soil or groundwater with a low pH; (2) salt water or tidal marshes; (3) cinder, ash, or slag fills; (4) organic fills containing humic acid; (5) peat bogs; and (6) acid mine drainage or industrial waste. Classification of ground aggressivity should consider the possibility of changes during the service life of the ground anchor, which may cause the ground to become aggressive, such as, might occur near mining operations, chemical plants, or chemical storage areas.

Consequences of Failure of the Anchored System For permanent anchors, if failure of the anchored system could result in serious consequences such as loss of life or significant financial loss, a minimum of Class I protection is required. The consequences of failure are considered serious for: (1) anchored systems used in urban areas where there are nearby structures behind the wall; (2) anchored systems used for a FHWA NHI-07-071 Earth Retaining Structures


highway retaining wall where the closure of one or two lanes would cause a major disruption of traffic; and (3) landslide stabilization walls where the retained slope has experienced past movement.

Cost for a Higher Level of Protection The final criterion for selecting the class of corrosion protection is the increased cost for changing from Class II (Simple) protection to Class I (Double) protection. For the same tendon, Class I protected anchors require a larger drill hole as compared to a Class II protected anchor. Encapsulating an anchor tendon increases the required drill hole size which may result in increased installation costs. In an uncased drill hole, the additional drilling costs can be small, and the owner may elect to use Class I protection. In a cased hole or in rock, the additional drilling costs can be higher, and the owner will decide if the benefit of providing a higher level of corrosion protection is worth the additional cost. The increase in drill hole diameter may result in a need to increase bearing plate dimensions, trumpet diameter, and the opening in the soldier beam to insert the tendon. 8.11.2 Step 4: Select Earth Pressure Diagram Recommended apparent earth pressure diagrams for sands, stiff to hard clays, and soft to medium clays are provided in Figures 8-67, 8-68, and 8-69. These “apparent earth pressure diagrams” are semi-empirical diagrams developed to account for the relatively complex deformation pattern associated with relatively flexible anchored systems. 8.11.2.1 Apparent Earth Pressure Diagram for Sands For the apparent earth pressure diagram for sands (i.e., Figure 8-67), the value for KA is the Rankine earth pressure coefficient, γ is the effective unit weight of the retained ground and H is the total height of the excavation. 8.11.2.2 Apparent Earth Pressure Diagram for Stiff to Hard Clays The apparent earth pressure diagram for stiff to hard clays (Figure 8-68) should only be used where for support of a temporary excavation (i.e., relatively short duration) and there is no available free water. If these either of these conditions are not met, an apparent earth pressure diagram for long-term (permanent) conditions using drained strength parameters should be evaluated. In this case, use the drained friction angle of the stiff to hard clay with the sand apparent earth pressure diagram shown in Figure 8-67 and explicitly add in water pressures.



Figure 8-67. Apparent Earth Pressure Diagram for Sand.

When using Figure 8-68, a maximum pressure ordinate of 0.3γH should be used where anchors are locked off at 75 percent or less of the anchor load (based on unfactored loads) and 0.4γH where anchors are locked off at 100 percent of the anchor load (based on unfactored loads).



Figure 8-68. Apparent Earth Pressure Diagram for Stiff to Hard Clays.

8.11.2.3 Apparent Earth Pressure Diagram for Soft to Medium Clays The apparent earth pressure diagram for use in designing temporary wall systems in soft to medium clays (Figure 8-69) is based on the well-known Terzaghi and Peck (1967) diagram for soft to medium clays. The Terzaghi and Peck (1967) diagrams, however, did not account for the development of soil failure below the bottom of the excavation. Observations and finite element studies have demonstrated that soil failure below the bottom of the excavation can lead to very large movements for temporary retaining walls in soft clays. Henkel (1971)



used a limit equilibrium method to obtain KA for use in evaluating the maximum pressure ordinate for the soft to medium clay apparent earth pressure diagram.

Figure 8-69. Apparent Earth Pressure Diagram for Soft to Medium Clay. For the apparent earth pressure diagram for soft to medium clays (Figure 8-69), the maximum pressure ordinate, pa, is given as pa=KAγH where KA is given as:

K A =1 −

4S u d  1 − 5.14S ub + 2 2  γH γH H

  ≥ 0.22 

(8-2)

where: Su = Undrained shear strength of retained ground Sub = Undrained shear strength of ground below the bottom of the excavation d = depth of potential base failure below base of excavation This method is based on a total stress approach, i.e., water pressures are already accounted for in the diagram. Additional information on the use of this diagram is provided in Sabatini et al. (1999). A computation showing the evaluation of KA according to Eq. 8-2 is provided in Student Exercise 7.



8.11.3 Step 5: Evaluate Factored Loads For this step, the apparent earth pressures, water, surcharge loads, seismic pressures, and other sources of lateral load on the wall are factored using the load factors in AASHTO (2007) Tables 3.4.1-1 and 3.4.1-2 for all appropriate limit states. Apparent earth pressure diagrams were originally developed based on wall pressures backcalculated from field measurements of strut loads in internally braced excavations. Application of these diagrams will generally produce conservative design loads, implying that if a strut (or ground anchor) load would be equivalent to the calculated load from the apparent pressure diagram at that location, the other strut (or ground anchor) loads would necessarily be less than that calculated from the apparent pressure diagrams. That is, the apparent pressure diagrams represent an envelope of maximum pressures. Using AASHTO (2007), a load factor of 1.35 is applied to these apparent earth pressure distributions. The use of this load factor represents a relatively large amount of conservatism for LRFD-based designs because of the manner in which the unfactored apparent earth pressure diagrams were developed. 8.11.4 Step 6: Evaluate Individual Anchor Loads and Subgrade Reaction Force Factored horizontal anchor loads may be calculated using either the tributary area method or the hinge method as shown in Figure 8-70 for a one-level wall and in Figure 8-71 for a multi(i.e., 2 or more) level anchored wall. Both methods, as shown, assume that a hinge (i.e., zero bending moment) develops at the elevation of the base of the excavation and that the excavation subgrade acts as a support. This latter assumption is reasonable for walls that penetrate into competent materials. For walls constructed in competent materials, a reaction force, R, is assumed to be developed by the passive resistance of the soil or rock below the elevation of the base of the excavation. The wall must be embedded sufficiently deep to develop this passive resistance. In this case, the lowest anchor carries only the tributary area of the apparent pressure diagram and the reaction force is equivalent to the load from the apparent pressure diagram from the base of the excavation to the midheight between the base of the excavation and the lowest anchor (i.e., using the tributary area method in Figure 8-71, Tn computed based on loading to a distance of Hn+1/2).



Figure 8-70. Calculation of Nominal Anchor Loads for One-Level Wall. Lateral load resistance is limited below the base of the excavation in very loose to loose granular soils or soft to medium clays. In both of these material types, the wall elements must experience relatively large movements to fully develop passive resistance. The vertical wall element may become overstressed (in bending) prior to achieving these movements. When undrained conditions exist in soft to medium clays (e.g., as for a temporary excavation in these soils) with stability number NS (=γH/Su)> 4, it can be shown that the passive resistance that develops on the excavation side of the wall will not be greater than the active pressure applied on the retained side of the excavation and therefore, regardless of how long the wall element is, lateral load resistance is not developed. Referring to Figure 8-72 for a temporary wall with continuous elements in soft to medium clay, it can be noted that, to provide lateral resistance, the quantity 2Su (on passive side) must be greater than γH – 2Su (on the active side): 2 S u ( passive side) ≥ γH − 2 S u (active side) 4 S u ≥ γH ⇒ if N s =

γH Su

then

N s ≤ 4 for passive resis tan ce to develop FHWA NHI-07-071 Earth Retaining Structures


Figure 8-71. Calculation of Nominal Anchor Loads For Multi-Level Wall.

Referring to the above discussion, if Ns > 4, from a loading point of view, increasing embedment has the effect of essentially creating a longer cantilever beam fixed at the location of the lowest anchor support. Therefore, as this length increases, the bending moments in the wall would increase considerably. If this condition exists, it is generally recommended to limit the depth of wall penetration to 20 percent of the excavation depth unless deeper embedment is necessary to develop sufficient capacity to resist vertical loads, provide basal stability or groundwater cutoff, or limit ground movements. This condition would also result in greater amounts of load being shifted to the lowermost anchor supports since the soils below the elevation of the excavation base cannot provide adequate support.



Figure 8-72. Pressure Diagram for Temporary Continuous Wall in Soft to Medium Clay (after AASHTO, 2007). For the case described above, the embedded portion of the wall should be designed as a cantilever beam fixed at the lowest anchor. For design, the wall section should be selected based on the maximum bending moment evaluated, i.e., either the maximum bending moment in the exposed portion of the wall above the lowermost anchor or the calculated cantilever bending moment about the lowermost anchor. Alternatively, soil-structure interaction analyses (e.g., beam on elastic foundation) (see Section 8.12.2) may be used to design continuous beams with small toe reactions as it may be overly conservative to assume that all load is carried by the lowest anchor. The values calculated using Figures 8-70 and 8-71 (for the unfactored (nominal) anchor loads) are the horizontal component of the unfactored (nominal) anchor load per unit width of wall, Thi. The total horizontal anchor load, Th, is calculated as: Th = Thi s

(8-3)

where s is the horizontal spacing between adjacent anchors. The anchor load, T, to be used in designing the anchor bond zone is calculated as:

T= FHWA NHI-07-071 Earth Retaining Structures

Th cos θ

(8-4)


where θ is the angle of inclination of the anchor below the horizontal. component of the total anchor load, Tv, is calculated as: Tv = T sin θ

The vertical

(8-5)

These unfactored (nominal) anchor loads should be multiplied by the appropriate load factor for apparent earth pressures, i.e., γAEP = 1.35. Other sources of loading (e.g., surcharges) should be distributed to individual anchors using the same procedures noted above (e.g., tributary area) and multiplied by the appropriate load factor for the limit state being considered. 8.11.5 Step 7: Evaluate Anchor Inclination

Ground anchors are commonly installed at angles of 15 to 30 degrees below the horizontal although angles of 10 to 45 degrees are within the capabilities of most contractors. Regardless of the anchor inclination, the anchor bond zone must be developed behind potential slip surfaces and in soil or rock layers that can develop the necessary design load. Steep inclinations may be necessary to avoid underground utilities, adjacent foundations, right-of-way constraints, or weak soil or rock layers. Anchors should be installed as close to horizontal as possible to minimize vertical loads resulting from anchor lock-off loads, however grouting of anchors installed at angles less than 10 degrees is not common unless special grouting techniques are used. 8.11.6 Step 8: Select Tendon Type and Check Tensile Resistance

The prestressing steel element of the tendon (i.e., strand or bar) must be capable of safely transmitting load in the anchor bond zone to the structure without tendon breakage. The maximum proof test load shall not exceed 80 percent of guaranteed ultimate tensile strength (GUTS) of the prestressing steel bar or strand (i.e., tendon breakage resistance factor = 0.8) (see Table 11.5.6-1 in AASHTO 2007 for resistance factors for other than high strength steel). Also, the unfactored (nominal) load in each anchor shall not exceed 60 percent of GUTS of the prestressing steel bar or strand. Dimensions and strengths of bars and strands commonly used in the U.S. for highway applications are provided in Tables 8-15 and 8-16, respectively. Larger size strand tendons (i.e., strand tendons with more strands than those shown in Table 8-16) are available for applications requiring greater ground anchor design loads. Specific property values should be obtained by strand and bar providers.



Table 8-15. Properties of Prestressing Steel Bars (ASTM A722). Steel grade

Nominal diameter

Ultimate stress fpu

Nominal cross section area Aps

Ultimate strength fpu Aps (= GUTS)

(ksi)

(in.) 1 1-1/4 1-3/8 1-3/4 2-1/2 1 1-1/4 1-3/8

(ksi) 150 150 150 150 150 160 160 160

(in.2) 0.85 1.25 1.58 2.62 5.19 0.85 1.25 1.58

(kips) 127.5 187.5 237.0 400.0 780.0 136.0 200.0 252.8

150

160

Prestressing force 0.8 fpu Aps (kips) 102.0 150.0 189.6 320.0 624.0 108.8 160.0 202.3

0.7 fpu Aps (kips) 89.3 131.3 165.9 280.0 546.0 95.2 140.0 177.0

0.6 fpu Aps (kips) 76.5 112.5 142.2 240.0 468.0 81.6 120.0 151.7

Table 8-16. Properties of 0.6 in. Diameter Prestressing Steel Strands (ASTM A416, Grade 270). Number of 0.6 in. diameter strands

Cross section area

Ultimate strength (=GUTS)

Prestressing force 0.8 fpuAps

0.7 fpuAps

0.6 fpuAps

(in.2)

(kips)

(kips)

(kips)

(kips)

1

0.217

58.6

46.9

41.0

35.2

3

0.65

175.8

140.6

123.0

105.5

4

0.87

234.4

187.5

164.1

140.6

5

1.09

293.0

234.4

205.1

175.8

7

1.52

410.2

328.2

287.2

246.1

9

1.95

527.4

421.9

369.2

316.4

12

2.60

703.2

562.6

492.3

422.0

15

3.26

879.0

703.2

615.3

527.4

19

4.12

1113.4

890.7

779.4

668.0

The type and size of the anchors should be evaluated prior to design of the anchor bond zone because the required hole diameter varies as a function of the tendon size. Table 8-17 can be used to estimate the minimum trumpet opening for strand or bar tendons. FHWA NHI-07-071 Earth Retaining Structures


Table 8-17. Guidance Relationship between Tendon Size and Trumpet Opening Size (after Sabatini et. al., 1999).

Tendon type

Minimum suggested trumpet opening size (in.) Class II corrosion Class I corrosion protection protection

Number of 0.6–in. diameter strands 4 7 9 11 13 17 Bar diameter (in.) 1 1-1/4 1-3/8

4 4-1/2 5 5-1/2 6 6-1/2

6 6-1/2 7 7-1/2 8 8-1/2

2-1/2 2-3/4 3

3-1/2 3-3/4 4

8.11.7 Step 9: Evaluate Anchor Bond Length

Estimates of load transfer in the anchor bond length are typically based on previous field experience. When estimating resistance using previous field results, potential variations in resistance due to differing installation and grouting methods must be considered. In a given soil deposit, the actual resistance achieved in the field will depend on the method of drilling including quality of drill hole cleaning and period of time that the drill hole is left open, the diameter of the drill hole, the method and pressure used in grouting, and the length of the anchor bond zone. Except for certain minimum values, the selection of these items should be left to the discretion of the specialty anchor contractor. The main responsibility for the designer is to define a minimum anchor resistance that can be achieved in a given ground type. Because of the similarity of many projects, some fairly typical anchor characteristics can be summarized. These are intended to provide a range of typical design values to engineers who are unfamiliar with anchor design. •

Maximum Test Load Between 60 and 240 kips: Anchor tendons to resist these loads can be handled without the need for unusually heavy or specialized equipment. In addition, stressing equipment can be handled by one or two workers without the aid of mechanical lifting equipment. The drill hole diameter is generally 6 in. or less,



except for hollow stem augered anchors that are typically approximately 12 in. in diameter. •

Total Anchor Length Between 30 and 60 ft: Because of geotechnical or geometrical requirements, few anchors for walls (or for tiedown structures) are less than 30 ft long. A minimum unbonded length of 10 ft for bar tendons and 15 ft for strand tendons should be used. These minimum unbonded lengths are required to avoid unacceptable load reduction resulting from seating losses during load transfer and prestress losses due to creep in the prestressing steel or the soil.

Soil Anchors Anchor bond lengths for gravity-grouted, pressure-grouted, and post-grouted soil anchors are typically 15 to 40 ft since significant increases in load carrying capacity for bond lengths greater than approximately 40 ft cannot be achieved unless specialized methods are used to transfer load from the top of the anchor bond zone towards the end of the anchor. Pressure grouting in cohesionless soils significantly increases the normal stresses acting on the grout body (i.e., increases confinement). Small increases may also be observed in the effective diameter of the anchor bond zone, but capacity estimates should be based on the asdrilled hole diameter. Post-grouting is an effective means of increasing capacity in cohesive soils. Post grouting increases the radial stresses acting on the grout body and causes an irregular surface to be developed around the bond length that tends to interlock the grout and the ground. It is difficult to predict load capacity in post-grouted anchors owing to the complexity of the grouting procedures used, however, post-grouting of ground anchors in cohesive soil can increase load capacity of a straight shaft anchor by 20 to 50 percent or more per phase of post-grouting with three phases being the common limit (PTI, 1996). Rock Anchors Typical anchor bond lengths in rock range from 10 to 30 ft. The load transferred from the bond length to competent sound rock may be estimated from the rock type, however, reduced values may be recommended after input from a geologist especially if the rock mass strength is controlled by discontinuities. In an allowable stress design (ASD) approach, the maximum allowable anchor design load in competent rock was determined by multiplying the bond length by the ultimate transfer load and dividing by a factor of safety of 3.0. This relatively high value of the factor of safety (compared to a factor of safety of 2.0 for anchors in soil) was used to account for uncertainties associated with potential discontinuities in the rock mass such as joints, fractures, and clay-filled fissures. In LRFD, this difference between soil FHWA NHI-07-071 Earth Retaining Structures


and rock is evidenced by the resistance factors for rock being less than that for soil (as shown subsequently). In weak rocks such as clay shales, bond stress transfer is relatively uniform as compared to bond stress transfer in more competent rock. These weak rocks may be termed “intermediate geomaterials” and have unconfined compressive strengths defined as varying from 69 to 694 psi. Anchor Bond Length Calculation For each anchor, the minimum estimated anchor bond length, Lb, is computed as:

Lb ≥ T ϕ Qn

(8-6)

where: Lb= Anchor bond length T = Factored anchor load (calculated according to Eq. 8-4 with all appropriate load factors for applicable limit states) ϕ =Resistance factor (AASHTO (2007) Table 11.5.6-1) = 0.65 (cohesionless soil) = 0.70 (cohesive soil) = 0.50 (rock) The nominal anchor pullout resistance, Qn, is given as: Q n = π × d × τa × L b

(8-7)

where: d = Diameter of anchor drill hole τa= Nominal anchor bond stress (see Table 8-18) As described in Sabatini et al. (1999), various procedures are available for estimating the nominal resistance of ground anchors in soil and rock using semi-empirical correlations or in-situ testing. As a guide, Table 8-18 may be used to estimate the nominal bond for anchors installed in cohesive soils, cohesionless soil, and rock, respectively. The values in this table FHWA NHI-07-071 Earth Retaining Structures


are sometimes referred to as presumptive values and should be used for preliminary design only. Alternatively, PTI (2004) suggests that the nominal bond stress between rock and grout can be approximated as 10 percent of the unconfined compressive strength of the rock up to a maximum value for nominal bond stress of 450 psi. 8.11.8 Step 10: Evaluate Factored Bending Moment and Flexural Resistance of Wall

Discrete vertical wall elements must be designed to resist all applicable earth and water pressure, surcharge, anchor and seismic loadings, and the vertical component of the anchor loads and other vertical loads within the tributary area between adjacent vertical wall elements. For continuous wall elements, the analysis is performed on a per unit length basis. In designing these elements, fixed horizontal support can be assumed at each anchor level and at the bottom of the wall if the elements are sufficiently embedded below the base of the wall. If the soils at the base of the wall are relatively weak, the bending moment about the lowest anchor may be calculated assuming that the wall element acts as a cantilever below the elevation of the lowest anchor. Unless beam on elastic foundation, finite element, or other soil-structure interaction analysis methods are used, the factored bending moment in the vertical wall element for an anchored wall may be calculated using the hinge method or tributary area method (see Figures 8-73 and 8-74). For walls constructed in competent soils such as most sands and stiff clays, the maximum (factored) bending moment, Mmax, occurs in the exposed portion of the wall. For walls that penetrate deep deposits of weak material, the maximum bending moment may occur in the embedded portion of the wall. The embedded portion of the wall refers to the length of wall that is below the base of the excavation. The negative bending moment calculated at the location of the uppermost anchor is evaluated by summing moments about the first anchor location. The vertical wall elements are commonly assumed to be continuous between each vertical support location. The maximum positive bending moment between each ground anchor is, for the tributary area method, assumed equal to 1/10 p l 2 where p is the maximum ordinate of the apparent pressure envelope (with a load factor of 1.35) and l is the vertical spacing between adjacent anchors. For the hinge method, the maximum positive bending moment between each ground anchor corresponds to the point of zero shear. Additional information on computations using the hinge method is provided in Weatherby (1998). These methods provide conservative estimates of the calculated bending moments, but may not accurately predict the specific location. For continuous sheet-pile walls, the bending moment per unit of wall is used to



120 - 200

30 – 120

120 - 250

100 - 120

30 - 155

25 - 35

Slates and hard shales

Soft shales

Sandstones

Weathered Sandstones

Chalk

Weathered Marl

5 - 15

• Stiff clay, med. to high plasticity

40 - 55

20 - 50

15 - 35

10 - 25

5 - 10

• Silty clay

• Very stiff clay, med. to high plasticity • Stiff clay, med. plasticity • Very stiff clay, med. plasticity • Very stiff sandy silt, med. plasticity

5 - 10

• Soft silty clay

Cohesive Soil Anchor type Average ultimate bond stress (psi) Gravity-grouted anchors 5 - 10 (straight shaft) Pressure-grouted anchors (straight shaft)

• Sandy gravel, med. dense-dense • Sandy gravel, densevery dense

• Dense glacial till

• Silty sands



40 - 200

31 - 200

45 - 75

25 - 60

Cohesionless Soil Anchor type Average ultimate bond stress (psi) Gravity-grouted anchors 10 - 20 (straight shaft) Pressure-grouted anchors (straight shaft) • Fine-med. sand, 12 - 55 med. dense – dense • Med.–coarse sand 15 - 95 (w/gravel), med. dense • Med.–coarse sand 35 - 140 (w/gravel), dense very dense

Note: Actual values for pressure-grouted anchors depend on the ability to develop pressures in each soil type.

200 - 400

150 - 200

Soft limestone

Concrete

200 - 300

Dolomitic limestone

Average ultimate bond stress (psi)

250 - 450

Rock

Granite and basalt

Rock type

Table 8-18. Presumptive (Nominal) Bond Stress for Ground/grout Interface along Anchor Bond Zone (after PTI, 1996).

Figure 8-73. Calculation of Wall Bending Moments Using Hinge Method (after Sabatini et al., 1999). select an appropriate sheet-pile section. To evaluate the maximum bending moment for design of a soldier pile, the maximum bending moment per unit of wall calculated from Figure 8-73 and 8-74 is multiplied by the tributary length of each wall element. Soil-structure interaction analyses may also be used for evaluating wall bending moments (see Section 8.12.2.2). When designing permanent anchored walls in relatively uniform competent materials, it is usually only necessary to check the final stage of construction provided that: (1) the ground can develop adequate passive resistance below the elevation of the base of the wall; (2) apparent earth pressure diagrams are used to assess the loading on the wall; and (3) there is minimal over excavation below each anchor level (Weatherby, 1998). For cases where there are large concentrated surcharges or berms at the ground surface, it is prudent to check wall bending moments for the initial cantilever stage (i.e., stage just prior to installation and lockoff of uppermost anchor). FHWA NHI-07-071 Earth Retaining Structures


Figure 8-74. Calculation of Wall Bending Moments Using Tributary Area Method (after Sabatini et al., 1999). In most cases, wall elements of an anchored wall will be subject to both flexure and axial compression loads. Axial compression loads result from vertical anchor forces, weight of lagging, and, weight of permanent facing rigidly attached to the vertical wall elements. AASHTO (2007) Section 6.9.2.2 should be used to evaluate combined axial load and flexure FHWA NHI-07-071 Earth Retaining Structures


for a steel soldier beam wall. Student Exercise 9 provides detailed computations for the structural design (i.e., evaluation of flexural and axial resistance) of a steel soldier pile. For reinforced concrete, AASHTO (2007) Section 5.7.4 is used to evaluate combined axial load and flexure. 8.11.9 Step 11: Evaluate Bearing Resistance of Vertical Wall Element

8.11.9.1 Bearing Resistance of Continuous Wall Elements The bearing resistance for continuous wall elements (e.g., slurry walls) is discussed in Section 8.12.3.3, 8.12.4.3, and 8.12.6.3. 8.11.9.2 Bearing Resistance of Discrete Wall Elements AASHTO (2007) Section 11.9.4.1 provides information on the evaluation of bearing resistance for discrete vertical wall elements. That section indicates that all vertical loads on the wall should be assumed to be carried by the embedded portion of the wall (i.e., portion of wall element below excavation subgrade). Vertical loads will generally include the vertical component of anchor lock-off loads, weight of lagging and permanent facing, and other external vertical loads. To evaluate bearing resistance, the user is directed to the section of the code on driven piles (AASHTO (2007) Section 10.7.3) or drilled shafts (AASHTO (2007) Section 10.8.3) depending on whether the elements are driven piles or drilled-in elements. Because of this reference, the user is directed to resistance factors for static analysis methods to compute nominal resistance of single piles or shafts, i.e., from 0.25 to 0.45 for driven piles and from 0.40 to 0.55 for drilled shafts. Since soldier piles will not typically be subject to dynamic analysis and/or load testing (as for deep foundation elements), the AASHTO LRFD code therefore requires resistance factors for bearing resistance of soldier piles to range from approximately 0.25 to 0.55 (i.e., the range reported above for static analysis of single piles and shafts). The corresponding ASD factor of safety value (for the lower end resistance factor) would be on the order of FS = 4.0. Because potential settlements of an anchored wall will not typically result in undesirable or unsafe performance (as compared to a similar amount of settlement for a deep foundation element used), GEC No. 4 (Sabatini et al., 1999) recommends factor of safety values of 2 to 2.5 for calculating bearing resistance of driven and drilled-in soldier piles for permanent walls.



In the absence of specific calibration, design engineers should consider adjusting resistance factors for the evaluation of bearing resistance for driven and drilled-in soldier piles for anchored walls to be consistent with a factor of safety value of 2.0. The corresponding resistance factor is 0.50. This recommendation for a resistance factor of 0.5 includes: • •

•

the vertical component of the anchor lock-off loads typically provides the largest source of vertical loading on the wall; a lock-off load is a prescribed load that is imparted to the wall and therefore, in the context of LRFD, the uncertainty in this load is relatively small resulting in an appropriate load factor of 1.0; and other sources of vertical load (which would use a maximum load factor of 1.25) include dead weight of the wall element itself, weight of lagging, and weight of facing.

Using the equations shown in Figure 4-1 and assuming that only lock-off loads impose vertical loads on the wall, results in a resistance factor of 0.5 (for a corresponding ASD FS=2.0). If other vertical loads are acting on the wall (such as weight of lagging, etc), because those loads would be factored using a load factor greater than 1.0, the corresponding resistance factor (for FS = 2.0) would be slightly greater than 0.50, therefore implying that the use of a resistance factor of 0.5 is reasonable. 8.11.10

Step 12: Evaluate Overall Stability of Anchored Wall at Service Limit State

8.11.10.1

Basal Stability

Significant basal heave and substantial increases in lateral earth pressures result when the weight of the retained soil (and surcharge loads) exceeds or approaches the soil bearing resistance at the base of the excavation. Basal heave should be checked for excavations made through soft to medium clay. Traditional methods for assessing the potential for bottom heave are based on the performance of braced excavations in soft to medium clays. Figure 8-75 shows a cut in soft clay H deep and B wide. The block of retained soil exerts a vertical pressure qapplied on strip cd equal to its weight minus the shear resistance of the soil along plane bd. The bearing capacity of a cohesive soil is equal to NcSu where Nc is the bearing capacity factor. It is noted that this analysis does need to be written in an LRFD format. The calculated FS value provides an index of likely lateral deformations (and hence potential basal instability) and is not associated with a particular strength limit state.



Based on the geometry of the failure surface, B' cannot exceed B 2 . Thus, the minimum FS is: FS =

N c Su  S 2 H  γ − u B 

   

(8-8)

The width, B', is restricted if a stiff stratum is near the bottom of the cut (Figure 8-75). For this case, B' is equal to depth D. Substituting D for B' in Equation 8-8, results in: FS =

N c Su Su   H γ −  D 

(8-9)

In relation to anchored wall designs in shallow deposits, Equation 8-9 may be used. However in moderate to deep soil deposits where the width of the excavation is very large, the contribution of the shearing resistance along the exterior of the failure block is negligible and equation 8-8 reduces to: FS =

Nc N = c γH Ns Su

(8-10)

where Ns is the stability number defined as γH/Su. The bearing capacity factor used in Equation 8-10 is affected by the height/width ratio (H/B), and the plan dimensions of the cut (B/L). Values of the bearing capacity factor, Nc, proposed by Janbu et al. (1956) for analysis of footings may be used in Equation 8-10 and these values are shown on Figure 8-75. Note from Figure 8-75 that Nc values are greater for excavations constructed in short lengths (e.g., slotted excavation) as compared to excavation of the entire length of the wall. Unless the designer specifically requires staged lengths of excavation, the design should be based on the assumption that the contractor will remove the entire length of each lift of excavation.Current practice is to use a minimum factor of safety against basal heave of 1.5 for temporary facilities in soft to medium clays. As the factor of safety decreases, loads on the lowest ground anchor increase.



B

D

B' b Su H

a HB' H

H c

d

D

B/ 2

(a) Failure planes, deep deposits of weak clay

(b) Failure plane, stiff layer below bottom of excavation

10

Square and circular B/L = 1

9

B/L = 0.5

8

Nc

H = Excavation depth B = Excavation width L = Length of excavation

7

B/L = 0, Rectangular

6 5 4 0

1

2

3

4

5

H/B (c) Bearing Capacity Factor, Nc

Figure 8-75. Analysis of Basal Stability (modified after Terzaghi et al., 1996.).

8.11.10.2

External Stability

To evaluate the overall stability of an anchored system at the Service I limit state, potential failure surfaces passing behind or through the anchors need to be checked. For walls with multiple levels of anchors, failure surfaces should be checked that pass just behind each anchor (Figure 8-76). In checking a failure surface that passes behind a level of anchors, the failure surface may cross in front of or through the anchor bond zone of another level(s) of anchors. In this case, the analysis may be amended to include a portion of the restraint force FHWA NHI-07-071 Earth Retaining Structures


from the other anchor(s). If the failure surface passes in front of an anchor bond zone, the full anchor (nominal) load can be modeled as a restraint force. If the failure surface crosses the anchor, a proportional magnitude of load assuming that anchor bond stress is distributed uniformly over the anchor bond length can be modeled. Many commercially available slope stability software packages include methods to model ground anchor restraining forces. Where external stability requirements cannot be met, the anchors may be lengthened or methods to improve anchor bond or load transfer mechanisms may be used.

Denotes potential failure surface for external stability analyses

Figure 8-76. Failure surfaces for external stability evaluations.

8.11.11 Step 13: Estimate Maximum Lateral Wall Movements and Settlements at the Service Limit State

The displacement of anchored walls is evaluated at the Service I limit state for all applicable load combinations. Because evaluations of structure displacements by LRFD are made at the Service I limit state where γ = 1.0 and φ = 1.0, methods used to estimate settlement and lateral displacement by LRFD are identical to those used in ASD. The vertical and lateral displacement of anchored walls is a complex soil-structure interaction problem, and deformation analyses can be performed using modified forms of beam on elastic foundation theory or finite element analyses. Depending on project constraints, requirements with respect to control of wall and ground movements will vary. For example, permanent anchored walls constructed in granular soils with no nearby structures pose little concern with respect to movements. Wall and ground movements, however, may be the primary design issue for a temporary excavation support system located in an urban area.



Several types of movement are associated with anchored walls. These include: (1) cantilever movements associated with installation of first anchor; (2) wall settlement associated with mobilization of end bearing; (3) elastic elongation of the anchor tendon associated with a load increase; (4) anchor yielding or load redistribution in the anchor bond zone; and (5) mass movements behind the ground anchors. The last three components of deformation result in translation of the wall and are relatively small for anchored walls constructed in competent soils. Excessive vertical settlements of the wall may induce significant lateral wall movements in addition to causing high stresses at the wall/anchor interface. Wall settlements may be minimized by installing ground anchors at flat angles (if possible), and by designing the embedded portion of the wall to carry applied axial loads. For earth retaining structures installed using top-down construction methods, numerous factors influence the amounts of wall and ground movements that may occur. These factors, identified by Clough et al. (1989) are listed below: Soil conditions & stratigraphy Support stiffness Depth of excavation Installation of supports Lagging installation Groundwater table location Excavation sequence

Construction surcharges Wall integrity Weather Preloading level Construction vibrations Dewatering Level first support placed

Support spacing Wall stiffness Installation of wall Over excavation Construction inside excavation Time excavation open Shape of excavation

Some of the above factors are not under the control of the designer, but are governed by the contractor’s selected construction procedure and on the quality of their workmanship. Nevertheless, some semi-empirical guidelines have been developed by combining field experience with analytical tools, as discussed below. Figure 8-77 presents correlations for estimating vertical soil movements behind externally supported walls. Several key points associated with this figure are as follows (Clough and O’Rourke, 1990): •

Only the basic excavation and support process has been considered. Other factors listed above have to be evaluated separately and could result in more movements.

•

Excavations in stiff to very hard clays show variable behaviors as they are influenced by the in situ horizontal stress, degree of fissuring, degree of weathering, and plasticity. Heave may also be possible for some conditions. For these materials, the dimensionless diagram in Figure 8-77(b) should be used as a conservative estimate, provided that the wall is stable and not affected by poor construction practice. In



making judgments about stiff to very hard clays, it often is valuable to refer to local construction experience. •

The family of curves shown in Figure 8-77(c) is based on numerical studies and assumes good workmanship and that cantilever deformation (i.e., lateral wall movement of the very top of the wall (above the uppermost support level)) represents a small fraction of the total movement. Curves I and II are commonly used for permanent anchored walls. Settlements increase rapidly for walls constructed in soft to medium clays where basal stability is marginal. The cantilever stage movements can be idealized assuming a point of fixity at an appropriate depth below the ground surface.

•

Use Figure 8-77(d) with caution, especially where the factor of safety against basal heave is below 1.5. In these conditions, construction variables can cause significant increases in movements.

Maximum lateral wall movements for anchored walls constructed in sands and stiff clays average approximately 0.2%H with a maximum of approximately 0.5%H where H is the height of the wall. Maximum vertical settlements behind a wall constructed in these materials average approximately 0.15%H with a maximum of approximately 0.5%H.

8.12

DESIGN OF STIFF CANTILEVERED AND ANCHORED WALLS

8.12.1 General

In general, the geotechnical (e.g., wall embedment) and structural design of stiff (e.g., slurry, secant/tangent pile, jet-grouted, and DMM) wall systems may be performed using the methods previously discussed in Section 8.9 for flexible nongravity cantilevered walls and Section 8.11 for flexible anchored walls. It is specifically noted, however, that these relatively simple methods are becoming more limited in today’s practice because the simplifying assumptions previously required to perform wall design are no longer required. Today, relatively simple soil-structure interaction analysis programs are available to model all stages of wall construction leading to better estimates of wall bending moments, shear forces, and lateral wall movements. In this section, information on the use of soil-structure interaction analysis methods (which are particularly well-suited to stiff wall systems) is described. Specific design issues relative to stiff wall systems are also summarized in this section.



Figure 8-77 a and b. Evaluation of Externally Supported Wall Movements (after Clough and O’Rourke, 1990).



Figure 8-77 c and d. Evaluation of Externally Supported Wall Movements (after Clough and O’Rourke, 1990).



8.12.2 Soil-Structure Interaction Analysis Methods

8.12.2.1 General Soil-structure interaction (SSI) methods are available and being used to model wall behavior considering various construction stages (e.g., excavation and brace/anchor installation.). In general, this is the preferable method of analysis because both the temporary construction conditions and the final design condition are considered. Example staged analyses methods include modeling the wall as a beam with springs to represent various layers of soil stiffness and strength (e.g., beam on elastic foundation) and also the finite element approach. There are a variety of finite element programs, with both two and three dimension capabilities and numerous other modeling abilities, such as dewatering, anchor representation, etc. Most of these more sophisticated methods can provide a more accurate, refined estimate of lateral bracing loads, wall moments and shears, and displacement in the surrounding soil mass, assuming that appropriate model parameters and excavation stages are used in the analyses. As a result, a more economical design can often be achieved. In most staged analysis methods, the type, spacing and stiffness of the brace or anchor can be modeled and input in the analysis. Also, the wall stiffness is input in these approaches based on the modulus of elasticity, E, and the moment of inertia, I. Fine tuning (e.g., cracked vs. uncracked section, composite, etc.) the bending stiffness value, EI, may be appropriate in some cases where a calibrated model is used with well defined soil parameters. In general, the primary influence of the EI value (within typical ranges for stiff wall systems) is on movements, not wall bending moments. However, there are other parameters (such as soil shear strength) which are typically known with less certainty that influence the strength and performance requirements for the wall system to a greater degree. 8.12.2.2 Analysis Method Herein, the so-called “p-y” analysis method is described as a viable approach to perform SSI analyses. The p-y analysis method predicts soil resistance along a pile (or wall element in this case) as a function of lateral pile deflection by treating the soil resistance as a series of non-linear springs. Common practice is to impose a lateral pressure distribution above the excavation subgrade and to perform the p-y analysis for the embedded (or passive side) portion of the wall. Distinct soil and rock layers on the passive side with distinct p-y curves can be modeled. Anchors and braces can also be treated as springs and walls with either discrete or continuous wall elements can be modeled. Commercially available computer programs such as LPILE (ENSOFT, 1999), PYWALL (ENSOFT, 1999), WALLAP (GEOSOLVE, 2002), GT STRUDL (GATECH, 2003), and others are available to perform this SSI analysis. FHWA NHI-07-071 Earth Retaining Structures


An alternative method termed the “strain wedge” method is becoming increasingly used for assessing lateral response of foundation elements. Unlike the p-y analysis approach, the effect of structural element (e.g., pile) stiffness, cross section shape, and pile head condition on the soil reaction can be modeled. In other words, the resultant soil reaction, p, is really a soil-pile reaction that depends upon the neighboring soil layers and pile properties (Ashour et al, 1998). It is noted, that although this method allows for modeling all stages of construction and thus would require the design engineer to develop an earth pressure diagram for each stage of the analysis, the method can also be used to just model a final excavation height condition. In that case, an apparent earth pressure diagram is developed to envelope the earth pressures imposed by all stages of construction. In many cases, however, stiff wall systems used for temporary support are incorporated into a final structure such as a building. In that case, a stage construction analysis is required to evaluate the stresses and strains that are locked into the final structure at the end of construction. Semi-empirical formulations have been developed for establishing p-y curves based on the results of full-scale instrumented load tests. For larger projects, the use of pressuremeter data can be used to establish site-specific p-y curves. For example, for the Marquette Interchange Project (ca. 2004) in Milwaukee, Wisconsin, 11 distinct p-y curve categories were developed based on the results of 37 site-specific pressuremeter tests for a particular segment of the project. These categories are based on soil type, consistency (or relative density), vertical effective stress (or overburden pressure) at the testing elevation, and similarity of the p-y curves. Typically, the predicted wall displacements from a SSI analysis are much more sensitive to the values used for the p-y curves than the predicted wall bending moments. For this reason, conservative selection of the parameters for the p-y curves should provide conservative estimates of ground movements without significantly increasing the structural demand of the wall (and, if used, bracing system) (Pearlman et al., 2004). Figure 8-78 shows a conceptual model for a SSI analysis. In this case, only the final excavation condition is shown. An LRFD-based procedure for designing walls using SSI is not provided in AASHTO (2007). Herein, a general sequence of analysis for a cantilevered wall (without bracing or anchors) is provided: 1. Evaluate unfactored loads (e.g., earth, surcharge, water) for the portion of the wall above the excavation level. 2. Determine factored loads at all appropriate strength and service limit states. 3. Assume a certain structural section to allow for computation of bending stiffness. FHWA NHI-07-071 Earth Retaining Structures


4. Evaluate overall stability at the Service I limit state using methods presented in Section 8.11.10. 5. Using factored axial and lateral loads for each appropriate strength limit state, perform SSI analysis. This may be done using a p-y analysis approach. 6. Based on above, compare factored loads (i.e., axial and shear) and factored bending moments from SSI analysis to factored resistances for structural section selected for all appropriate strength limit states. If unacceptable results, increase structural section (e.g., increase diameter of shaft, reduce distance between supports, etc.). 7. Evaluate service limit states and compute lateral and vertical deformations of the wall (and ground settlements behind the wall). If computed deformations exceed tolerable deformations, then modify design (e.g., increase embedment depth or structural section). 8. Repeat until all limit states have been satisfied. 9. Check applicable extreme event limit states and modify design as necessary.

TH1 Anchor Forces

Lateral Earth Pressure

TH2

Subgrade resistance Modeled with p-y curves Figure 8-78. Model for Soil-Structure Interaction Analysis.

Since the soil/rock response is linked to deformation (via a p-y curve), it is recommended that for all strength and service limit state evaluations, soil p-y curves should use a resistance factor of 1.0. This implies, however, that uncertainty in predicted resistance is not addressed. 8.12.2.3 Overall Stability and Wall Seepage The evaluation of overall stability is concerned, in part, with the evaluation of potential heave into the bottom of the excavation. Procedures described in Section 8.10.10 are used to FHWA NHI-07-071 Earth Retaining Structures


evaluate bottom heave potential in clayey soils. For excavations in sand, bottom heave is only a concern if upward seepage gradients directed towards the excavation become large enough to liquefy the sand. Figures 8-79, -80, -81, and -82 identify significant features of excavation geometry in relation to groundwater control. Seepage can enter the excavation through the walls and at the bottom of the excavation. Seepage quantities should be evaluated before construction using flow nets or other analytical techniques. Seepage through a wall should be assessed on the basis of differential head, retained soil characteristics, and effective permeability of the wall where the effective permeability of the wall is controlled primarily by defects resulting from construction. For example, a torn or jumped interlock of a steel sheet pile wall could allow an order of magnitude more seepage through the wall (and movement of fines) than indicated by laboratory or design studies of seepage through “tight” interlocks. Similarly, a local zone of honeycombed concrete could allow significant seepage compared to properly placed concrete. Qualitative comparisons for various wall types are shown in Table 8-19 (ASCE, 1997). 8.12.2.4 Pile Spacing Effects For the SSI analysis, the lateral pressures above the excavation line are multiplied by the center-to-center spacing of the piles. Below the excavation, the passive resistance (as defined by the p-y curves for the below excavation ground) must be modified to account for spacing effects. The resistance of a group of closely spaced piles is less than would be calculated based on the sum of the resistances of the same number of widely spaced piles. Therefore, p-y curves may need to be adjusted by a so-called “p-multiplier”, Pm, for closely spaced piles. The selection of a p-multiplier is site specific and is generally taken as a function of the ratio of the spacing of the piles (s) to the diameter of the pile (b). For example, for the TREX project in Denver involving cantilevered drilled shaft walls (i.e., tangent pile walls), a Pm of 1.0 was used for s/b=3 and 0.5 was used for s/b=1 (Sisson et al., 2004). AASHTO (2007) Section 10.7.2.4 recommends that a Pm value less than 1.0 only be used if the pile spacing is 5b or less (e.g., use Pm = 0.7 for s/b =3). According to AASHTO (2007), p-y analysis methods apply to elements that have the ability to bend and deflect. For large diameter, relatively short elements, the element tends to rotate rather than bend in which case strain wedge theory may be more applicable (Ashour et al, 1998). FHWA NHI-07-071 Earth Retaining Structures


Figure 8-79. Significant Features of Excavation Geometry in Relation to Groundwater Control (after ASCE, 1997). FHWA NHI-07-071 Earth Retaining Structures








Table 8-19. Cutoff Functions and Watertightness of Excavation Walls (after ASCE, 1997). Type Of Wall

Potential Leakage Through Potential for Piping beneath Wall Wall Wood, concrete, or steel Deflection opens joints between To avoid piping, penetration below cantilever sheeting of sheets. Leakage typically 0.1 to 4 subgrade must exceed ¼ of exterior low height gpm per 100 linear ft for low head where there is limited depth of head. Also, depends on joint pervious layer or more than ½ of head for deep pervious layer. detail. Braced, interlocked steel Leakage typically 1 to 10 gpm per As above, but in potentially sheeting 100 linear ft for low head. Lesser “running” soil (i.e., non-plastic silt, quantity of leakage if movement silty fine sand, or narrowly graded is minimized, locks are filled and sand), piping may occur in a path tensile stress acts along the wall along the face of the sheeting. in longitudinal direction. Steel soldier piles with No impediment to leakage; Not intended for cutoff below wood or concrete exterior drawdown and water subgrade. If exterior head is not reduced by leakage, lagging pressure in active wedge of greatly retained soil can be analyzed by dewatering to control uplift may be flow net. Inflow depends on soil necessary. permeability and recharge. Concrete cylinder piles: Equivalent permeability of wall is Could accomplish partial or complete tangent, secant, or typically 5 x 10-4 to 1 x 10-5 cm/s. cutoff below subgrade, but usually Layout staggered Leakage typically 0.5 to 8 gpm not seated in rock. per 100 linear ft per 10 ft of head convenient to work around shallow depending principally on quality obstructions. of joints between cylinders. Slurry trench concrete Equivalent permeability of wall is Cutoff enhanced by chiseling into wall typically 2 x 10-4 to 1 x 10-6 cm/s. underlying rock or by pre-positioning Leakage typically 0.1 to 4 gpm grout pipe in wall elements to per 100 linear ft per 10 ft of head. facilitate grouting in strata beneath Much influenced by anchor wall. penetration through wall and joint quality. Details of penetration are important. Note: Leakage through wall is expressed as quantity of inflow per 100 ft length of wall for each 10 ft difference in head across the wall, assuming leakage is not limited by permeability of the retained soil.



8.12.2.5

Modeling the Ground Anchors

Weatherby et al. (1998) provides a detailed discussion on the use of SSI for anchored walls. In that report, a detailed description of a modeling approach for the ground anchor is provided. A summary is provided below.

T-y curves are used to model ground anchors in a SSI analysis after they have been stressed and locked off. Figure 8-83 shows how a non-linear T-y curve for a SSI analysis is developed. Since ground anchors are installed at an angle, the horizontal components of anchor load and tendon elongations are used in developing the T -y curves. The anchor lockoff load is the starting point for the ground anchor in the analysis, and the deflection associated with the lock-off load corresponds to the wall deflection from the previous construction stage (anchor stressing). If the wall moves out, the ground anchor load will increase, and if the wall moves back into the soil, the lock-off load will decrease. High ground anchor loads will move the wall back into the ground. Low anchor loads will result in the wall deflecting outward until the ground anchor load increases. The initial slope of the T-y curve is the horizontal component of the anchor tendon stiffness and it is given as: k=

As Es cos α Lu

(8-11)

where: k = anchor stiffness As = area of anchor tendon Es = Young’s modulus for the anchor tendon Lu = effective unbonded length α = anchor inclination In Equation 8-11, the effective unbonded length of the anchor tendon is assumed to be the sum of the unbonded length plus half the tendon bond length. This value is assumed to permit the T-y curve to be constructed, but the actual elastic behavior of the ground anchor will be different. Bending moments are not sensitive to changes in elastic length. If the ground anchor load changes during the analysis, wall deflections will vary depending upon the unbonded length used to construct the T-y curves. At the yield load, the T -y curve changes slope. The second portion of the anchor curve represents the ground anchor behavior between the yield and ultimate tendon strength.



Figure 8-83. Diagram Illustrating a Ground Anchor T-y Curve (after Weatherby, 1998).



8.12.3 Slurry Walls

The design procedures previously presented for soldier pile and lagging walls are generally applicable for the design of slurry walls. However, there are important differences in the design of slurry walls since: (1) they are generally continuous structures for their full depth of penetration; (2) they are considerably stiffer than soldier pile and lagging systems and are usually used where strict deformation control is required; and (3) they are typically designed for full hydrostatic water pressure behind the wall. The design considerations specific to slurry walls are discussed below. 8.12.3.1 Earth Pressure Distribution Earth pressures greater than those corresponding to active conditions may be appropriate if the wall system is highly preloaded and/or stiff and workmanship is of high quality (ASCE, 1997). Based on this, it is reasonable to assume that most slurry walls should be designed for higher than active earth pressure loadings. Since slurry walls will likely only be used for projects where strict deformation control must be maintained, it is reasonable to use earth pressures consistent with those obtained from apparent earth pressure diagrams. Slurry walls are often used in urban environments wherein different soil types may exist over the depth of the excavation. In this case, it may be difficult to select a single apparent earth pressure diagram for the entire depth of the excavation so more than one apparent pressure diagram may be used to estimate loads on the wall. Usually, unless a dewatered condition is achieved, hydrostatic water pressures would be added to the apparent earth pressure diagrams for slurry walls. Alternatively, water pressures consistent with seepage conditions under and around the wall maybe assumed. The previous discussion relates primarily to temporary loadings for slurry walls. Usually, if the slurry wall is used for a permanent system, the permanent lateral earth pressures are assumed to be consistent with at-rest earth pressures. For the permanent case, appropriate long-term water pressures are added. It is noted that when more sophisticated analyses such as finite element methods are used to analyze all stages of slurry wall construction (including excavation, support (i.e., anchor) loads, etc.), that the actual computed earth pressures are less than those from apparent earth pressure diagrams. These more sophisticated analyses are often used for large projects or where significant construction cost savings may be realized through the optimization of structural elements. Reinforced concrete slurry walls are typically analyzed as a beam spanning vertically between support levels. For Soldier-Pile-Tremie-Concrete (SPTC) walls, however, the FHWA NHI-07-071 Earth Retaining Structures


reinforced concrete wall is typically analyzed as a beam spanning horizontally between the steel piles, and the steel piles are used to transmit the applied loads vertically to the support levels. 8.12.3.3 Support of Vertical Loads In many cases, slurry walls are incorporated into a permanent structure and must be designed to support vertical loads. These loads result from vertical anchor forces as well as permanent vertical loads (such as from a building if the slurry wall is a basement wall). For design, these vertical forces may be carried, at least partially, above the excavated subgrade, however since slurry walls are often used in relatively weak soils, it is conservative to assume that all load is carried by the excavation subgrade. It is important, however, to consider the specific ground conditions for the site to evaluate the conservatism of this assumption. The portion of the load carried below the excavation subgrade is resisted by skin friction over two sides of the panel and through end bearing. Because the width of the base of a slurry wall is relatively large (e.g., 2 ft or greater), a relatively significant amount of vertical movement would be required to mobilize end bearing. For ASD, a reasonable assumption is that no more than 20 percent of axial resistance is achieved through end bearing with the remainder of resistance generated in skin friction. A similar distribution of resistance should be assumed for LRFD-based computations. 8.12.4 Tangent and Secant Pile Walls

8.12.4.1 Earth Pressure Distribution The earth pressure distribution diagrams, types of wall support, methods of analysis for design of tangent pile and secant pile walls are the same as those discussed for slurry walls. 8.12.4.2 Methods of Analysis When a steel reinforcing cage or steel structural section is placed in all of the bored piles, the design load for each pile is determined by multiplying the appropriate soil and water pressure diagrams by the center-to-center spacing of the piles. When only alternate piles are reinforced, the design load for the reinforced piles is determined by multiplying the appropriate pressure diagrams by the center-to-center spacing of the reinforced piles. 8.12.4.3 Support of Vertical Loads Procedures used for the axial resistance evaluation of drilled shafts (see AASHTO Section 10.8.3) may be used to calculate required embedment depths for tangent/secant pile walls FHWA NHI-07-071 Earth Retaining Structures


designed to carry vertical forces. In general, an analysis of a single shaft is performed and this capacity is reduced to account for overlapping between shafts and group effects resulting from very closely spaced piles. The recommendations provided in Section 8.11.9.2 of this manual may be used for LRFD-based evaluations. For intermittent pile walls, individual capacities can be reduced to account for group effects in the same way that the capacity of a drilled shaft group is reduced from that of the sum of the capacities of individual shafts. In cases where these walls (or other stiff walls systems) support permanent vertical loads, site-specific load testing may be performed. In that case, resistance factors consistent with the level of static load testing performed can be used for bearing resistance design. 8.12.5 Jet Grouted Walls

For a jet-grouted wall, steel reinforcements can be inserted into the wall elements before the soil-grout mix hardens. The resulting structural combination can resist lateral earth pressures, and when the wall is properly supported laterally (e.g., by ground anchors) it can support excavations of considerable depth. The overlapped columns can also be used as a cut-off wall for seepage control. The jet-grouted earth wall is designed based on conventional wall design procedures. The width of the wall is usually defined in terms of column diameter, configuration of overlapped columns, and minimum overlapping length. Jetting parameters, which achieve the required width, strength and permeability of the wall, are then established based on existing soil and groundwater conditions. The characteristics of the jet-grouted soil (strength, permeability, etc.) are influenced by the properties of the in situ material, the composition of the grout mix, and a number of operating parameters, such as injection pressure, flow rate, withdrawal rate (lifting speed), rotation rate, etc. The design of a jet-grouted column is influenced by a number of interdependent variables related to in situ soil conditions, materials used, and operating parameters. Table 8-20 presents a summary of the principal variables of the jet grouting system and their potential impact on the three basic design aspects of the jet-grouted wall: column diameter, strength and permeability. Table 8-21 gives typical ranges of operating parameters and results achieved by the three basic injection systems of jet grouting. It should be noted, however, that the grout pressures indicated in Table 8-21 are based on certain equipment and can vary. Table 8-21 can be used in feasibility studies and preliminary design of jet-grouted wall systems. The actual operating parameters used in production are usually determined from initial field trials performed at the beginning of construction. FHWA NHI-07-071 Earth Retaining Structures


Table 8-20. Summary of System Variables and Their Impact on Basic Design Elements. Principal Variables

General Effect of the Variable on Basic Design Elements (Strength, Permeability and Column Diameter)

(a) Jet-Grouted Soil Strength Degree of mixing of soil and grout Soil type and gradation Cement Factor Water/cement ratio of grouted mass Jet grouting system Grout composition Age of grouted mass

Strength is higher and less variable for higher degree of mixing. Sands and gravels tend to produce stronger material while clays and silts tend to produce weaker material. Strength increases with an increase in cement factor (weight of cement per volume of jet-grouted mass.) Strength of the jet-grouted soil mass decreases with increase in in-situ water/cement ratio. The strength of the double fluid system may be reduced due to air entrapment in the soil-grout mix. Richer grouts produce higher strengths than leaner grouts. As the jet-grouted soil mass cures, the strength increases but usually at a slower rate than that of concrete.

(b) Wall Permeability

Wall continuity

Grout composition

Soil composition

Overall permeability of a jet grout wall is almost entirely contingent on the continuity of the wall between adjacent columns or panels. Plumb, overlapping multiple rows of columns would produce lower overall permeability. In case of obstructions (boulders, utilities, etc.) if complete encapsulations is not achieved then overall permeability may be reduced due to possible leakage along the obstruction-grout interfaces. Assuming complete wall continuity and complete replacement of in situ soil, the lowest permeability which can be obtained is that of the grout (typically 10-6 to 10-7 cm/sec). Lower permeabilites may be possible if bentonite or similar waterproofing additive is used. If complete replacement is obtained (as may be possible with a triple fluid system) then soil composition does not matter. Otherwise, if uniform mixing is achieved then finer grained soils would produce lower permeabilities as compared to granular soils.

(c) Column Diameter Jet grouting system Soil density and gradation Degree of mixing of soil and grout

The diameter of the completed column increases in size as the number of fluids is increased from the single to the triple fluid systems. As density increases, column diameter reduces. For granular soils, the diameter increases with reducing uniformity coefficient (D60/D10). Larger and more uniform diameters are possible with higher degree of mixing.

Due to the large number of variables involved in the jet grouting process, it is very difficult to establish a rigorous design process. A typical design procedure is dependent to a great extent on the field and laboratory test results, the specific equipment and method used, and the experience of the equipment operator. The following steps are followed in the design of a jet-grouted wall: FHWA NHI-07-071 Earth Retaining Structures


Table 8-21. Typical Range Of Jet Grouting Parameters And Jet-Grouted Soil Properties (adapted from Kauschinger and Welsh, 1989) Parameter

Units

Single Fluid

(a) Injection Pressure Water jet MPa Not used Grout jet MPa 30-55 Compressed air MPa Not used (b) Flow Rates Water jet liters/min Not used Grout Jet liters/min 60-150 3 Compressed air m /min Not used (c) Nozzles Sizes and Number Water jet mm Not used Grout jet mm 1.8-3.0 Number of water jets Not used Number of grout jets 2-6 (d) Cement Grout w/c ratio 0.8 - 2.1 3 Cement Consumption kg/m 400-1000 (e) Drill Rod Rotation and Lifting Speed Rotation speed rpm 10-30 Lifting Speed min/m 3-8 (f) Column Diameter Coarse-grained soil m 0.5-1 Fine-grained soil m 0.4-0.8 (g) Jet-Grouted Soil Strength Sandy soil MPa 10-30 Clayey soil MPa 1.5-10

Double Fluid

Triple Fluid

Not used 30-55 0.7-1.7

30-55 1-4a 0.7-1.7

Not used 100-150 1-3

70-100 150-250 1-3

Not used 2.4-3.4 Not used 1-2

1.8-2.6 3.5-6.0 1-2 1

0.8 - 2.1 150-550

0.8 -2.1 150-650

10-30 3-10

3-8 10-25

1-2 1-1.5

1.5-3 1-2

7.5-15 1.5-5

10-20 1.5-7.5

Note: aHigher grout pressures of 10-18 MPa are used with certain equipment where large separation exists between the water/air nozzle and the grout nozzle. 1 m = 3.3 ft, 1 MPa = 145 psi, 1 kg/m3 = 0.062 pcf, 1 liters = 0.26 gallons.

1. Establish earth pressure diagrams. For determining the lateral earth pressures, the jetgrouted wall may be considered as a gravity structure due to its large mass. For thinner, externally supported and reinforced jet-grouted walls subject to bending, the earth pressures acting on the wall may be similar to those of a slurry diaphragm wall. FHWA NHI-07-071 Earth Retaining Structures


2. Perform stability and wall seepage analyses and determine wall thickness based on assumed strength, permeability, column diameter and overlapping length. Use Table 8-21 and 8-22 in establishing these assumptions. 3. Conduct a field trial, coupled with laboratory tests, to confirm the design assumptions and select the parameters to use for production.

Table 8-22. Range of Typical Soilcrete Strengths (Three Fluid System) (after Elias et al., 2006). Soil Type Clean Sands & Gravels Silty Sands Sandy Silts & Clayey Sands Silts and Low Plasticity Clays

Soilcrete Unconfined Compressive Strength (psi) 1,000 – 2,000 600 – 1,450 350 – 1,000 300 – 1,000

8.12.6 DMM Walls

8.12.6.1 General The design of a DMM wall is influenced, to a certain extent, by its planned function. When used as a cut-off wall, for instance, the wall is designed to provide a required permeability. When used for excavation support, on the other hand, vertical reinforcing members are usually provided to resist bending moments as well as shear stresses along the longitudinal direction of the wall. Between the reinforcements, the soil-cement structure is designed to resist and redistribute the horizontal stresses to neighboring reinforcing members. Table 8-23 summarizes the range of anticipated engineering parameters that can be achieved by deep mixing techniques. Note that the wide range of reported improved properties is a function of the in-situ soils, reagent content, equipment used for mixing, and the in-situ water and organic content. Additional information on engineering properties of wet (and dry) mix treated soils is provided in Elias et al. (2006).



Table 8-23. Typical Improved Engineering Characteristics of Soils Treated with DMM (wet mix) (after Elias et al., 2006) Property Unconfined Compressive Strength, qu

Permeability, k Secant Modulus at 50% qu, E50

Tensile Strength (wet mix) Poisson’s Ratio

Typical Range Up to 175 psi (organic and very plastic clays, sludges) 60 – 220 psi (soft clays) 100 – 360 psi (medium to hard clays) 145 – 435 psi (silts) 220 – 725 psi (fine to medium sands) 10-4 to 10-7 cm/s 50 – 1,000 qu for lab samples (Japan) 50 -150 qu for lab samples (U.S) 100 – 300 qu (typical range) 8 – 14 percent of qu 0.19 – 0.45 (typical value of 0.26)

8.12.6.2 Factors Affecting Design Strength, permeability and modulus of elasticity are the primary engineering properties of concern for the DMM wall. These properties are affected by the soil type, cement and other admixture proportions, water/cement ratio of the grout, degree of soil-cement mixing, curing environment and age. Based on actual wall projects in various soil types, Taki and Yang (1991) presented design guidelines, many of which are included in the following sections. The strength is commonly expressed in terms of unconfined compressive strength, qu. It is primarily influenced by the soil type and cement content. In general, an increase in cement content increases the unconfined compressive strength. Higher strength increases are observed for sandy or gravelly soils as compared to cohesive soils where the strength gain with increasing cement content may be minimal. For a given cement proportion, the strength increases with age. The 28-day strength is typically 1.4 to 1.5 times the 7-day strength for clays, and 2 times the 7-day strength for sands. The 56-day unconfined compressive strength may be 1.5 times to 28-day strength. Strength continues to increase with time for perhaps 6 months. For design purposes, one third of the unconfined compressive strength can be considered as the shear strength, τf, of the soil-cement, i.e.:



1 3

τ f = qu ( 28)

(8-12)

For a well mixed soil-cement, the coefficient of permeability ranges from 10-5 to 10-7 cm/sec. In general, the coefficient of permeability and the porosity of the soil-cement matrix decrease with decreasing the sand content and the water-cement ratio, and increasing the curing age. It should be noted, however, that the above indicated coefficients of permeability are those of the soil-cement mix and do not necessarily represent the overall permeability of the DMM wall. This permeability is influenced by the tightness of the wall and the potential presence of “windows” which may be caused by boulders or column deviations. The overall permeability of the DMM wall is best determined by performing in situ permeability tests in the completed wall. 8.12.6.3 Design Considerations The structural design of DMM walls includes: (a) the design of reinforcing members to resist bending moments, shear stresses and deflections along the height of the wall; (b) the design of soil-cement elements to resist and redistribute the lateral pressures to the reinforcing members; and (c) the determination of the minimum depth of embedment required for base stability or vertical load support. The design of reinforcing members and the required depth of embedment are carried out as in the case of conventional soldier pile and lagging walls. The lateral earth pressures are similar to those discussed for slurry walls. The design information presented herein is for informational purposes only and has not, to date, been calibrated for a LRFD format. The design of the soil-cement lagging proceeds by considering it as a horizontal member supported on two adjacent reinforcing members. If a reinforcing member is installed in every column, as shown in Figure 8-84(a), it is only necessary to consider the punch-through shear force Q in calculating the shear stresses. Where reinforcement is not installed in every soil-cement column, the soil-cement element may be analyzed using the hypothetical model shown in Figure 8-84(b). In this case, in addition to the punch-through shear stresses, the analysis must consider compressive stresses along a hypothetical parabolic arch with a configuration as shown. In both cases, although the resistance to the punching shear (Q) is calculated at the point of maximum shear along the edge of the reinforcing member (H-pile), it should also be checked at the necking point between columns. Although this point has lesser shear, it also has the minimum section and should be checked.



An empirical design guideline for spacing reinforcing members to avoid bending failure in the soil-cement element is presented in Figure 8-84(c). This guideline was based on finite element simulation and full-scale model tests to study various failure modes of soil-cement between reinforcing members (Taki and Yang, 1991).

wall loading

d

Q

(a)

l2 wall loading

(b)

Center of H-pile

D h e 45º

Center of DMM Column

L2 If L2 ≤ D + h -2e, no bending failure (c) Figure 8-84. DMM Wall Design: (a) Analysis for Punch-Through Shear; (b) Analysis for Compressive Action of Arching Effects; (c) Empirical Guideline for Avoiding Bending Failure (after Taki and Yang, 1991)



For DMM walls, axial resistance of steel reinforced columns needs to consider the potential for the steel section to punch through resulting from insufficient load transfer from the steel section to the soil cement column. If sufficient load transfer can be assured, the entire DMM can be designed to carry axial loads in much the same way as tangent/secant piles, i.e., development of skin friction and end bearing (over diameter of soil cement column) modified to account for group effects.

8.13

DRAINAGE SYSTEMS FOR CUT WALLS

The need for drainage in cut wall system applications varies with project requirements. Drainage systems may be omitted in cases where groundwater drawdown in the retained soil is prohibited or undesirable. In other cases, drainage is used as a means to control surfacewater infiltration and groundwater seepage. Other beneficial effects of drainage include: • • • •

allowing excavations to be constructed in relatively dry conditions; reducing seepage so that instabilities such as soil heave at the base of the excavation, soil softening due to saturation, or piping of materials at the wall face is minimized; reducing water levels which increases passive resistance of the soils below the excavation depth; and enabling anchors and soil nails to be installed without the use of special drilling and grouting techniques (ASCE, 1994).

For cut wall systems, collection of subsurface flow is usually achieved with prefabricated drainage elements (i.e., geocomposites) placed between the wall and the permanent facing. With this type of system, vertical drainage strips are extended over the full height of the wall. Single strips can be placed at appropriate horizontal spacing along the wall or a continuous sheet can be placed over the entire wall face, depending on the project drainage requirements and the expected flow rate. Water intercepted in the drainage elements flows downward to the base of the wall where it is conveyed through the permanent facing in longitudinal/outlet pipes or weepholes. In other applications, the drainage elements are extended into the subgrade to a footing drain. Similar drainage systems may be used for face drainage of soilnailed walls where the prefabricated drainage elements are typically placed between the excavated face and the back of the shotcrete facing layer. In applications where subsurface flow rates are large, horizontal drains may be used to remove water from behind the wall. A horizontal drain is a small diameter perforated pipe that is advanced into a nearly horizontal drill hole in an existing slope. For example, a cut wall constructed on or at the base of a steep slope will likely interfere with pre-existing FHWA NHI-07-071 Earth Retaining Structures


natural drainage paths. This interference may cause hydrostatic pressures resulting from trapped water to build-up against the wall. To relieve these pressures, horizontal drains can be installed at appropriate vertical and horizontal spacing along the wall alignment. For wall system applications, horizontal drains extend back from the wall face a sufficient distance to intercept subsurface flow before it enters the retained soil mass. General surface water drainage is discussed in Section 5.5.8.3 in Chapter 5. For cut wall, dikes can be constructed on the ground surface near the top of the wall or the vertical wall element can be extended above the ground surface grade to minimize surface water that can enter the excavation during construction and weaken the soils inside the excavation (Sabatini et al., 1997).





CHAPTER 9 IN-SITU REINFORCED WALLS 9.1

INTRODUCTION

In-situ reinforced walls are constructed from the top-down to support temporary and permanent excavations. Construction of these walls involves insertion of reinforcing elements in the in-situ soils to create a composite earth structure. This chapter presents information on two different types of in-situ walls: (1) soil nail walls and (2) micropile walls (a less common wall type). The main components of an in-situ reinforced wall are the in-situ material, the reinforcing inclusions, and the wall facing. The reinforcing inclusions typically consist of metal bars (soil nail walls) and small diameter steel pipe piles (micropile walls). Shotcrete, welded-wire mesh, cast-in-place concrete, or precast concrete panels are typically used for the facing. Insitu reinforced walls can be used in a wide range of ground conditions. In-situ reinforced earth walls have been successfully used for a variety of applications including: • • • • • •

Temporary and permanent walls for excavations in urban areas; Cut-slope retention for roadway widening and construction of depressed roadways; Support of existing bridge abutments for roadway widening projects; Support of cuts at tunnel portals; Landslide protection and stabilization of natural slopes; and Repair or reconstruction of existing retaining walls.

9.2

SOIL NAIL WALLS

9.2.1

General

Soil nail walls are one of the in-situ reinforced walls that are constructed to support temporary and permanent excavations (Figure 9-1). This chapter presents information on feasibility, construction materials and methods, cost, design, load testing, and construction inspection of soil nail walls. A more comprehensive coverage of soil nail walls is provided in Geotechnical Engineering Circular No. 7 “Soil Nail Walls” (Lazarte et al., 2003).


9-1

9 – Soil Nail and Micropile Walls June 2008

Figure 9-1. Soil Nail Wall Application (a) Temporary Shoring; (b) Roadway Widening Under Existing Bridge; (c) Slope Stabilization; and (d) Roadway Cut (after Porterfield et al., 1994)


9-2


Soil nail walls use horizontal to subhorizontal reinforcements to improve the shearing resistance of the soil. The reinforcements, known as nails, are closely spaced (i.e., typical spacing of 5 ft horizontal and vertical) and unlike ground anchors, are not prestressed. Shear stresses in the ground are transferred as tensile forces in the nails through friction (or adhesion) mobilized at the ground-nail interface. The nails develop tension as the retained ground moves towards the excavation as the excavation progresses. The nails may also develop some bending and shear forces; however these effects are considered secondary and are not included in the recommended design method presented later in this chapter. Nails can be installed by (1) driving; (2) drilling and grouting; (3) jet grouting; and (4) launching. Conventional U.S. practice includes constructing a drill hole in which the bar is placed and then grouted under gravity (Figure 9-2). Driven soil nails have been more commonly used in France and Germany than in the U.S. and are typically used for temporary applications. Jet-grouted and launched nails are not common in the United States however launched nails have been used for a small number of temporary slope stabilization projects. Also, so called “hollow-core” soil nails are becoming increasingly used. In this method, drilling and grouting of the nail is accomplished in one step (i.e., Steps 2 and 3 shown in Figure 9-2 are combined into one step). The components of a soil nail wall installed using the techniques listed above vary from one technique to another. Although the concept of soil nail walls is not proprietary, several specialized components or procedures are under U.S. or international patents. Patented components include, but are not limited to, some threaded bars, corrosion-protection systems, and nail installation systems (e.g., self-drilling, jet-grouted, and launched soil nails). Hollow bars are proprietary and have been used primarily for temporary walls because of concerns regarding the consistency of the corrosion protection; however, research is on-going with a focus towards identifying uses of hollow bars for permanent applications. Common applications for soil nail walls include: (1) temporary shoring for construction of CIP walls or other permanent facilities; (2) a permanent wall for roadway widening at a bridge abutment; (3) a permanent retaining wall for a roadway cut; and (4) slope stabilization. Examples of each of these applications are shown in Figure 9-3. 9.2.2

Feasibility of Soil Nails

The feasibility evaluation of a soil nail wall should encompass technical and economical considerations and include: (1) an evaluation of the prevailing ground conditions; (2) an assessment of the advantages and disadvantages of a soil nail wall for the particular application being considered; and (3) comparison with alternative systems (e.g., ground FHWA NHI-07-071 Earth Retaining Structures

9-3


anchor wall systems). The following sections present a discussion of these aspects of the feasibility evaluation.

Figure 9-2. Typical Nail Wall Construction Sequence (after Porterfield et al., 1994).


9-4


(d)

(c)


9-5


Figure 9-3. Soil Nail Wall Applications (a) Temporary Shoring, (b) Roadway Widening Under Existing Bridge, (c) Roadway Cut, and (d) Slope Stabilization.

(b)

(a)

9.2.2.1 Evaluation of Ground Conditions for Soil Nail Walls Soil nail walls can be used for a wide range of soil types and conditions. Project experience has shown that certain favorable ground conditions make soil nailing cost effective over other techniques. Conversely, certain soil conditions can be considered marginal for soil nailing applications and may make the use of soil nails too costly when compared with other techniques. Soil nail walls can generally be constructed without complications in a mixed stratigraphy, as long as the individual layers of the soil profile consist of suitable materials. The following two sections present the soil conditions that are considered most and least suitable for soil nail walls. Intermediate soil conditions, for which the feasibility of soil nailing is not readily apparent, are also described. Favorable Soil Conditions for Soil Nailing Soil nail walls have been constructed successfully in various types of soils. Construction difficulties and long-term complications can generally be avoided when specific favorable soil conditions prevail. Soil nailing has proven economically attractive and technically feasible when: •

the soil in which the excavation is constructed is able to stand unsupported in a 3- to 6-ft high vertical or nearly vertical cut for one to two days;

•

all soil nails within a cross section are located above the groundwater table; and

•

if the soil nails are below the groundwater table, the groundwater does not adversely affect the face of the excavation, the bond strength of the interface between the grout and the surrounding ground, or the long-term integrity of the soil nails (e.g., the chemical characteristics of the ground do not promote corrosion).

Although not an absolute requirement, it is advantageous that the ground conditions allow drillholes to be advanced without the use of drill casings and for the drillhole to be unsupported for a few hours until the nail bars are installed and the drillhole is grouted. Alternatively, soil nails have been installed with success using the hollow-stem drilling method in fully and temporarily cased drillholes. It is important to note that the selection of the drilling method is typically left to the discretion of the soil nail installation contractor. Soil conditions are presumed to be favorable for the construction of soil nail walls when results from field tests indicate competent soils. The Standard Penetration Test (SPT, see next chapter) provides the SPT value, N, which can be used to preliminarily identify favorable soil conditions. FHWA NHI-07-071 Earth Retaining Structures

9-6


Based on the general criteria for favorable conditions noted above, the following ground types are generally considered well suited for soil nailing applications. •

Stiff to hard fine-grained soils. Fine-grained (or cohesive) soils may include stiff to hard clays, clayey silts, silty clays, sandy clays, sandy silts, and combinations thereof. Fine-grained soils can be tentatively classified as stiff if they have SPT N-values of at least 9 blows/ft. However, the consistency characterization of fine-grained soils should not rely solely on SPT N-values. Instead, the consistency (and thereby shear strength) characterization should be supplemented with other field and/or laboratory testing. To minimize potential long-term lateral displacements of the soil nail wall, fine-grained soils should be of relatively low plasticity (i.e., in general, plasticity index (PI) < 15).

•

Dense to very dense granular soils with some apparent cohesion. These soils include sand and gravel with SPT N-values larger than 30 (Terzaghi et al., 1996), and with some fines (typically no more than about 10 to 15 percent of fines) or with weak natural cementation that provide cohesion. Capillary forces in moist fine sands may also provide an apparent cohesion. To avoid excessive breakage of capillary forces and thereby significant reduction of this apparent cohesion, the movement of water toward the excavation face needs to be minimized including by redirecting surface water away from the excavation face.

•

Weathered rock with no weakness planes. Weathered rock may provide a suitable supporting material for soil nails as long as weakness planes occurring in unfavorable orientations are not prevalent (e.g., weakness planes dipping into the excavation). It is also desirable that the degree of weathering be approximately uniform throughout the rock so that only one drilling and installation method will be required. Conversely, a highly variable degree of rock weathering at a site may require changes in drilling equipment and/or installation techniques and thereby cause a costly and prolonged soil nail installation.

•

Glacial soils. Glacial outwash and glacial till materials are typically suitable for soil nailing applications as these soils are typically dense, well-graded granular materials with a limited amount of fines.

Unfavorable or Difficult Soil Conditions for Soil Nailing Examples of unfavorable soil types and ground conditions are provided below: •

Dry, poorly graded cohesionless soils. When poorly graded cohesionless soils are completely dry, contain no fines, or do not exhibit any natural cementation, apparent


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cohesion is not available. Therefore, the required vertical or nearly vertical cuts are difficult to achieve. •

Soils with high groundwater. Perched groundwater occurring behind the proposed soil nail wall will require significant drainage, which is necessary to stabilize the mass of soil in this location. Additionally, large amounts of groundwater can cause drillholes (particularly in loose granular soils) to collapse easily, thus requiring a costly soil nail installation. Excessive groundwater seeping out to the excavation face may cause significant difficulties for shotcrete application.

•

Soils with cobbles and boulders. A large proportion of cobbles and boulders present in the soil may cause excessive difficulties for drilling and may lead to significant construction costs and delays. When only a few boulders and cobbles are present, modifying the drilling orientation from place to place may minimize or eliminate most of the difficult drilling. However, this approach has practical limitations when too many boulders are present.

•

Soft to very soft fine-grained soils. These soils typically have SPT N-values less than 4 and are unfavorable for soil nailing because they develop relatively low bond strengths at the nail-grout-soil interface, thereby requiring unreasonably long nail lengths to develop the required resistance. Long-term deformations (creep) of the soils may be a concern for highly plastic clays. Concerns for creep deformations are generally less critical for temporary applications. As with any retaining system constructed in a top-down manner, the potential for instability at the bottom of the excavation is high in soft fine-grained soils. Additionally, high-plasticity soils may be expansive and may induce additional localized pressure on the facing due to swelling.

•

Organic soils. Some organic soils such as organic silts, organic clays and peat typically exhibit very low shear strengths and thereby low bond strengths, which causes uneconomical nail lengths. While some organic soils can exhibit acceptable shear strengths, other organic soils like fibrous peat may be highly heterogeneous and highly anisotropic. In this case, while the soil shear strength can be reasonable along some orientations, it may be significantly low along other orientations. These unfavorable orientations may have a detrimental impact on the wall stability and very long soil nails will be required. In addition, organic soils tend to be more corrosive than inorganic soils.

•

Highly corrosive soil (cinder, slag) or groundwater. These conditions may lead to the need of providing expensive corrosion protection. These conditions are obviously more disadvantageous for permanent applications of soil nail walls.


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•

Weathered rock with unfavorable weakness planes and karst. Weathered rock with prevalent unfavorable weakness planes such as joints, fractures, shears, faults, bedding, schistosity, or cleavage may affect the drillhole stability and make grouting difficult. In addition, the presence of these discontinuities may cause the formation of potentially unstable blocks in the retained mass behind the wall during excavation. The marginal stability of blocks may rapidly deteriorate due to various factors, such as gouge in the joints, uplift and lateral hydrostatic pressures, and seepage forces. The stabilization of individual blocks may be necessary and can make this solution uneconomical when compared to conventional soil nails. In addition, grouting in rock with very large open joints or voids will be very difficult and/or expensive due to excessive grout loss. Grouting soil nails in karstic formations is not appropriate.

•

Loess. When it is dry, loess may exhibit acceptable strengths that would allow economical installation of soil nails. However, when sizable amounts of water ingress behind the proposed soil nail wall, the structure of the loess may collapse and a significant loss of soil strength may take place. Therefore, the collapse potential upon wetting of these soils must be evaluated. Appropriate measures to avoid excess water migration to the soil nail area must be provided in loess exhibiting significant collapse potential. Additionally, considerably low soil shear strengths may arise for the wetted condition. In these cases, unusually long soil nail lengths may result in using conventional methods of nail installation. Regrouting (an atypical and more costly step) has been used to increase bond strengths in loess.

In addition to the difficulties described above, other aspects related to soil conditions must be considered when assessing the feasibility of soil nail walls: •

The prolonged exposure to ambient freezing temperatures may cause frost action in saturated, granular soils and silt; as a result, increased pressures will be applied to the temporary and permanent facings.

•

Repeated freeze-and-thaw cycles taking place in the soil retained by the soil nail wall may reduce the bond strength at the soil nail grout-ground interface and the adhesion between the shotcrete and the soil. To minimize these detrimental effects, a suitable protection against frost penetration and an appropriate shotcrete mix must be provided.

•

Granular soils that are very loose (N ≤ 4) and loose (4 < N ≤ 10) may undergo excessive settlement due to vibrations caused by construction equipment and traffic.

•

Loose and very loose saturated granular soil can be susceptible to liquefaction in seismically exposed regions. Several ground modification techniques (typically with


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significant associated costs) may be utilized to densify granular soils and thereby minimize these damaging effects. Despite the difficulties associated with unfavorable soil conditions described above, soil nail walls may still be built. It should be recognized that these wall systems would typically be more expensive to design and construct when compared to conventional walls in a more suitable soil. It is likely that significant extra effort and cost is needed in the design and construction of soil nail walls in these marginal conditions and that more strict long-term performance requirements might be necessary to allow soil nailing in such challenging conditions. Intermediate Soil Conditions for Soil Nailing There exists some soil conditions that are intermediate to the two conditions described previously. Examples of intermediate soil conditions are presented below: •

Engineered fill. Soil nails can be installed in engineered fill if it is a mixture of wellgraded granular material and fine-grained soil with low plasticity (typically, PI < 15).

•

Residual soils. Residual soils (i.e., those soils created from the in-place weathering of the parent rock material) may be an acceptable material for soil nailing. Similarly, lateritic soil, a highly weathered tropical soil, may be acceptable. For these types of soil, specific consideration should be given to the soil spatial variability and its ability to drain.

9.2.2.2 Advantages of Soil Nail Walls Soil nail walls exhibit numerous advantages when compared to ground anchors and alternative top-down construction techniques. Some of these advantages are described below: Construction •

requires smaller ROW than ground anchors as soil nails are typically shorter;

•

less disruptive to traffic and causes less environmental impact compared to other construction techniques;

•

provides a less congested bottom of excavation, particularly when compared to braced excavations;


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•

there is no need to embed any structural element below the bottom of excavation as with soldier beams used in ground anchor walls;

•

installation of soil nail walls is relatively rapid and uses typically less construction materials than ground anchor walls;

•

easy adjustments of nail inclination and location can be made when obstructions (e.g., cobbles or boulders, piles or underground utilities) are encountered; on the other hand, the horizontal position of ground anchors is more difficult to modify almost making adjustments in the field costly;

•

because significantly more soil nails are used than ground anchors, adjustments to the design layout of the soil nails are more easily accomplished in the field without compromising the level of safety;

•

overhead construction requirements are smaller than those for ground anchor walls because soil nail walls do not require the installation of soldier beams; this is particularly important when construction occurs under a bridge; and

•

soil nailing is advantageous at sites with remote access because smaller equipment is generally needed.

Performance •

soil nail walls are relatively flexible and can accommodate relatively large total and differential settlements;

•

measured total deflections of soil nail walls are usually within tolerable limits; and

•

soil nail walls have performed well during seismic events owing to overall system flexibility.

Cost •

soil nail walls are more economical than conventional concrete gravity walls when conventional soil nailing construction procedures are used;

•

soil nail walls are typically equivalent in cost or more cost-effective than ground anchor walls when conventional soil nailing construction procedures are used; and

•

shotcrete facing is typically less costly than the structural facing required for other wall systems.


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9.2.2.3 Limitations of Soil Nail Walls Some of the potential limitations of soil nail walls are listed below: •

soil nail walls may not be appropriate for applications where very strict deformation control is required for structures and utilities located behind the proposed wall, as the system requires some soil deformation to mobilize resistance; post tensioning of soil nails can overcome this shortcoming in most cases, but this step increases the project cost;

•

the occurrence of utilities may place restrictions on the location, inclination, and length of soil nails in the upper rows;

•

soil nail walls are not well-suited where large amounts of groundwater seeps into the excavation because of the requirement to maintain a temporary unsupported excavation face;

•

permanent soil nail walls require permanent, underground easements; and

•

construction of soil nail walls requires specialized and experienced contractors.

9.2.2.4 Comparison with Ground Anchor Wall Systems To gain further insight into the soil nail wall concept, it is useful to compare the main features of a soil nail wall with those of a ground anchor wall, which is a commonly used top-to-bottom system for retaining wall construction. Detailed information on ground anchor walls can be found in Sabatini et al. (1999). •

Installation Equipment: Ground anchor walls require a two-step process for construction. First, vertical elements (typically soldier beams) are driven or predrilled to depths below the level of the final excavation using relatively large equipment. Second, ground anchors are installed and then post-tensioned to transfer loads behind the potential failure using equipment different than that used for anchor installation. Construction using different equipment may be a sizeable portion of both the project cost and schedule in ground anchor wall projects. Soldier beams and dedicated pre-tensioning equipment are not used in soil nail walls.

•

Nail/Anchor Density: Under similar project conditions, the number of required soil nails per wall unit area is larger than the number of ground anchors per wall unit area. The use of more reinforcing elements in a soil nail wall adds a degree of redundancy that can contribute to the stability of a soil nail wall. Consequently, the failure of one reinforcing element will have a smaller effect on the stability of a soil nail wall than


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that of a ground anchor wall. Typically, only five percent of production soil nails are load tested, whereas all ground anchors are tested for acceptance. •

Load on Wall Facing: The density of soil nails implies that the facing in soil nail walls supports a smaller portion of the soil pressure due to a smaller tributary area compared to the facing in ground anchor walls, which supports a much greater soil pressure. This difference is more due to the dissimilar design approaches in the two systems rather than differences in the controlling load transfer mechanisms.

•

Load Transfer: Soil nail transfer load along the entire length of the nails, whereas ground anchors are designed to transfer load only in the anchor zone behind the potential failure surface.

•

Load Distribution: The resisting force provided by soil nails is variable along its entire length. In a ground anchor, one portion of the anchor length is unbonded while the remaining portion is bonded. The load in a ground anchor is approximately constant in the unbonded length and variable in the bonded zone.

•

Load Mobilization: Ground anchors develop tensile capacity through post-tensioning after anchor installation. Soil nails are passive and only develop tensile capacity by deformation of the surrounding soil.

•

Stability of Excavation Bottom: In ground anchor walls, soldier beams are embedded to elevations below the bottom of the excavation. The shear resistance derived from the embedded portion of the soldier beams provides additional stability of the bottom of the excavation in ground anchor walls. This favorable effect is absent in soil nail systems.

•

Deflection: Field measurements in ground anchor walls indicate that the maximum wall lateral deflection is generally at midheight of the wall. In soil nail walls, the maximum lateral deflection takes place at the top of the wall. Also, maximum wall deformations are generally greater in soil nail walls than in ground anchor walls.

9.2.3 Construction Materials and Methods This chapter presents information on construction materials and methods used for the construction of soil nail walls typically used in U.S. highway applications. Conventional U.S. practice includes the use of drilled and grouted soil nails. However, other soil nail installation methods and materials have been developed and are briefly introduced here.


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Nail installation techniques can be classified as: •

Drilled and grouted soil nails: These are approximately 4- to 8-in. diameter nail holes drilled in the foundation soils. These holes are typically spaced about 5 ft apart. Steel bars are placed and the holes are grouted. Grouted soil nails are the most commonly used soil nails and they can be used as temporary and permanent applications (provided that appropriate corrosion protection is in place).

•

Driven soil nails: These soil nails are relatively small in diameter [¾ to 1 in.] and are mechanically driven into the ground. They are usually spaced approximately 3 to 4 ft apart. The use of driven soil nails allows for a faster installation (as compared to drilled and grouted soil nails); however, this method of installation cannot provide good corrosion protection other than by sacrificial bar thickness. For this reason, driven nails are only used in the United States for temporary applications. At this time, this method is not recommended for soil nail installations for permanent soil nail walls.

•

Self-drilling soil nails: These soil nails consist of hollow bars that can be drilled and grouted in one operation. In this technique, the grout is injected through the hollow bar simultaneously with the drilling. The grout, which exits through ports located in a sacrificial drill bit, fills the annulus from the top to the bottom of the drillhole. Rotary percussive drilling techniques are used with this method. This soil nail type allows for a faster installation than that for drilled grouted nails and, unlike, driven soil nails, some level of corrosion protection with grout is provided. This system is most commonly used for temporary nails.

•

Jet-grouted soil nails: In this technique, jet grouting is performed to erode the ground and allow the hole for the nail (subsequently installed) to be advanced to the final location. The grout provides corrosion protection to the central bar. In a second step, the bars are typically installed using vibro-percussion drilling methods. At this time, this method is not recommended for soil nail installations for permanent soil nail walls.

•

Launched soil nails: In this method, bare bars are “launched” into the soil at very high speeds using a firing mechanism involving compressed air. Bars are ¾ to 1 in. in diameter and up to 25 ft in length. This technique allows for a fast installation with little impact to project site; however, it may be difficult to control the length of nail that penetrates the ground. These types of soil nails are only used for temporary nails. At this time, this method is not recommended for soil nail installations for permanent soil nail walls.


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9.2.3.1 Components of a Soil Nail Wall The remainder of this section presents a description of the main components of a typical soil nail used in the U.S. practice (Figure 9-4). PERMANENT FACING (CIP CONCRETE/SHOTCRETE) TEMPORARY FACING (SHOTCRETE)

GEOCOMPOSITE STRIP DRAIN BEARING PLATE STEEL BAR

BEARING NUT AND BEVELED WASHER

GROUT / SHOTCRETE CONTACT

HEADED STUD (TYP)

~ REINFORCEMENT

CENTRALIZER (TYP.)

GROUT

DRILLHOLE

WELDED WIRE MESH TOTAL

DRILL LENGT H NAIL B AR LE NGTH

Figure 9-4. Main Components of a Typical Soil Nail (after Porterfield et al., 1994).

Nail Bars Steel reinforcing bars used for soil nails are commonly threaded and may be either solid or hollow. Bars generally have a nominal tensile strength of 60 ksi (Grade 60) 75 ksi (or Grade 75). Common U.S. practice of soil nailing involves the use of solid steel bars of Grade 60 or 75. All steel bars must be continuous without splices or welds, straight and undamaged. They can be bare or epoxy coated or encapsulated, as required based on corrosion protection considerations. Bars with a tensile strength of Grade 95 and as high as Grade 150 may be considered for soil nailing. Bars with lower grades are preferred because they are more ductile, less susceptible to corrosion, and readily available. Grade 150 bars should not be used in conventional soil FHWA NHI-07-071 Earth Retaining Structures

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nail applications because this material tends to be more brittle, particularly under shear, and more susceptible to stress corrosion than steel with lower grades. The use of high-grade steel is more typical for ground anchors where the design loads are usually much greater than for soil nails. Bars of Grade 95 may be acceptable for soil nails as long as the steel ductility is comparable to lower-grade steels. Bars of Grade 60 or 75 should conform to the standard ASTM A615, whereas the Grade 150 steel should conform to ASTM A722. Threaded bars for typical soil nail wall applications are available in No. 6, 7, 8, 9, 10, 11, and 14 up to approximately 60 ft in length. Bars having diameters smaller than No. 8 should not be used or used with great care in applications where long bars are required (e.g., high walls) because they tend to bend excessively during handling and installation. If needed, couplers can be used to extend the length of bars in excess of 60 ft; however, soil nails in excess of this length are typically not required for most highway projects. As with solid bar nails, steel for hollow-core soil nails must meet the requirements of ASTM A615. Grade 60 is minimum requirement, however, higher grade steels are available and commonly used. Currently, two types of hollow core nails are available, the primary difference between the two being the type of thread on the exterior of the nail. One thread type is the “rope” thread or the R-thread and is the standard thread for continuously threaded drill rods. The other thread type is similar to that used on continuously threaded deformed reinforcement bars. Additional information on hollow core bars is provided in Samtani and Nowatzki (2006). Nail Head The nail head comprises two main components: the bearing-plate, hex nut, and washers; and the headed-stud (Figure 9-4). The bearing plate is made of Grade 36 (ASTM A36) steel and is typically square, 8- to 10-in. side dimension and ¾-in. thick. The purpose of the bearing plate is to distribute the force at the nail end to the temporary shotcrete facing and the ground behind the facing. The bearing plate has a central hole, which is inserted over the nail bar. Beveled washers are then placed and the nail bar is secured with a hex nut or with a spherical seat nut. Washers and nuts are steel with a grade consistent with that of the nail bar commonly of Grade 60 or 75. Nuts are tightened with a hand-wrench. The head-stud connection may consist of four headed studs that are welded near the four corners of the bearing plate to provide anchorage of the nail head into the permanent facing. For temporary walls, the bearing plate is on the outside face of the shotcrete facing.


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Grout Grout for soil nails is commonly a neat cement grout, which fills the annular space between the nail bar and the surrounding ground. In ground with potential for drillhole caving, a neat cement grout is always used. Sand-cement grout can also be used in conjunction with open hole drilling (i.e., for non-caving conditions) for economic reasons. Cement types I, II, III, or V conforming to ASTM C 150 can be used. Cement Type I (normal) is recommended for most applications. Cement Type III is grounded finer, hardens faster, and can be used when a target grout strength is required to be achieved faster than for typical project conditions. Cement Type II hardens at a slower rate, produces less heat, and is more resistant to the corrosive action of sulfates than Cement Type I. The water/cement ratio for grout used in soil nailing applications typically ranges from 0.4 to 0.5. In some cases, a stiffer grout with a slump on the order of 1½ in. may be used. The need for a stiffer grout may arise when the hollow-stem auger drilling method is used or it is desired to control leakage of grout into highly permeable granular soils or highly fractured rock. Occasionally, the stiff consistency of the grout may cause difficulties with the installation of the centralizers. In this case, the grout itself may provide sufficient support to centralize the nail bar within the drillhole. Regardless of the ability of the stiff grout to support the nail bar, centralizers should always be used to assure that a minimum grout cover around the nail bar is achieved. The characteristics of the grout have a strong influence on the ultimate bond strength at the grout-ground interface. The grout should have a minimum 3-day compressive strength of 1,500 psi and a minimum 28-day unconfined compressive strength of 3,000 psi in accordance with AASHTO T106 or ASTM C109. Admixtures are not typically required for most applications, but plasticizers can be used to improve grout workability for projects located in high-temperature climates or where project constraints dictate that the grout must be pumped over long distances. Typically, the improved workability of grout due to plasticizers can be extended up to approximately one hour. The use of air entrainment agents can improve workability and reduce cracking potential, but they cause the grout to develop a more open matrix and lose some of the chemical corrosion protection provided by cement. Therefore, its use should be approved only when other corrosion protection methods, other than grout cover, are present, or the thickness of the grout cover is increased. Some proprietary grouts contain chemicals that provide zero-volume shrinkage, which is desirable to minimize cracking and enhance bond strength. Where admixtures are being considered for use, tests should be performed to verify that the grout and bond properties of the grout are not adversely affected.


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Figure 9-5. Grout Placement (Tremie) through Pipe (after Porterfield et al., 1994). Grout is pumped shortly after the nail bar is placed in the drillhole to reduce the potential for hole squeezing or caving. In solid nail bar applications, the grout is injected by tremie methods through a grout pipe, which is previously inserted to the bottom of the drillhole, until the grout completely fills the drillhole (Figure 9-5). The grout pipe typically consists of heavy-duty plastic tubing varying between 3/8-in. and 3/4-in. outside diameter (OD). Grout pipes are removed when used as part of the installation of production nails and commonly left in place when used for soil nails that are to be load tested. Grout injection must be conducted smoothly and continuously in such a way that the space between the drillhole and the nail bar is filled completely, with no voids or gaps. The bottom of the grout pipe must remain below the grout surface at all times while grout is being pumped into the drillhole. During grouting operations, the portion of the soil nail near the back of the temporary facing may not be completely filled with grout. Because this area is the most vulnerable to corrosion, it is critical that this area be subsequently filled with shotcrete, or less commonly with a stiff grout to assure complete grout coverage. Centralizers Centralizers are devices made of polyvinyl chloride (PVC) or other synthetic materials that are installed at various locations along the length of each nail bar to ensure that a minimum thickness of grout completely covers the nail bar (Figure 9-6). They are installed at regular intervals, typically not exceeding 8 ft, along the length of the nail and at a distance of about 1.5 ft from each end of the nail. FHWA NHI-07-071 Earth Retaining Structures

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Figure 9-6. Typical PVC Centralizers Attached to a Nail Bar Prior to Nail Installation (after Porterfield et al., 1994). Wall Facing Nails are connected at the excavation surface (or slope face) to a facing system, which most commonly consists of a first-stage, temporary facing of shotcrete during construction and, a second-stage, permanent facing of CIP concrete. The purpose of the temporary facing is to support the soil exposed between the nails during excavation, provide initial connection among nails, and provide protection against erosion and sloughing of the soil at the excavation face. The purpose of the permanent facing is to provide connection among nails, a more resistant erosion protection, and an aesthetic finish. Temporary facing typically consists of shotcrete with welded wire mesh (WWM) and additional shorter reinforcement bars (referred to as waler bars) around the nail heads, which are applied after each row of nails is installed in the ground. Permanent facing is commonly constructed of CIP reinforced concrete and WWM-reinforced shotcrete. Prefabricated panels may also be used to construct the permanent wall facing, especially for projects with special aesthetic requirements or where prefabricated panels are more cost-effective.


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Drainage System Groundwater is a major concern in both the construction of soil nail retaining walls and in their long term performance. Soil nail walls are best suited to applications above the water table. Excess water at the face can result in face stability problems during construction together with an inability to apply a satisfactory shotcrete construction facing. In addition, long-term face drainage is required to prevent the generation of localized high groundwater pressures on the facing. A commonly adopted design for controlling surface runoff consists of a surface interceptor ditch, excavated along the crest of the excavation and lined with concrete applied during the shotcreting of the first excavation lift. The ditch should be contoured to drain away from the working area, with collector drain pipes installed at appropriate locations, if necessary. Where larger graded slope areas exist above the wall, installation of plastic film slope protection sheeting above the interceptor ditch provides another quick and inexpensive means of controlling surface water during construction. Similar permanent surface drainage measures are generally required to prevent surface waters from infiltrating behind the facing, or flowing over the top of the wall, during the operational life of the structure. For stepped or benched walls vegetation can also be used to inhibit infiltration and lower soil water contents by evapotranspiration. Long-term groundwater drainage measures may include the following: •

Face Drains: These typically consist of 16-in-wide prefabricated geocomposite drains that are placed in vertical strips down the excavation face, on a horizontal spacing corresponding to the nail horizontal spacing, and discharging either into a base drain or through weep holes at the bottom of the wall.

•

Shallow Drains (Weep Holes): These are typically 16-in. long, 2- to 4-in. diameter PVC pipes discharging through the face and located where heavier seepage is encountered.

•

Horizontal Drains: If it is determined that the retained soil will be subjected to groundwater pressures and there is no serious impediment to construction of the wall, then the wall must be designed to support the anticipated driving and uplift groundwater forces. Usually in such cases, however, deep horizontal drains, typically consisting of 2-in. diameter slotted or perforated tubes and inclined upward at 5 to 10 degrees to the horizontal, are installed to relieve the groundwater pressures imposed on the retained soil mass. The design spacing and depth of these drains are site specific, but they will typically be longer than the


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length of the nails and with a density of approximately one drain per 100 square feet of face. Deep horizontal drains may also be used to control unanticipated water flow during construction. For aesthetic reasons, the drain outlets may have to be plumbed and carried down the wall face between the shotcrete construction facing and permanent CIP facing, and then outlet at the wall base. Horizontal drain flow exiting directly out through and flowing down the exterior permanent face is unsightly. 9.2.3.2 Construction Sequence of a Soil Nail Wall Excavation Prior to any excavation, surface water controls should be constructed to prevent surface water from flowing into the excavation as this condition will adversely affect construction and potentially cause instability of the excavated face. Collector trenches behind the limits of the excavation are used to intercept and divert surface water. Subsequently, soil excavation is performed using conventional earth-moving equipment from a platform, and final trimming of the excavation face is typically carried out using a backhoe or excavator from a platform. The initial lift is typically 3 to 4 feet high (Figure 9-7). The excavated face profile should be reasonably smooth and not too irregular to minimize excessive shotcrete quantities. Soil profiles containing cobbles and/or boulders may require hand excavation. A level working bench on the order of 30-ft wide is required to accommodate the conventional drilling equipment used for nail installation. Track drills smaller than the conventional drilling equipment can work on benches as narrow as 15 ft and with headroom clearance as low as 9 ft. For cases where the excavated slope face cannot stand unsupported for the required period of time, a continuous berm may be employed to stabilize the unsupported face section (Figure 9-8a). In this case the soil nails are installed and grouted first through the stabilizing berm. Subsequently, the berm is excavated and shotcrete is applied along the entire excavation level. Alternatively, the stabilizing berm can be removed in sequence in which alternating excavation slots are excavated and panels of the stabilizing berm are created in between (Figure 9-8b). The width of the excavation slots should not exceed approximately the soil nail horizontal spacing. Next, soil nails are installed and grouted, and shotcrete is applied on the face of the excavation slots. Subsequently, the stabilizing berm panels are removed and the intermediate soil nails are installed and shotcrete is applied. The use of stabilizing berms should be used only as a last resource because this procedure may have a significant impact FHWA NHI-07-071 Earth Retaining Structures

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Figure 9-7. Initial Excavation Lift and Nail Installation (after Porterfield et al., 1994). construction costs and project schedule. If temporary excavation stand up time is a major concern and represents a potential risk, alternative top-down construction methods such as ground anchors may be considered. Drilling of Nail Holes Nail holes are drilled at predetermined locations using one of several available drilling methods, including rotary (Figure 9-9), percussion, auger, and rotary/percussion drilling. Open-hole installation using auger drilling (in particular, hollow-stem augers) is most commonly used on soil nailing projects in the U.S. because no casing of the drillhole is necessary, high installation rates can be obtained, and costs are relatively low. Nail holes drilled using auger drilling can range between 4 to 12 in. in diameter. More commonly, drillholes are 6 to 8 in. in diameter. Contractors will usually select a relatively large drillhole diameter (e.g., 8 in.) to reasonably assure that the ultimate soil nail bond strength required in the construction specifications can be achieved without difficulties minimizing drilling equipment costs.


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GEOCOMPOSITE DRAIN STRIP TEMPORARY SHOTCRETE FACING

SOIL NAIL

BOTTOM OF PREVIOUS LIFT NAIL DRILLHOLE

SV

TEMPORARY STABILIZING BERM - SLOPE AS NECESSARY TO MAINTAIN FACE STABILITY; TO BE REMOVED AFTER NAIL INSTALLATION AND GROUTING, AND BEFORE SHOTCRETE APPLICATION

OVEREXCAVATION TO EXTEND WWM (AND GEOCOMPOSITE IF REQUIRED) PROVIDE ADDITIONAL LENGTH OF WWM TO MEET SPLICING LENGTH REQUIREMENT

1 n

FINAL WALL LAYOUT

NEXT STABILIZING BERM

(a) SH

APPLY SHOTCRETE BEFORE REMOVING STABILIZING BERMS

SV

WWM

SV

STABILIZING BERMS ALTERNATING SLOTS - WIDTH MUST BE COMPATIBLE WITH INSTALLATION EQUIPMENT AND FACE STABILITY REQUIREMENTS

MAX: SH (APPROX)

(b) Figure 9-8. Examples of Alternative Temporary Excavation Support (a) Stabilizing Berm and (b) Slot Excavation (after Porterfield et al., 1994). FHWA NHI-07-071 Earth Retaining Structures

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Figure 9-9. Typical Drilling of Soil Nails with Rotary Method (after Porterfield et al., 1994). Nail Bar Installation The selection of the drilling method may be controlled by the local availability of equipment and the specific ground conditions to be encountered. Typical soil nail wall contract documents allow the contractor to select the drilling method. However, the design engineer may occasionally restrict the choice of drilling methods and/or procedures based on the subsurface conditions or other project needs. The most common practice for placing nail bars is inserting them into a predrilled, straightshafted drillhole. After the nail bar is inserted in the drillhole, the drillhole is filled with clean cement grout, as discussed previously. This method is referred to as open-hole installation. As the grout sets, it bonds to the nail bar and the surrounding ground. The open-hole installation is by far the most commonly used method in soil nail wall construction. The most common U.S. practice of grouting by gravity provides bond strengths that are sufficient for soil nailing to be a feasible and cost-effective solution. However, in cases where poor soil conditions are encountered, higher bond strengths might be required. High bond strength may be achieved in granular soils and weak fissured rocks by injecting grout or FHWA NHI-07-071 Earth Retaining Structures

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regrouting under pressure. Grouting under pressure increases resistance to pullout relative to the open-hole method due to several mechanisms including an increase in the confining pressure around the grout bulb, higher compaction of the material surrounding the bulb, an interlocking mechanism (“fingers”), and an increase of the grout bulbs effective diameter. Grouting under pressure is seldom used for fine-grained soils. If used, only minor increases in bond strength should be expected because the frictional component and the interlocking mechanism are absent or not significant. Drainage Installation The migration of groundwater towards the excavation must be prevented. Therefore, conventional methods of surface water control and drainage are necessary during construction. These methods include intercepting ditches at the top of the excavation. Additionally, vertical geocomposite strip drains are installed behind the wall facing to prevent groundwater pressure to build up. The drain strips are secured against the excavation face and are placed with the geotextile side against the ground. During shotcrete application, the contamination of the geotextile side with shotcrete must be avoided to prevent loosing the flow capacity of the drains. Strip drains must be spliced at the bottom of each excavation lift and must have at least a 12-in. overlap such that the water flow is not impeded. The groundwater collected at strip drains is removed by a series of footing drains at the bottom of the excavation. The footing drain consists of a trench at the bottom of the excavation, which is filled with aggregate free of fines and has a PVC slotted collection pipe. The drainage geotextile must envelope the footing drain aggregate and pipe and conform to the dimensions of the trench. Additionally, weep holes can be installed through the wall facing at the lower portions of the wall. In special situation when the groundwater behind the proposed soil nail wall is high, conventional, deeper horizontal pipe drains are necessary. Temporary Wall Facing Construction Temporary wall facings for soil nail wall applications are usually constructed using shotcrete. The thickness of the temporary shotcrete facing is typically between 3 and 4 in. It is noted that this thickness range is based on the typical soil nail spacings described herein. Wider nail spacings would require the use of a thicker shotcrete facing. Shotcrete provides a continuous supporting layer over the excavated face that can also serve to fill voids and cracks on the excavated face. Temporary shotcrete applications have been constructed using both WWM or fiber reinforcement and bars. WWM is the preferred method among contractors because it requires less time to install while the excavated face is unsupported. A shotcrete facing for a wall under construction is shown in Figure 9-10. FHWA NHI-07-071 Earth Retaining Structures

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Figure 9-10. Shotcrete Temporary Facing (after Porterfield et al., 1994). Two types of shotcrete methods are commonly used: dry mix and wet mix. In the dry mix method, the aggregate and cement are blended in the dry and fed into the shotcrete gun while the mix water is added at the nozzle. Admixtures can be added at the mix plant or with the water. The addition of water at the nozzle allows the plasticity of the shotcrete to be adjusted at the nozzle, if required. In the wet mix method, the aggregate, cement, water, and admixtures are mixed in a batch plant and conveyed to the nozzle by a hydraulic pump. The plastic mix is applied at higher velocities by compressed air. Both shotcrete methods produce a mix suitable for wall facings. Dry mix and wet mix shotcrete use a water-cement ratio of about 0.4 and produce roughly the same mix quality, although shotcrete obtained with the wet mix process yields a slightly greater flexural strength. Keeping water cement ratios at about 0.4 and using air entrainment, which is difficult with the dry-mix process, enhances the durability of shotcrete. Low water-cement ratios (i.e., < 0.45) result in high strength, high durability, and low permeability as long as proper in-situ compaction (i.e., elimination of entrapped air in the shotcrete) is achieved. Steel fiber reinforcement has been added to shotcrete as part of a wet mix to increase ductility, toughness, and impact resistance. Fibers tend to reduce the shotcrete brittleness and thereby reduce crack propagation, but they have little effect on compressive strength and produce only a modest increase in flexural strength. Wet mix is often preferred for the construction of shotcrete facing walls because:


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•

wet mix yields higher production of fresh shotcrete, typically 2.2 to 3.0 yd3/hour versus .5 to 3.0 yd3/hour for dry mix;

•

the shotcrete rebound (i.e., loss of material due to lack of “stick”) for a wet mix is typically only about 5 percent, compared to 15 percent for a dry mix;

•

there is no need to add water at the nozzle, as in the case of a dry mix, thus it is less dependent on the nozzle operator’s experience;

•

equipment (e.g., concrete pump) is more readily available because shotcrete gun and moisturizer are not needed as with dry mix; and

•

supply of ready-mix concrete from commercial batch plants is readily available and convenient.

Welded wire mesh is commonly used as reinforcement for temporary facing but occasionally is also used in permanent facing. The cross-sectional area and mesh opening of the WWM are selected to satisfy structural requirements (i.e., flexural and punching shear capacities) and constructibility constraints. The selected WWM must have a width that is consistent with the excavation lift height (equivalent to the vertical nail spacing), plus an overlap of at least 8 in. The dimensions of the WWM (i.e., bar size and spacing) are evaluated as part of the soil nail wall design. Permanent Wall Facing Construction Several methods to provide a permanent facing have been used by contractors. These methods include reinforced shotcrete, cast-in-place (CIP) reinforced concrete, and precast concrete. The thickness of permanent facing consisting of reinforced shotcrete is typically between 6 and 12 in. This thickness does not include the thickness of the temporary facing. The permanent facing can be applied at the same time as the lift is excavated or can be applied later after the final lift of the temporary facing is completed and advance in phases from the bottom. The shotcrete facing is reinforced with WWM or rebar mesh. The final wall facing thickness is obtained by applying successive layers of shotcrete, each layer 2- to 4-in. thick. To achieve a good connection between the permanent facing and the nail bar, the nail head must be fully encapsulated in the permanent shotcrete facing. Headed studs are welded to the bearing plate to provide this connection (Figure 9-11).


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Figure 9-11. Headed-Studs Welded to Bearing Plate (after Porterfield et al., 1994). The reinforcement of permanent facing using CIP concrete typically consists of a mesh (standard reinforcing bars) and occasionally waler bars placed over the nail head. Reinforcement is placed approximately at the center of the facing section thickness. Because bonding between temporary shotcrete and permanent CIP concrete facings cannot be assured, the temporary facing is typically disregarded as a resisting element in the section design. A variety of finishes can be implemented by using commercially available form liners, as shown in Figure 9-12. The advantage of the CIP reinforced concrete is that the finish is more aesthetically pleasing and the quality of the concrete tends to be more homogeneous. The main disadvantage is the need for formwork and potentially longer construction time for facing installation. The use of reinforced shotcrete as permanent facing has the potential benefit of cost savings and efficiency as the same shotcrete equipment used for the temporary facing can be utilized. A major limitation of this technique is that the conventional finish of shotcreted walls is typically relatively rough and may not meet aesthetic requirements for a finished wall face. Precast concrete facing has been used in permanent applications to meet a variety of aesthetic, environmental, and durability criteria. A project using this method is shown in Figure 9-13. Precast facings also provide a means of integrating a continuous drainage blanket behind the facing and a frost protection barrier in cold climates. FHWA NHI-07-071 Earth Retaining Structures

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Figure 9-12. Cast-in-Place Facing.

Figure 9-13. Precast Panel Facing (after Elias et al., 2001). FHWA NHI-07-071 Earth Retaining Structures

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Precast concrete panels are either small segmental (i.e., nonstructural) elements or full-height tilt-up (i.e., structural) elements. Small segmental panels resist smaller, localized loads than structural elements, which can resist large loads. If non-structural elements are used for the permanent facing, the temporary facing requires adequate strengthening to resist the design loads. 9.2.4

Soil Nail Corrosion Protection

9.2.4.1 Evaluation of Soil Corrosion Potential Subsurface conditions exhibiting high corrosion potential usually do not preclude the use of soil nails, providing the design life, type of structure, and proper corrosion protection for the soil nail bars are properly considered. Various ground conditions promote corrosion including: (1) low electrical resistivity of soil; (2) high concentration of chlorides or sulfates; and (3) too low or too high hydrogen potential (pH) of soil or groundwater. Examples of soils with corrosion potential include: (1) acidic soils; (2) organic soil; and (3) soils with materials of industrial origin (slag, fly ash, fills with construction debris, mine tailings, and acid mine waste). Corrosion potential is of primary concern in aggressive soil applications and is evaluated based on tests results of pH, electrical resistivity, chloride content, sulfate content, and presence of stray currents. The test-based criteria listed in Table 9-1 are used to classify the corrosion potential of the ground. The ground is classified with a strong corrosion potential or aggressive if any one of the conditions listed in the first column of Table 9-1 exceeds the limits listed in the second column of the table during the service life of the soil nail wall, otherwise, the ground is classified as non-aggressive. 9.2.4.2 Corrosion Protection Systems Grout Protection This method of corrosion protection involves fully covering the bar with neat cement grout. After the bar is centered in the drillhole, neat grout is injected and fills up the annular space around the steel bar. Grout encapsulation provides both physical and chemical corrosion protection. When a minimum grout cover is in place, components such as carbonates and chlorides in the soil, and oxygen and humidity in the air are prevented or delayed in reaching the bar due to passivation. Additionally, the grout must have low permeability to ensure the effectiveness of the encapsulation. The grout provides an alkaline environment that reduces FHWA NHI-07-071 Earth Retaining Structures

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Table 9-1. Criteria for Assessing Ground Corrosion Potential. Test

Strong Corrosion Potential/Aggressive

Mild to no Corrosion Potential/Non-Aggressive

Test Method

PH Resistivity Sulfates Chlorides Stray current

< 4.5, > 10 < 2,000 ohm-cm > 200 ppm > 100 ppm Present

5.5 < pH < 10 Greater than 5,000 Less than 200 Less than 100 –

T 289-91 T 288-91 T 290-91 T 291-91 –

corrosion potential. A minimum grout cover of 1 in. between the bar and the soil should be specified. Epoxy Coating and Grout Corrosion protection with epoxy (Figure 9-14) consists of coating the nail bar with a fusionbonded epoxy that is applied by the manufacturer prior to shipment to the construction site. Cement grout is placed around all epoxy-coated nail bars. The minimum required thickness of epoxy coatings is 16 mils. The epoxy coating provides physical and chemical protection, as epoxy is a dielectric material. In transporting and handling bars, the epoxy coating may be damaged before nail installation. Therefore, it is not uncommon to spray epoxy coating in the field on chipped or nicked surfaces. Applicable standards for epoxy coating are found in ASTM A-775. Galvanized Coating and Grout A common method of providing corrosion protection is galvanization, which consists of applying a zinc coating on the steel surface. The process is performed by hot-dipping bars and other metallic pieces with zinc. The protection provided by galvanized coating is both physical and chemical, as this process forms a protective layer of zinc oxide. Cement grout is placed around all galvanized nail bars. Encapsulation For maximum corrosion protection, a polyvinyl chloride (PVC) or high-density polyethylene (HDPE), minimum 0.04-in. thick corrugated sheathing may be installed around the grouted nail bar (Figure 9-15), similar to what is done for ground anchors in permanent applications (whether the ground is aggressive or not). This practice is common in soil nail applications used for permanent soil nail walls built in an aggressive or unknown environment. The FHWA NHI-07-071 Earth Retaining Structures

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PERMANENT FACING TEMPORARY FACING

BEARING PLATE HEADED-STUD (TYP.)

GEOCOMPOSITE STRIP DRAIN

EPOXY-COATED BAR [MINIMUM THICKNESS OF EPOXY COATING: 0.4 mm (16 mil)] 16 mil

NUT AND BEVELED WASHER CENTRALIZER

2 in. 50 mm (2 in.) MINIMUM COVER


NAIL GROUT

6 in.(6 in.) 150 mm MINIMUM 2.5 8mft (8 ft) MAXIM UM

0.5 1m .5(1 ft.5 MAXIM ft) UM

FACING REINFORCEMENT NOT SHOWN

Figure 9-14. Grouted Epoxy-Coated Nail (after Byrne et al., 1998). sheathing is corrugated to transfer the effect of anchorage to the surrounding grout. Grout must completely fill the annular spaces inside and outside the sheathing. The minimum grout cover between the sheathing and the nail bar is 0.4 in. This distance allows the injected grout to flow without difficulty and provides sufficient physical protection. Outside the sheathing, the minimum grout cover between the sheathing and the drillhole wall must be 0.8 in. In some systems, the inner annular space is grouted in the shop and the whole assembly transported to the project site. The sheathing must be sufficiently strong to resist transportation, handling, and installation. Additionally, sheathing must be non-reactive with concrete, chemically stable, ultra-violet-light resistant, and impermeable. The minimum sheathing wall thickness is typically 0.04 in. Certain sheathing techniques may be proprietary. Nail-Head Corrosion Protection The use of materials made of galvanized steel and a minimum cover of 2 in. of concrete or permanent shotcrete provide corrosion protection of bearing plates, washers, and nuts. Epoxy coating can be applied on bearing plates and nuts.


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PERMANENT FACING TEMPORARY FACING

GEOCOMPOSITE STRIP DRAIN

BEARING PLATE

UNCOATED NAIL BAR

HEADED STUD (TYP)

INNER GROUT

NUT AND BEVELED WASHER

PVC SHEATHING OUTER GROUT CENTRALIZER END CAP ASSEMBLY


~

20 mm (0.8 in.) 1 in. MINIMUM COVER

6 in. 150 mm (6 in.) MINIMUM 2.5 m 8 ft (8 ft) MAXIM UM

0.5 1m .5(1ft.5 MAXIM ft) UM

BAR INNER GROUT

0.4(0.4 in. in.) 10 mm MINIMUM COVER

PVC SHEATHING

FACING REINFORCEMENT NOT SHOWN

OUTER GROUT

Figure 9-15. Sheathing-Encapsulation on Grouted Nail (after Byrne et al., 1998).

9.2.4.3 Corrosion Protection Levels Table 9-2 presents the protection levels and the protection systems commonly used in soil nail applications.

Table 9-2. Recommendations for Minimum Levels of Corrosion Protection for Soil Nails.

Temporary Permanent

Non-Aggressive Ground grout grout + epoxy coating; or grout + galvanization

Aggressive Ground See Note 1 grout + encapsulation; or grout + epoxy coating + encapsulation; or grout + galvanization + encapsulation

Note 1: Since temporary soil nails are oftentimes used for temporary support of excavation projects in which design may be the responsibility of the Contractor, it is recommended that the specific corrosion protection system be evaluated on projectspecific basis. If the temporary condition can be assured to be of a controlled duration (say less than 18 months) and if the incremental costs associated with encapsulation are prohibitive, the Owner may wish to consider, as a minimum, FHWA NHI-07-071 Earth Retaining Structures

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either grout and epoxy coating or grout and galvanization. If there is a potential for the service life of the soil nails to extend beyond 18 months and/or the incremental costs of providing encapsulation are not prohibitive, then the additional protection offered by encapsulation may be considered. When using corrosion protection measures shown in Table 9-2, it is not necessary to incorporate a sacrificial thickness into the design. In general, current U.S. public-sector projects do not use the approach of sacrificial thickness as a means to address potential corrosion for soil nail applications, especially for permanent applications. In temporary applications, unprotected, bare bars can be driven, as long as the soil corrosion potential is mild or insignificant. A preliminary and safe (for most conditions) estimate of the required sacrificial total thickness for unprotected bars is 0.08 in over the entire surface area. 9.2.5

Cost

The cost range for walls up to 60 ft in height is typically $15 to $35/ft2 of exposed wall face. A typical cost of a soil nail is $35/linear ft, which is less than the per unit length cost of a ground anchor. Reasons for the cost difference include: •

soil nails are usually not as long as ground anchors because soil nails develop load capacity starting right behind the wall whereas ground anchors develop load in a bond zone which may be located at a relatively large distance behind the wall;

•

soil nails are usually smaller in diameter than ground anchors; and

•

load testing is only performed on a small number of soil nails (typically 5 percent of the total production soil nails) compared to all of the ground anchors.

Soil nail walls are also less expensive than anchored walls because, unlike anchored walls, soil nail walls do not need a vertical structural wall element. The cost range provided above does not include the cost for the permanent wall facing. Typically, permanent concrete wall facing can add $20 to $30/ft2 to the cost of a soil nail wall. If casing of the drillhole is required for the proper installation of the soil nails, the costs for the soil nail wall system can increase significantly.


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9.2.6

Design Concepts For Soil Nails

9.2.6.1 Load Transfer in Soil Nail Walls The following wall response and load transfer mechanisms take place during a conventional soil nail construction: •

Soil excavation is initiated from the ground surface and the Excavation Phase 1 is completed (Figure 9-16). Because of the soil ability to stand unsupported, the upper portion of the soil behind the excavation is stable (or at least marginally stable) before the first row of nails (Nails 1) is installed. Soil strength is mobilized along the uppermost potential critical failure surface to allow the unsupported soil wall to stand.

•

As Nails 1 and the temporary facing are installed, some load derived from the deformation of the upper soil is transferred to these nails through shear stresses along the nails and translate into and axial forces. The top portion of Figure 9-16 shows schematically the axial force distribution in Nails 1 at the end of excavation Phase 1. At this point, the temporary facing supports the excavation surface and provides connectivity between adjacent nails in row of Nails 1.

•

As excavation proceeds to Excavation Phase 2, the uppermost and the unsupported portions of the soil nail wall deforms laterally. At this point, another potential sliding surface, one originated from base of Excavation Phase 2 is formed (Figure 9-16). The critical failure surface at this excavation level is different than that at the previous excavation level.

•

Nails 2 are then installed. Subsequently the temporary facing between the bottom of excavation Phases 1 and 2 is installed and integrated to the facing constructed in Phase 1. Subsequent movements of the soil above the Phase 2 depth will cause additional loads to be transferred to Nails 1 and generate loads in Nails 2. Note the increased nail force distribution for Nails 1 at the end of excavation Phase 2.

•

To provide global stability, the soil nails must extend beyond the potential failure surface. As lateral deformation increases due to subsequent excavation, additional shear stresses along the soil nail/soil interface and axial forces of the previously installed nails are mobilized. As the depth of excavation increases, the size of the retained soil mass increases, as shown in Figure 9-16.

•

As the size of the retained zone increases, the stresses at the soil/nail interface and the axial forces in the nails increase. The induced tensile stresses are transferred behind the retained zone in an anchorage effect. These stresses ultimately tend to stabilize the potentially sliding mass.


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Schematic Distribution of Nail Axial Force, T, in Nail 1 after each excavation phase

Deflection pattern at end of each phase

N

2 1

Excavation Phase 1

Nail 1

H

Excavation Phase 2

Critical Failure Surfaces for Excavation Phases 1, 2, ..., N

Excavation Phase N

Nail N

Figure 9-16. Potential Failure Surfaces and Soil Nail Tensile Forces (after Lazarte et al., 2003).

•

While the tensile force in the intermediate and lower nails may increase as the excavation depth increases, the tensile force in some of the upper nails may decrease due to load redistribution. For example, the upper portion of Figure 9-16 shows schematically that the axial force distribution for Nails 1 at the end of the last excavation Phase N does not exhibit the largest values.

•

As the critical failure surface becomes deeper and larger, the contribution of the upper nails to the stabilization of this larger sliding mass diminishes. In some cases, upper nails may be entirely ineffective in the assessment of deep critical failure surfaces. However, the upper nails should not be considered superfluous, because they contribute to the stability during earlier stages of excavation and help reduce lateral displacements.


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The analysis of soil nail walls must consider both “during construction” and “post construction” loading conditions to establish the most critical case at each soil nail level. The most critical situation may arise after the wall is completed due to a combination of longterm design loads (e.g., dead load, live load, and traffic) and extreme loads (e.g., earthquake). In other situations, the most critical case may occur during construction when the then lowest excavation surface remains temporarily unsupported and the soil nails and shotcrete are not yet installed (Figure 9-17). These critical short-term loading conditions can be exacerbated by temporary seepage conditions.

H

Potential seepage

Critical Failure Surface for Temporarily Unsupported Excavation

Figure 9-17. Potential Critical Stability During Construction (after Lazarte et al., 2003). 9.2.6.2 External Failure Modes General External failure modes refer to the development of potential failure surfaces passing through or behind the soil nails (i.e., failure surfaces that may or may not intersect the nails). For external failure modes, the soil nail wall mass is generally treated as a block. Stability calculations take into account the resisting soil forces acting along the failure surfaces to establish the equilibrium of this block. If the failure surface intersects one or more soil nails, the intersected nails contribute to the stability of the block by providing an external stabilizing force that must be added to the soil resisting forces along the failure surface. FHWA NHI-07-071 Earth Retaining Structures

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The design of a soil nail wall should ensure that the system is safe against all of the potential failure conditions presented in Figure 9-18. In this section, discussion of external failure modes including global stability, sliding, and bearing capacity is presented. Information on global stability associated with seismic forces is provided in GEC No. 7 (Lazarte et al., 2003). EXTERNAL FAILURE MODES

SOIL STRENGTH SOIL STRENGTH

NAIL RESISTANCE

HEAVE

FAILURE SURFACE

SOIL STRENGTH AT BASE

(a) GLOBAL STABILITY FAILURE

(b) SLIDING STABILITY FAILURE

(c) BEARING FAILURE (BASAL HEAVE)

INTERNAL FAILURE MODES GROUT

BAR BREAKAGE

V M

FAILURE SURFACE

M = Moment V = Shear

(d) NAIL-SOIL PULLOUT FAILURE

(e) BAR-GROUT PULLOUT FAILURE

(f) NAIL TENSILE FAILURE

(g) NAIL BENDING AND/OR SHEAR FAILURE

FACING FAILURE MODES HEADED-STUD BREAKAGE

FAILURE SURFACE PLASTIC MOMENT

(h) FACING FLEXURE FAILURE

(i) FACING PUNCHING SHEAR FAILURE

(j) HEADED-STUD FAILURE

Figure 9-18. Principal Modes of Failure of Soil Nail Wall Systems (after Lazarte et al., 2003).


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Global Stability Global stability refers to the overall stability of the reinforced soil nail wall mass. As shown in Figure 9-17, the slip surface passes behind and beneath the soil nail wall system. In this failure mode, the retained mass exceeds the resistance provided by the soil along the slip surface and the nails, if intersected. The global stability of soil nail walls is commonly evaluated using two-dimensional limit-equilibrium principles, which are used for conventional slope stability analyses. As with traditional slope stability analyses, various potential failure surfaces are evaluated until the most critical surface (i.e., the one corresponding to the lowest factor of safety) is obtained. To illustrate the elements of a global stability analysis for soil nail walls, a simple, singlewedge failure mechanism is shown in Figure 9-19.

Q

β i

W

S = R c + Rφ

H

T

α

φ'

N

LS

ψ

Figure 9-19. Global Stability Analysis of Soil Nail Wall using a Single-Wedge Failure Mechanism (after Lazarte et al., 2003).


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The destabilizing forces consist of the driving components of the weight (W) and the surcharge loads (Q). The stabilizing forces along the failure surface are the shear force (S) and the equivalent nail tensile force (T). It is noted that the force T is a resultant force that combines the effect of all nails installed to that depth H. The factor of safety against global failure (FSG) is expressed as the ratio of the resisting and driving forces, which act tangent to the potential failure plane:

FSG =

∑ resisting forces ∑ driving forces

(9-1)

The normal and tangent forces on the failure plane are:

∑ Normal Forces = (W + Q) cos ψ + T cos (ψ − i) − N = 0

(9-2)

∑Tangent Forces = (W + Q) sin ψ − T sin (ψ − i) − S = 0

(9-3)

S = R c + R f = c * L s + N tan φ *

(9-4)

where:

tanφ * =

tanφ' FSG

(9-5)

where: α β φ’ c’ ψ i Ls W Q T N S Rc Rφ

= = = = = = = = = = = = = =

wall face batter angle (from vertical); slope angle; soil effective angle of internal friction; soil effective cohesion; inclination of failure plane; nail inclination; length of failure plane; weight of sliding mass; surcharge load; equivalent nail force; normal force on failure surface; shear force on failure surface; cohesive component of S; and frictional component of S.


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c* =

c' FSG

(9-6)

and φ* is the mobilized friction angle, and c* is the mobilized cohesion. A single global factor of safety is used for the cohesive and frictional strength components of the soil (c’ and tanφ’, respectively). However, it is possible to select different safety factors for each strength component. The simplistic analysis presented above only considers force equilibrium. More rigorous limit equilibrium slope stability analysis methods allow establishing simultaneously moment and force equilibrium equations. The simple model shown in Figure 9-19 and presented in the above equations may be used to perform an independent verification of the computer’s solution. A global stability analysis can be used to complete either (or both) of the following two tasks related to the analysis of soil nail walls: 1) calculate the critical (minimum) factor of safety FSG of the sliding mass for a given soil nail length pattern; or 2) determine the required force T in all nails that will yield a selected target factor of safety against global failure. Global stability analyses are performed using computer programs specifically developed for the design of soil nail walls. The two computer programs most commonly used in the United States for the analysis and design of soil nail walls are SNAIL and GOLDNAIL. These programs can consider failure surfaces that are more complex than the simple planar, singlewedge. SNAIL uses two-part planar wedges; GOLDNAIL uses circular failure surfaces that consider multiple slices in lieu of wedges. These programs are similar, in many respects, to general slope stability computer programs (e.g., search routines, closed form force/moment equilibrium equations, etc.). However, computer programs dedicated to soil nail design include the iterative and interactive design of the soil nail length and the consideration of other failure modes (e.g., soil nail tensile force and facing punching shear failure). Sliding

Sliding stability analysis considers the ability of the soil nail wall to resist sliding along the base of the retained system in response to lateral earth pressures behind the soil nails. Sliding failure may occur when additional lateral earth pressures, mobilized by the FHWA NHI-07-071 Earth Retaining Structures

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excavation, exceed the sliding resistance along the base (Figure 9-20). In general, the factor of safety against sliding, FSSL is 1.3 for temporary walls and 1.5 for permanent walls. Concepts similar to those used to assess sliding stability of gravity retaining structures (in which Rankine or Coulomb theories of lateral earth pressures are used) can be applied to assess the sliding stability of a soil nail wall system. Sliding stability analyses are presented in Section 9.2.7.4 in subsection titled “Sliding Stability”. Bearing Capacity

Bearing capacity may be a concern when a soil nail wall is excavated in fine-grained, soft soils. Because the wall facing does not extend below the bottom of the excavation (unlike soldier piles in cantilever or ground anchor walls), the unbalanced load caused by the excavation may cause the bottom the excavation to heave resulting in a bearing capacity failure of the foundation (Figure 9-21). Bearing capacity analyses are presented in Section 9.2.7.4 in subsection titled “Bearing Capacity”.

Q

2H

For infinite slopes:

∆H

β

βeq

For broken slopes:

βeq = β ∆H

βeq = tan-1(2H ) 1

γ, φ', c' = 0 α

H1

W

PA

δ = βeq

H

H/3

θ B

Strength parameters

cb and φ'b

ΣR

Figure 9-20. Sliding Stability of a Soil Nail Wall (after Lazarte et al., 2003).


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Load expressed as equivalent soil overburden: γ∆H

B

B'

(width of excavation is typically very large)

γ∆H

D

HB'

HB'

Su H

soft fine-grained soil

H

Su H

H

soft fine-grained soil

D

B/ 2

Failure surface b) Shallow deposit of soft fine-grained soil underlain by stiff layer

a) Deep deposit of soft fine-grained soil 10

B/L = 1, Square and Circular

9

B/L = 0.5

H = Excavation depth B = Excavation width L = Excavation Length

8

NC

7

B/L = 0, Rectangular

6 5 4 0

1

2

3

4

5

H/B c) Bearing Capacity Factor, NC

Figure 9-21. Bearing Capacity (Heave) Analysis (after Terzaghi et al., 1996).

Factors of safety against heave for soil nail walls should be selected to be consistent with those typically used for heave analysis at the bottom of excavations. In general, minimum FSH can be adopted as 2.5 and 3 for temporary and permanent walls, respectively. As the great majority of soil nail walls are not constructed in soft fine-grained soils, this failure mode is not critical for most soil nail projects. 9.2.6.3 Internal Failure Modes General

Internal failure modes refer to failure in the load transfer mechanisms between the soil, the nail, and the grout. Soil nails mobilize bond strength between the grout and the surrounding FHWA NHI-07-071 Earth Retaining Structures

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soil as the soil nail wall system deforms during excavation. The bond strength is mobilized progressively along the entire soil nail with a certain distribution that is affected by numerous factors. As the bond strength is mobilized, tensile forces in the nail are developed. Depending on the soil nail tensile strength and length, and the bond strength, bond stress distributions vary and different internal failure modes can be realized. Typical internal failure modes related to the soil nail are (Figures 9-18 d–g): •

Nail Pullout Failure: Nail pullout failure is a failure along the soil-grout interface due to insufficient intrinsic bond strength and/or insufficient nail length, Figure 918d.

•

Slippage of the Bar-Grout Interface: The strength against slippage along the grout and steel bar interface (Figure 9-18e) is derived mainly from mechanical interlocking of grout between the protrusions and “valleys” of the nail bar surface. Mechanical interlocking provides significant resistance when threaded bars are used and is negligible in smooth bars. The most common and recommended practice is the use of threaded bars, which reduces the potential for slippage between the nail bar and grout.

•

Tensile Failure of the Nail: The nail can fail in tension if there is inadequate tensile strength, Figure 9-18f.

•

Bending and Shear of the Nails: Due to relatively modest contribution resulting from shear and bending stresses in nails, the shear and bending strengths of the soil nails are conservatively disregarded in the guidelines contained in this document. The demand in flexure and shear is minimal as well. A discussion of a methodology to evaluate shear and bending behavior is included in Elias and Juran (1991).

A discussion of the two most common internal failure modes (i.e., nail pullout and nail tensile failure) is presented in the following two sections. Nail Pullout Failure

Pullout failure is the primary internal failure mode in a soil nail wall. This failure mode may occur when the pullout capacity per unit length is inadequate and/or the nail length is insufficient. In general, the mobilized pullout per unit length, Q, (also called the load transfer rate) can be expressed as: Q = π q D DH


9-44

(9-7)


where: q = mobilized shear stress acting around the perimeter of the nail-soil interface; and DDH = average or effective diameter of the drill hole. Actual distributions of mobilized bond shear stress (and load transfer rates) are not uniform, as illustrated in Figure 9-22, and depend on various factors including nail length, magnitude of applied tensile force, grout characteristics, and soil conditions. As a simplification, the mobilized bond strength is often assumed to be constant along the nail, which results in a constant load transfer rate, Q. As a result, the nail force at the end of the pullout length, Lp, is:

T(L p ) = To = Q L p

(9-8)

The pullout capacity, Rp, is mobilized when the ultimate bond strength is achieved and is expressed as: R p = Tmax = Q u L p

(9-9)

Q u = π q u D DH

(9-10)

with:

where: Qu = pullout capacity per unit length (also referred to as load transfer rate capacity); and qu = ultimate bond strength. Typical values of ultimate bond strength for various soils and drilling methods are presented in Table 9-3. The lower and upper bounds provided in Table 9-3 correspond approximately to the least and most favorable conditions for a particular ground type and construction method. These values inherently contain some level of conservatism and can be used as preliminary values for design. It is common practice to require preproduction soil nail load tests to verify the bond strengths included in the construction specifications and to establish the minimum required nail length to support a specified nail design load.


9-45


q T + dT

T

dx

q

To (end load)

DDH

x q (x)=constant Mobilized shear stress at interface

Actual distribution of nail tensile force

T (x) Q

To

1

Lp Figure 9-22. Single Nail Stress-Transfer Mode (after Lazarte, 2003). The following allowable values of the bond strength or pullout capacity per unit are used in design: q ALL =

qu FS p

R p ALL =


Rp FS P

9-46

(9-11)

(9-12)


Table 9-3. Estimated Ultimate Bond Strength of Soil Nails in Soil and Rock (after Elias and Juran, 1991). Material

Rock

Construction Method

Soil/Rock Type

Rotary Drilled

Marl/limestone Phyllite Chalk Soft dolomite Fissured dolomite Weathered sandstone Weathered shale Weathered schist Basalt Slate/Hard shale Sand/gravel Silty sand Silt Piedmont residual Fine colluvium Sand/gravel low overburden high overburden Dense Moraine Colluvium Silty sand fill Silty fine sand Silty clayey sand Sand Sand/gravel

Rotary Drilled

Cohesionless Soils

Driven Casing

Augered Jet Grouted Rotary Drilled Driven Casing

Silty clay Clayey silt Loess Soft clay Stiff clay Stiff clayey silt Calcareous sandy clay

Fine-Grained Soils Augered

Ultimate Bond Strength (psi) 43.5 – 58.0 14.5 – 43.5 72.0 – 86.5 58.0 – 86.5 86.5 – 144.5 29.0 – 43.5 14.5 – 21.5 14.5 – 25.5 72.0 – 86.5 43.5 – 58.0 14.5 – 26.0 14.5 – 21.5 8.5 -11.0 6.0 – 17.5 11.0 – 21.5

27.5 – 34.5 40.5 – 62.5 55.0 – 69.0 14.5 – 26.0 3.0 – 6.0 8.0 – 13.0 8.5 – 20.5 55.0 101.0 5.0 – 7.0 13.0 – 20.5 3.5 - 11 3.0 – 4.5 6.0 – 8.5 6.0 – 14.5 13.0 – 20.5

where FSP is the factor of safety against pullout failure. In general, a minimum factor of safety of 2 is recommended against pullout failure. Soil-Nail Interaction and Tensile Force Distribution

The soil-nail interaction that occurs behind the wall facing is complex. The loads applied to the soil nails originate as reactions to the outward wall movement during excavation of the soil in front of the wall, as discussed earlier. The portion of the nail behind the failure FHWA NHI-07-071 Earth Retaining Structures

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surface (i.e., the anchoring zone) is pulled out of the soil slope. The tensile forces in the soil nail, T, vary from the anchoring zone to the facing as follows: they start as zero at the end of the nail, increase to a maximum, Tmax, value in the intermediate length, and decrease to a value To at the facing (Figure 9-23). q(x)

Facing

DDH

(a)

L x

q(x) (+)

(b)

(-)

T(x) Tmax (c)

To Tmax = Maximum Nail Force To = Tensile force at nail head

Figure 9-23. Soil Nail Stress-Transfer Mechanism (after Lazarte, 2003). The maximum nail tensile force in the nail bar nail does not necessarily occur at the point where the nail crosses the failure surface. The mobilized shear stress along the grout-soil interface, q, is not uniform and, in fact, changes from “positive” to “negative”, as shown in Figure 9-23a and b. The schematic distribution of the tensile force (T) along the soil nail is shown in Figure 9-23c. For design, the tensile force distribution along the nail shown in Figure 9-23 can be simplified as shown in Figure 9-24. The tensile force the nail increases at a constant slope Qu (equal to the pullout capacity per unit length), reaches a maximum value, Tmax, and then decreases at the rate Qu to the value To at the nail head. With reference to Figure 9-24, the following three conditions related to the maximum tensile force are noted. The value Tmax is bounded by three limiting conditions: the pullout capacity, RP, the tensile capacity, RT, and the facing capacity, RF. If RP < RT and RF, pullout failure controls the value of Tmax. If RT FHWA NHI-07-071 Earth Retaining Structures

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RF

T(x)

RT

RP

1

To

1

Qu

Tmax

LP RT = Nail Tensile Capacity RF = Facing Capacity RP = Pullout Capacity Qu, qu = Ultimate load transfer rate and bond strength To ~ 0.6-1.0 Tmax 1) RP < RT < RF (pullout controls, shown in example above ) 2) RT < RP < RF (tensile failure controls) 3) RF < RP or RT (facing failure may control depending on To/Tmax)

Figure 9-24. Simplified Distribution of Nail Tensile Force (after Lazarte, 2003).

< RP and RF, tensile failure controls Tmax. Finally, if RF < RT and RP, failure of the facing may control, depending on the ratio of To/Tmax. To achieve a balanced design, all of the resisting components in a system should have comparable margins of safety; no component should be significantly oversized or undersized. In the case of nail tensile forces, a good design should balance the capacities of all resisting elements; therefore, values of RP, RT, and RF should be reasonably similar. Design for Nail Tensile Failure

To achieve a balanced design for all internal failure modes, the soil strength must be fully mobilized consistently with the full mobilization of the nail tensile strength at the same time. In other words, when FSG = 1.0 (full soil mobilization), the safety factor for the tensile strength, is FST = 1.0 (full nail tensile mobilization). The nail tensile force for this condition is the maximum design force in the nail (Tmax-s). It is intuitive that when the loads are kept constant, the design force Tmax-s will increase when FSG > 1.0. This is caused because for FSG FHWA NHI-07-071 Earth Retaining Structures

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>1, the soil strength is not fully mobilized and the tensile forces must compensate to achieve equilibrium. Therefore, calculating Tmax-s directly from the global stability analysis giving FSG > 1 is more conservative. The computation of Tmax-s is discussed below. The program SNAIL automatically reports the average nail tensile force, but not the maximum tensile force corresponding to FSG = l. Thus, to estimate the maximum nail tensile force for a FSG = l without performing an additional stability analysis, the following simplified method can be used. This procedure is based on the fact that the ratio of the maximum nail load calculated by SNAIL, Tmax, to the average nail load, Tavg, for FSG > 1, is similar to the ratio of the maximum nail load for FSG = 1, Tmax-s, to the average nail load, Tavg-s, for FSG = 1. Therefore, a good approximation of the maximum design nail load (Tmax-s) can be obtained by the following relationship: Tmax −s Tavg−s = Tmax Tavg

(9-13)

Tavg-s is the average design nail load and is reported by SNAIL in output files as the “Maximum Average Reinforcement Working Force”. The design nail force Tmax-s is compared to the tensile capacity of the nail, which is defined as follows. A tensile failure of a soil nail takes place when the longitudinal force along the soil nail, Tmaxs, is greater than the nail bar tensile capacity, which is defined as: R T = Atf y

(9-14)

where At is the nail bar cross sectional area and fy is the nail bar yield strength. The tensile capacity provided by the grout is disregarded, due to the difference in stiffness (i.e., modulus of elasticity) between the grout and the nail. To take into account uncertainties related to material strength and applied loads, allowable values of the nail tensile capacity are used in design as follows: R

T ALL

=

RT FST

(9-15)

where FST is the factor of safety against soil nail tensile failure. In general, a minimum factor of safety of 1.8 is adopted for static loads.


9-50


9.2.6.4 Facing Connection Failure Modes General

The most common potential failure modes at the facing-nail head connection are presented in Figure 9-18 and are shown in detail in Figure 9-25 as: •

Flexure Failure: This is a failure mode due to excessive bending beyond the facing’s flexural capacity. This failure mode should be considered separately for both temporary and permanent facings.

•

Punching Shear Failure: This failure mode occurs in the facing around the nails and should be evaluated for both temporary and permanent facings.

•

Headed-Stud Tensile Failure: This is a failure of the headed studs in tension. This failure mode is only a concern for permanent facings.

For each of these failure modes, the nail head and facing must be designed to provide capacity in excess of the maximum nail head tensile force (To) at the wall face. Appropriate dimensions, strength, and reinforcement of the facing and suitable nail head hardware (e.g., bearing plate, nut, and headed studs) must be provided to achieve the design capacities with adequate factors of safety for all potential failure modes. Tensile Forces at the Wall Facing

The nail tensile force at the wall face, To, is less than or equal to the maximum nail tensile force (Byrne et al., 1996). Discussion on measured nail tensile forces based on several research studies is provided in GEC No. 7 (Lazarte et al., 2003). For design, the following is recommended for the nail head tensile force: To = Tmax -s [0.6 + 0.057 (S [ft ] − 3)]

(9-16)

where: = Design nail head tensile force; To Tmax-s = Maximum design nail tensile force obtained from global stability analysis using results from a computer program (e.g., SNAIL); and S = Maximum soil nail spacing. Use maximum of SV and SH, the vertical and horizontal nail spacing, respectively, in Equations 9-16 and 9-17.


9-51


STEEL IN TENSION VERTICAL MOMENT mV

WWM OR BAR VERTICAL MOMENT mV

WWM BEARING PLATE HEADED STUD

FLEXURE FAILURE

CONICAL SURFACE

45°

RFP/2

(PUNCHING SHEAR RESISTANCE)

TO

BEARING PLATE

RFP/2

PUNCHING SHEAR FAILURE (TEMPORARY FACING) COMPOSITE CONICAL SURFACE

REINFORCEMENT

RFP/2

RFP/2 RFH HEADED STUD

PUNCHING SHEAR FAILURE (PERMANENT FACING)

HEADED-STUD TENSILE FAILURE

Figure 9-25. Facing Connection Failure Modes (after Lazarte, 2003). FHWA NHI-07-071 Earth Retaining Structures

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For a typical nail head spacing of 5 ft, the nail head tensile force corresponds to a recommended facing service load of about 0.7 times the maximum nail service load. Facing Flexural Capacity

The soil nail wall facing can be considered a continuous reinforced concrete slab where the loading is the lateral earth pressure acting on the facing and the supports are the tensile forces in the soil nails (Figure 9-26a and b). The loads from the lateral earth pressure and the “reaction” in the soil nails induce flexural moments in the facing section. Positive moments (i.e., tension on the outside of the section) are generated in the midspan between nails; negative moments (i.e., tension on the inside of the section) are generated around the nails (Figure 9-26b). If these moments are excessive, a flexural failure of the shotcrete may occur. As with other reinforced concrete/shotcrete structures, flexural failure occurs progressively. After the first yield of the facing section (Figure 9-26c), progressive cracking takes place on both sides of the facing as the lateral earth pressure increases. As the lateral pressure increases, fractures grow and deflections and nail tensile forces increase. Individual fractures indicate where the flexural capacity is achieved. Eventually, an ultimate stage of the structure is achieved when all fractures connect, act as hinges, and form a mechanism referred to as the critical yield line pattern. In theory, the soil pressure that causes facing failure (i.e., the critical yield line pattern) can be applied to an influence area around the nail head, and a nail tensile force (“reaction”) is obtained. This force is designated as the facing flexure capacity, RFF, and is related to the flexural capacity per unit length of the facing. The flexural capacity per unit length of the facing is the maximum resisting moment per unit length that can be mobilized in the facing section. Based on yield-line theory concepts, RFF can be estimated as the minimum of:  S h[ft]   × f y [ksi ] R FF [kip] = 3.8 × CF × (a vn + a vm ) in 2 /ft ×  H  Sv 

(9-17a)

 S h[ft]   × f y [ksi ] R FF [kip] = 3.8 × CF × (a hn + a hm ) in 2 /ft ×  v S H  

(9-17b)

[

[

]

]

where: CF

h

= factor that considers the non-uniform soil pressures behind the facing (Byrne et al., 1996); =

thickness of facing (Figure 9-27);


9-53


FRACTURES ON EXTERNAL FACE

A

SOIL REACTION

BEARING PLATE

IDEALIZED DEFLECTION PATTERN AT ULTIMATE LOAD

TO

INITIAL POSITION

HINGE (TYP)

FRACTURES ON INTERNAL FACE

A

(a) IDEALIZED YIELD LINE PATTERN

TO

To ultimate= RFF

(b) ULTIMATE DEFORMATION PROFILE SECTION A-A

PROGRESSIVE CRACKING

FIRST YIELD

DEFLECTION,

(c) IDEALIZED LOAD-DEFLECTION BEHAVIOR

Figure 9-26. Progressive Flexural Failure in Wall Facings (after Lazarte, 2003). avn avm ahn ahm

= cross sectional area of reinforcement per unit width in the vertical direction at the nail head; = cross sectional area of reinforcement per unit width in the vertical direction at midspan; = cross sectional area of reinforcement per unit width in the horizontal direction at the nail head; = cross sectional area of reinforcement per unit width in the horizontal direction at midspan;


9-54


SH SV fy fc′

= = = =

nail horizontal spacing; nail vertical spacing; reinforcement tensile yield strength; and concrete compressive strength.

The factor CF takes into account the non-uniform soil pressures behind the facing (Byrne et al., 1998) and represents nominally the ratio of soil pressure behind the nail to soil pressure in the midspan between nails. The soil pressure distribution behind the facing is generally non-uniform. Soil pressure is affected by soil conditions and the facing stiffness, which in turn affects the wall displacement. In the midspan between nails, the displacement of the facing occurs outward and the lateral earth pressure is relatively low. Around the nail heads, the soil pressure is larger than the soil pressure at midspan between nails. The pressure distribution in the facing also depends on the stiffness of the facing. When the facing is relatively thin (as with typical temporary facings), the facing stiffness is relatively low, causing the facing to deform in the midspan sections. As a result, the soil pressure tends to be relatively low in the midspan sections. When the facing is relatively thick, the facing stiffness increases and the resulting wall deformations are smaller than would result from a thin wall facing. As a result of the increased wall stiffness, the soil pressure is more uniform throughout. Table 9-4 shows factors (CF) for typical facing thickness. For all permanent facings and “thick” [≥ 8 in.] temporary facings, the soil pressure is assumed to be relatively uniform. In Equations 9-17a and 9-17b, it is assumed that the maximum moments in the facing are around a horizontal axis and the design of reinforcement in the vertical direction is more critical than the design of the horizontal reinforcement. In practice, the cross section area of reinforcement in the horizontal direction is the same as for the vertical direction (i.e., ahm = avn and ahm = avm); therefore, the most critical case is the one that gives the minimum of SH/SV and SV/SH. When the same nail spacing and reinforcement are used in the horizontal and vertical directions, and Grade 60 is used, Equations 9-17a and 9-17b simplify as: R FF [kip] = 230 × C F × (a vn + a vm ) [in 2 /ft] × h[ft]


9-55

(9-18)


SH d = 0.5 h

Rebar Mesh or WWM (Final facing)

A

h

avm

avn Waler Bar

Waler Bar (TYP)

ahn

SV

WWM (Temporary Facing)

ahm

h d = 0.5 h

A Section A-A

Figure 9-27. Geometry used in Flexural Failure Mode (after Lazarte, 2003).

Table 9-4. Factors CF (after Lazarte, 2003).

Type of Structure

Temporary Permanent


Nominal Facing Thickness

Factor CF

in. 4 6 8 All

2.0 1.5 1.0 1.0

9-56


These equations can be used for temporary or permanent facing. The thickness of the temporary concrete facing is generally conservatively disregarded when evaluating the flexural capacity of the permanent facing as shown in Figure 9-27. If (vertical) waler bars are used over the nail heads, the total reinforcement area per unit length in the vertical direction can be calculated as: a vn = a vm +

A vw SH

(9-19)

where Avw is the total cross sectional area of waler bars in the vertical direction. Similar concepts can be applied along the horizontal direction. If rebar is used in permanent facings instead of WWM, the total area of reinforcement must be converted to a per unit length basis as: a vm =

A vm SH

(9-20)

where Avm is the total cross sectional area of rebar reinforcement in the vertical direction (see Figure 9-27). Given the tensile force at the soil nail head, To, and the facing flexure capacity, the safety factor against facing flexural failure can be defined. FS FF =

R FF To

(9-21)

In general, a minimum factor of safety of 1.35 is adopted for static loads in temporary walls and 1.5 for static loads in permanent walls. Limit Reinforcement in Facing

As with other reinforced concrete structures, the quantity of reinforcement placed in the facing of soil nail wall generally falls within prescribed limits. The amount of reinforcement can be expressed as a reinforcement ratio: ρ=


a ij 0.5 h

9-57

100

(9-22)


where aij is the cross sectional area of reinforcement per unit width in the “i” direction (vertical or horizontal) and at location “j” (nail head or midspan). The minimum reinforcement ratio, expressed in SI and English units, typically occurs in midspan and is defined as:

ρ min [%] = 0.24

f c' [psi]

(9-23)

f y [ksi]

The maximum reinforcement ratio is recommended to be:  f c' [psi]  90   ρ max [%] = 0.05 f y [ksi]  90 + f y [ksi] 

(9-24)

Therefore, the placed reinforcement must be: ρmin ≤ ρ ≤ ρmax. In addition, the ratio of the reinforcement in the nail and midspan zones should be less than 2.5 to ensure comparable ratio of flexural capacities in these areas. Facing Punching Shear Capacity

Punching shear failure of the facing can occur around the nail head and must be evaluated at: 1) bearing-plate connection (used in temporary facings), and 2) headed-stud connection (commonly used in permanent facings). As the nail head tensile force increases to a critical value, fractures can form a local failure mechanism around the nail head. This results in a conical failure surface, as shown in Figure 9-28. This failure surface extends behind the bearing plate or headed studs and punches through the facing at an inclination of about 45 degrees, as shown schematically in Figure 928. The size of the cone depends on the facing thickness and the type of the nail-facing connection (i.e., bearing-plate or headed-studs). As is common for concrete structural slabs subjected to concentrated loads, the nail-head capacity must be assessed in consideration of punching shear, RFP, and can be expressed as: R FP = C P VF

(9-25)

where VF is the punching shear force acting through the facing section and CP is a correction factor that accounts for the contribution of the support capacity of the soil. FHWA NHI-07-071 Earth Retaining Structures

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The punching shear force can be calculated considering both SI and English units using standard equations for punching shear. These equations consider the size of a conical failure surface (with diameter D’C at the center of the facing and height hC, as shown in Figure 9-28) at the level of the concrete slab as: VF [kip] = 0.58 f c' [psi] π D c ' [ft] h c [ft]

(9-26)

where: D’C = effective diameter of conical failure surface at the center of section (i.e., an average cylindrical failure surface is considered); and hC = effective depth of conical surface.

The correction factor CP is used to take into account the effect of the soil pressure behind the facing that acts to stabilize the cone. If no subgrade reaction is considered, CP = 1.0. When the soil reaction is considered, CP can be as high as 1.15. For practical purposes, the correction is usually omitted and this is considered as CP = 1.0. These equations can be used for both temporary and permanent facing. However, the size of the conical surface (values of D’C and hC) must be adjusted to consider the specific type of facing. For the temporary facing, the dimensions of the bearing plate and facing thickness must be considered. For the permanent facing, the dimensions of the headed-studs (or anchor bolts) must be considered. Figure 9-29 shows details of a typical headed-stud connector. The effective size of the conical surfaces must be considered as follows: Temporary facing (Figure 9-28a)

D’C = LBP + h hC = h

(9-27a) (9-27b)

where LBP = bearing plate length, h = temporary facing thickness.


9-59


DC D'C

WALER BAR (TYP)

LBP CONICAL FAILURE SURFACE

h

Shear Resistance RF/2

RF/2

45° (TYP)

h/2

To DDH

IDEALIZED SOIL REACTION

(a) TEMPORARY BEARING-PLATE CONNECTION

DC D'C SHS COMPOSITE CONICAL SURFACE

RF/2

LS 45° (TYP)

h

tP

hC

h/2

To

IDEALIZED SOIL DEFLECTION

DDH (b) PERMANENT HEAD-STUDDED CONNECTION

Figure 9-28. Punching Shear Failure Modes (after Lazarte, 2003). Permanent facing (Figure 9-28b)

D’C = min of (SHS + hC and 2 hC) hC = LS - tS + tP


9-60

(9-28a) (9-28b)


DH

tH LS

DS

Figure 9-29. Geometry of a Headed-Stud (after Lazarte, 2003). where: SHS = headed-stud spacing, LS = headed-stud length, tH = headed-stud head thickness, and tP = bearing plate thickness. Given the tensile force at the soil nail head, To and the punching shear capacity of the facing, the safety factor against facing punching shear can be defined as: FS FP =

R FP To

(9-29)

In general, a minimum factor of safety of 1.35 is adopted for static loads in temporary walls and 1.5 for static loads in permanent walls. Headed-Stud Tensile Capacity for Permanent Facing

The tensile capacity of the headed-studs (or anchor bolts) connectors providing anchorage of the nail into the permanent facing must be verified, as shown in Figure 9-28. The nail head capacity against tensile failure of the connectors, RHT, is computed as: R HT = N A S f y


9-61

(9-30)


where: N = number of connectors (or anchor bolts) in the connection (usually 4); AS = cross-sectional area of the headed-stud shaft; and fy = tensile yield strength of the headed-stud. Given the tensile force at the soil nail head To and the tensile capacity of the headed-studs, the factor of safety against tensile failure of the headed-studs can be defined as: FS HT =

R HT To

(9-31)

A minimum factor of safety of 1.8 is adopted for static loads in temporary walls and 2.0 for static loads in permanent walls if steel A307 is used for the headed-stud connector. If steel A325 is used for the headed-stud connector, a minimum factor of safety of 1.5 is adopted for static loads in temporary walls and 1.7 for static loads in permanent walls (Byrne et al., 1996). The headed studs may also exert excessive compressive stress on the concrete bearing surface. The compression on the concrete behind the head of the headed-stud is assured to be within tolerable limits if the following geometric constraints are met (ACI, 1998): AH ≥ 2.5 AS

(9-32)

tH ≥ 0.5 (DH - DS)

(9-33)

where: AH AS tH DH DS

= = = = =

cross-sectional area of the stud head; cross-sectional area of the stud shaft; head thickness; diameter of the stud head; and diameter of the headed-stud shaft.

To provide sufficient anchorage, headed-stud connectors should be extended at least to the middle of the section, while maintaining a minimum 2 in. of cover. To provide additional anchorage capacity, the headed studs should be long enough that the head is located behind the reinforcement.


9-62


When threaded bolts are used in lieu of headed-stud connectors, the effective cross-sectional area of the bolts must be employed in the equations above. The effective cross-sectional area, AE, of threaded anchors is computed as follows: π AE = 4

  0.9743    D E −   n t  

2

(9-34)

where: DE = effective diameter of the bolt core; and nt = number of threads per unit length. 9.2.6.5 Drainage Design Considerations Surface Water and Groundwater Control

Dewatering measures during construction include, as a minimum, the control of surface water runoff and subsurface flow associated with either perched water or localized seepage areas. A surface water interceptor ditch, excavated along the crest of the excavation and lined with concrete, applied during the shotcreting of the first excavation lift, is a recommended element for controlling surface water flows. Additionally, if the design engineer believes that the groundwater impacts are localized or short-term conditions, wells or well points installed beyond the length of the nails may be used temporarily to lower the groundwater table. However, this approach may result in much higher construction costs and delays. Long-Term Groundwater and Surface Water Control Geocomposite Drain Strips. These elements are strips of synthetic material approximately 12 to 16 in. wide. They are placed in vertical strips against the excavation face along the entire depth of the wall (Figure 9-30). The horizontal spacing is generally the same as the nail horizontal spacing. The lower end of the strips discharges into a pipe drain that runs along the base of the wall or through weep holes at the bottom of the wall. For highly irregular excavation faces, the placement of prefabricated drain strips against the excavated face is difficult and often impractical. In some cases, the prefabricated drain strips may be sandwiched between the shotcrete construction facing and the permanent CIP facing, with the drain placed over 2- to 3-in. diameter weep holes passing through the construction facing. The design engineer needs to provide explicit construction and inspection guidance for this type application, to assure that the performance of the drainage system is not impacted during FHWA NHI-07-071 Earth Retaining Structures

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Concrete Ditch

Groundwater Table Geodrain Strips

Weephole

Drains

Toe Drain

Figure 9-30. Drainage of Soil Nail Walls (after Lazarte et al., 2003).

installation of the shotcrete. If appropriate performance cannot be guaranteed, the effect of the groundwater table needs to be considered in the analysis. Additional discussion on geocomposite drains is provided in Section 5.5.10. Shallow Drains (Weep Holes). These are typically 12- to 16-in. long, 2- to 4-in. diameter PVC pipes discharging through the face and located where localized seepage is encountered or anticipated. Weep holes are also used as the terminating point of the vertical strip drains to allow any collected water to pass through the wall. Drain Pipes. Horizontal or slightly inclined drain pipes may be installed where it is necessary to control the groundwater pressures imposed on the retained soil mass. Drain pipes typically consist of 2-in. diameter PVC slotted or perforated tubes, inclined upward at 5 to 10 degrees to the horizontal. Drain pipes are typically longer than the length of the nails and serve to prevent groundwater from being in contact with the nails or the soil nail wall mass, as shown in Figure 9-31. The lengths of the drains depend on the application. To provide drainage of shallow or perched groundwater occurring erratically close to the facing, drain pipes with lengths varying from 1 to 1.5 ft, and in some cases, up to 3 ft can be installed. They are installed at a density of approximately one drain per 100 square ft of face. FHWA NHI-07-071 Earth Retaining Structures

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Drain pipes are typically deployed after nail installation to prevent potential intrusion of nail grout into the slotted pipes. The pipes typically exit through the face of the wall. The PVC pipe should be slotted, as shown in Figure 9-31. Although drain pipes are typically installed after nails are in place and the shotcrete is applied to avoid either grout or shotcrete from entering the drain, they can be applied prior to shotcrete application. In this case, a plug of dry-pack and temporary PVC caps must be used to prevent the shotcrete from coming into the drain hole and obstructing the drain slots or perforations. Permanent Surface Water Control. Permanent surface water control measures include installing an interception ditch behind the wall to prevent surface water runoff from infiltrating behind the wall or flowing over the wall edge. A vegetative protective cap may be also be used to reduce or retard water infiltration into the soil. Design Considerations

Drain pipes require long-term maintenance. Analysis of soil nail walls for long term conditions may need to take into consideration the potential for clogging. Clogging of horizontal drains and a corresponding increase in water pressure will reduce the factor of safety against global stability and/or sliding, and may adversely impact the internal stability by affecting soil/nail interaction. 9.2.6.6 Deformation Behavior of Soil Nail Walls During construction and after its completion, a soil nail wall and the soil behind it tend to deform outwards. The outward movement is initiated by incremental rotation about the toe of the wall, similar to the movement of a cantilever retaining wall. Most of the movement occurs during or shortly after excavation of the soil in front of the wall. Post construction deformation is related to stress relaxation and creep movement, which are caused by postconstruction moderate increases in tensile force in the soil nail described previously. Maximum horizontal displacements occur at the top of the wall and decrease progressively toward the toe of the wall. Vertical displacements (i.e., settlements) of the wall at the facing are generally small, and are on the same order of magnitude as the horizontal movements at the top of the wall. In general, horizontal and vertical displacements of the facing depend on the following factors: • • •

wall height, H, (deformation increases approximately linearly with height); wall geometry (a vertical wall produces more deformation than a battered wall); the soil type surrounding the nails (softer soil will allow more deformation);


9-65


END CAP

~

PERMANENT SHOTCRETE

2 in.) in. MIN 50 mm (2 DIA. SCH 40 PVC PIPE

10° - 15°

PROTECTIVE PVC CAP (NOTE 1)

TED

SLOT

ION SECT

(2 ft) 0.6 m ED ORAT PERF NON ECTION S

m 50 m .) (2 in MIN

NOTES 1. PROTECTIVE CAP NEEDS TO BE REMOVED AFTER FINAL SHOTCRETE IS APPLIED

10 ft 2. SPACING OF DRAINS IS TYPICALLY 3.3 m (10 ft)

TYPICAL DRAIN DETAIL

GEOCOMPOSITE DRAIN STRIP

SOIL NAIL WALL DRAIN PIPE

HARDBOARD

COLLECTOR DITCH FINISHED GRADE

GEOCOMPOSITE DRAIN STRIP

CONTINUOUS GRAVEL DRAIN FINISHED GRADE

FILTER FABRIC

FILTER FABRIC

~ WEEP HOLE GEOTEXTILE

WEEP HOLE

FOOTING DRAIN

TOE DRAIN

Figure 9-31. Typical Drain Pipe Details to Provide Groundwater Control in Soil Nail Walls (after Byrne et al., 1998).


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• • • • •

nail spacing and excavation lift heights (larger nail spacing and thicker incremental excavation lifts generate more deformation); global factor of safety (smaller FSG are associated with larger deformation); nail-length-to wall-height ratio (shorter nail lengths in relation to the wall height generates larger horizontal deformation); nail inclination (steeper soil nails tend to produce larger horizontal deformation because of less efficient mobilization of tensile loads in the nails); and magnitude of surcharge (permanent surcharge loading on the wall increases deformation).

Empirical data show that for soil nail walls utilizing a typical nail-length to wall-height ratio between 0.7 and 1.0, negligible surcharge loading, and typical global factors of safety (FSG) values of 1.5, the maximum long-term horizontal and vertical wall displacements at the top of the wall, δh and δv, can be estimated as follows: δ  δh =  h  × H  H i

(9-35)

where: (δh/H)I H

= is a ratio dependant on the soil conditions “i” indicated in the table below; = wall height.

The size of the zone of influence (Figure 9-32), where noticeable ground deformation may take place, is defined by a horizontal distance behind the soil nail wall (DDEF) and can be estimated with the following expression: D DEF = C (1 − tan α ) H

(9-36)

where: α = Wall batter angle; and C = Soil dependent coefficient indicated in Table 9-5. The movements shown above are considered to be relatively small and comparable to those obtained with braced systems and anchored walls. These estimates of deformations have essentially become recommended design values. The adopted tolerable deformation criterion is project-dependent and should consider not only the magnitude of deformation but also the extent of the area behind the wall that may be affected by wall movements. As a first FHWA NHI-07-071 Earth Retaining Structures

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DDEF EXISTING STRUCTURE h

L

V

DEFORMED PATTERN H

INITIAL CONFIGURATION

SOIL NAIL (TYP)

Figure 9-32. Deformation of Soil Nail Walls (after Byrne et al., 1996).

estimate, horizontal deflections greater than 0.005 H during construction should be a cause for concern, as they generally represent an upper limit of acceptable performance. When excessive deformations are considered to be likely with a certain wall configuration, some modifications to the original design can be considered. Soil nail wall deformations can be reduced by using a battered wall, installing longer nails in the top portion of the wall, using a higher safety factor, or even using ground anchors in conjunction with the soil nails. Table 9-5. Values of (δh/H)i and C as Functions of Soil Conditions.

Variable

δh/H and δv /H C


Weathered Rock and Stiff Soil 1/1,000 1.25

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Sandy Soil 1/500 0.8

Fine-Grained Soil 1/333 0.7


Additionally, some contractors have used soil nails that are grouted partially along their length and then partially tensioned to mobilize some of the nail tensile strength without soil mass deformation near the wall face. In these cases, after the tensioning is complete, the nails are fully grouted and the shotcrete is applied before the next lift is excavated. Post-construction monitoring of soil nail wall displacements indicates that movements tend to continue after wall construction, sometimes up to 6 months, depending on ground type. Typically, the post construction deformation increases up to 15 percent of the deformations observed soon after construction. As a result of this movement, additional tension is developed in the nails. In general, fine-grained soils of high-plasticity (i.e., approximately PI > 20) and high water contents (such that LI > 0.2) tend to incur deformation for longer periods of time. 9.2.6.7 Other Design Considerations Frost Protection

The formation of ice lenses in the vicinity of the soil nail wall facing in frost-susceptible soils may lead to the development of high loads on both the facing and the head of the nail. This phenomenon may result in damage to the facing. In situations where the facing is designed to resist frost damage, the nail or to the connection between the nail and the facing can still be impacted by frost. The magnitude of the impact to the facing/nail depends on the depth of frost penetration, the intensity and duration of the freeze period, the availability of water, and the stiffness of the facing. Kingsbury et al. (2002) report that the force in the nail head caused by frost action can be as high as 2.5 times larger as the maximum seasonal nail force without frost action. Increases in nail and facing loads should be anticipated in areas where frost durations are generally greater than one week, where frost susceptible soils are encountered near the face, and where the face is in close proximity to a source of water. Soils susceptible to frost action are those exhibiting the following characteristics: (1) more than 3 percent of the solids fraction is smaller than 1 mil for non uniform soils (i.e., Cu > 5), or (2) more than 10 percent of the solids fraction is smaller than 1 mil for uniform soils (i.e., Cu ≤ 5) (Casagrande, 1931). Cu is the coefficient of uniformity which can be obtained from grain size gradation tests. In general, frost-loading effects can be eliminated or significantly mitigated by increasing the wall thickness to approximately the depth of frost penetration. This can be done by placing porous backfill (e.g., gravel) or insulating material either between the temporary facing and permanent facing, or outside the permanent facing. Figure 9-33 shows examples of proposed frost protection details using Styrofoam insulation. Because a 1-in. thick Styrofoam FHWA NHI-07-071 Earth Retaining Structures

9-69


insulation board is generally considered to be equivalent to 1-ft thickness of gravel, appropriate protection against frost action can be obtained with a relatively thin layer of this material. External Loads

External loads may be applied at the top of the soil nail wall and may vary from relatively light highway appurtenance loads (e.g., roadway lighting supports) to significant loads (e.g., loads resulting from the integration of a relatively large cantilever retaining structure on top of the wall). For relatively light loading conditions, the external loads can be used to define additional shear forces and flexural moments in the section of the wall above the first row of nails. These loads are then added to the calculated facing loads for subsequent analysis. For more significant loads (e.g., loads applied by bridge abutments), it may be necessary to perform a full soil-structure interaction analysis to define how the additional facing and nail loads are distributed throughout the entire soil nail structure. The magnitude and distribution of the load transferred to the wall depends on the distance of the load to the wall and the type of load foundation (shallow or deep). The magnitude of these loads can be significantly increased if the structure is subject to seismic forces. Design of Support for Facing Dead Load

The weight of temporary facing must be supported by the installed nails or other supplementary means until compressive stresses develop at the facing-nail contact. This is particularly important for the facing of the initial excavation lifts that becomes unsupported when the next excavation lift is performed. For typical construction facings consisting of 4in. thick shotcrete, experience has shown that the soil nails will support the weight of the facing without major difficulties. For thicker applied shotcrete facings, support for the shotcrete facing weight by considering the shear capacity of the nails and the bearing capacity of the soils beneath the nails should be formally evaluated. The maximum thickness of shotcrete facing that can be supported in this manner is dependent on the strength of the soils. In competent ground, shotcrete facings up to 8- to 10-in. thick have been successfully supported. If necessary, support of the shotcrete facing weight may be achieved by the installation of additional short, steeply inclined reinforcing elements acting as compression struts. An analysis method for this case is provided in GEC No. 7 (Lazarte et. al, 2003).


9-70


CLASS C STONE (1' DEPTH)

TIMBER CAP 1- 2"X10" 2-2"X8"

NONWOVEN GEOTEXTILE

TIMBER BOARDS (USE ONE SIZE FOR ENTIRE WALL 2" x 8" OR 2" x 10")

RIGID INSULATION (6")

TIMBER POST (4" x 6", SPACED AT 5'-0" O.C.) VARIES (19' MAX)

RIGID INSULATION (6 1/4")

GEOCOMPOSITE DRAIN STRIP (AT 5'-0" O.C.)

REINFORCED SHOTCRETE (1") CONCRETE BLOCK (12" x 12" x 24")

RIGID INSULATION (6" x 5' LONG)

SOIL NAIL (TYP)

CRUSHED STONE PVC PIPE (6" DIA.)

FILTER FABRIC

a) Example of Soil Nail Wall with Frost Protection VARIES

10-FOOT LONG TONGUE & GROOVE TIMBER PLANK FACING (10-FOOT LONG)

2

1

1

2

2-INCH TONGUE & GROOVE INSULATION BOARD (2-INCH) GEOCOMPOSITE DRAIN STRIP (AT 5'-0" O.C.)

CAST-IN-PLACE STRUCTURAL CONCRETE WALL

15° - 25°

SHOCRETE TEMPORARY FACING (4-INCH)

4:1

PVC SLEEVE (WEEPHOLES AT 5' O.C.; GEOCOMPOSITE LOCATION)

6:1

PERMANENT SOIL NAIL (TYP)

UNDERDRAIN

CONCRETE FOOTING PERFORATED PVC PIPE COLLECTION TRENCH

b) Example of soil nail wall with Frost Protection

Figure 9-33. Examples of Frost Protection of Soil Nail Walls (after Lazarte et al., 2003).

Design for Seismic Forces

Considerations for the effects of seismic forces on soil nail walls is provided in Section 5.4.5 of GEC No. 7 (Lazarte et al., 2003). FHWA NHI-07-071 Earth Retaining Structures

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9.2.7

Design of Soil Nail Walls

9.2.7.1 Introduction The purpose of this section is to present a step-by-step generalized method for soil nail wall design. The major design steps are outlined in Table 9-6 and Steps 1 and 2 are discussed in other chapters.

Table 9-6. Design Steps for Soil Nail Walls.

Step 1.

Establish project requirements including all geometry, external loading conditions (temporary and/or permanent, seismic, etc.), performance criteria, aesthetic requirements, and construction constraints.

Step 2.

Evaluate site subsurface conditions and relevant properties of in situ soil and rock.

Step 3.

Develop initial soil nail wall design criteria.

Step 4.

Perform preliminary design using simplified design chart solution.

Step 5.

Evaluate external stability including global stability, sliding, and bearing capacity.

Step 6.

Evaluate internal stability including nail pullout resistance and tensile resistance.

Step 7.

Perform facing design including: (1) evaluation of nail head load; (2) selection of temporary and permanent facing materials and thickness; (3) evaluation of facing flexural resistance; (4) evaluation of facing punching shear resistance; and (5) evaluation of facing stud tensile resistance.

Step 8.

Estimate maximum lateral wall movements.

Step 9.

Design wall subsurface and surface drainage systems.

After completing the design, the design engineer will prepare soil nail wall specifications and recommendations for construction monitoring. 9.2.7.2 Step 3: Develop Initial Soil Nail Wall Design Criteria As part of this step, design factors of safety (see Table 9-7) and corrosion protection requirements are selected. In addition, the following design elements are selected to permit preliminary design calculations: • •

Wall layout (e.g., wall height, wall length, face batter); Soil nail vertical and horizontal spacing;


9-72


• • • •

Soil nail pattern on wall face (e.g., square, staggered, other irregular patterns); Soil nail inclination; Soil nail length and distribution; and Soil nail material type (e.g., selection of steel bar grade)

Table 9-7. Minimum Recommended Factors of Safety for the Design of Soil Nail Walls Using the ASD Method (after Lazarte, 2003). Minimum Recommended Factors of Safety Failure Mode

Resisting Component

Global Stability (long-term) External Stability

Internal Stability

Facing Strength

Notes:

Symbol

FSG

Temporary Structure

Permanent Structure

Seismic Loads(2) (Temporary and Permanent Structures)

1.35

1.5(1)

1.1

Static Loads(1)

1.2-1.3

(2)

Global Stability (excavation)

FSG

NA

Sliding

FSSL

1.3

1.5

1.1

Bearing Capacity

FSH

2.5 (3)

3.0 (3)

2.3 (5)

Pullout Resistance

FSP

2.0

1.5

Nail Bar Tensile Strength

FST

1.8

1.35

Facing Flexure

FSFF

1.35

1.5

1.1

Facing Punching Shear

FSFP

1.35

1.5

1.1

H.-Stud Tensile (A307 Bolt)

FSHT

1.8

2.0

1.5

H.-Stud Tensile (A325 Bolt)

FSHT

1.5

1.7

1.3

(1) For non-critical, permanent structures, some agencies may accept a design for static loads and long-term conditions with FSG = 1.35 when less uncertainty exists due to sufficient geotechnical information and successful local experience on soil nailing. (2) The second set of safety factors for global stability corresponds to the case of temporary excavation lifts that are unsupported for up to 48 hours before nails are installed. The larger value may be applied to more critical structures or when more uncertainty exists regarding soil conditions. (3) The safety factors for bearing capacity are applicable when using standard bearing-capacity equations. When using stability analysis programs to evaluate these failures modes, the factors of safety for global stability apply.

Wall Layout

Establish the layout of the soil nail wall, including: (1) wall height; (2) length of the wall; and (3) wall face batter (inclination typically ranges from 0 to 10). The evaluation of the wall layout also includes developing the wall longitudinal profile, locating wall appurtenances (e.g., traffic barriers, utilities, and drainage systems), and establishing ROW limitations.


9-73


Battered wall face can be selected to improve temporary face stability, as a battered face exerts smaller forces on the wall, thus requiring shorter soil nails. The material savings resulting from the use of shorter nails may offset the increased cost of soil excavation incurred to create the batter. A mild batter (i.e., less than 10 degrees) is usually provided for aesthetic reasons, especially around horizontal curves, and may be enough to ensure temporary face stability. A batter angle greater than 10 degrees can enhance stability. Soil Nail Vertical And Horizontal Spacing

Horizontal nail spacing, SH, is typically the same as vertical nail spacing, SV (Figure 9-34). Nail spacing ranges from 4 to 6.5 ft for conventional drilled and grouted soil nails, and may be as low as 1.5 ft for driven nails. This reduced spacing for driven nails is required because driven soil nails develop bond strengths that are lower than those for drilled and grouted nails. A soil-nail spacing of 5 ft is routinely used and is preferred for conventional drilled and grouted soil nails. Soil nail spacing may be affected by the presence of existing underground structures. Soil nail spacing in horizontal and vertical direction must be such that each nail has an influence area SH × Sv ≤ 40 ft2 ft. The design engineer should specify a minimum horizontal soil nail spacing of about 3.3 ft. Design forces from global stability analysis and facing design are affected by soil nail spacing. In general, the larger the spacing, the greater the design forces. The purpose of the minimum nail spacing is to reasonably ensure that group effects between adjacent soil nails are minimized due to potential nail intersection as a result of drilling deviations. Group effects reduce the load-carrying capacity of individual soil nails. The maximum soil nail spacing should also be specified. The purpose of a maximum spacing [usually about 6.5 ft] is to provide for a soil nail system that is relatively easy to construct and that effectively supports the lateral earth pressures and imposed surcharge loads. Soil Nail Pattern on Face

The soil nail pattern is commonly one of the following (Figure 9-34): (1) square (rectangular); (2) staggered in a triangular pattern; and (3) irregular (at limited locations). A square pattern results in a column of aligned soil nails, and facilitates easier construction of vertical joints in the shotcrete facing (or easier installation of precast concrete panels). Also, a square pattern enables a continuous vertical installation of geocomposite drain strips behind the facing to be easily constructed. In practice, a square pattern is commonly adopted.


9-74


GEOCOMPOSITE DRAINAGE STRIPS

SH

SVO < SV NAIL #1 1 ftM 0.30 (TYP)

SV

i

BOTTOM OF EXCAVATION LIFTS

N

BOTTOM OF EXCAVATION

SVN < SV

NAIL SQUARE PATTERN GEOCOMPOSITE DRAINAGE STRIPS

SH / 2

SH / 2

SVO

1 ftM 0.30 (TYP)

SV

SVN

NAIL STAGGERED PATTERN

Figure 9-34. Soil Nail Patterns on Wall Face (after Lazarte et al., 2003). A staggered soil nail pattern results in a more uniform distribution of earth pressures in the soil mass. This effect is beneficial because an enhanced soil arching effect is achieved. This method should be considered in cases where marginally stable soils are present because such soils have less margin to redistribute loads. The main disadvantage of the use of a triangular FHWA NHI-07-071 Earth Retaining Structures

9-75


pattern is that it makes installation of geocomposite drain strips more complicated. In particular, it can be difficult to establish a vertically continuous drain system to the footing drain, especially for higher walls. The use of uniform nail spacing is beneficial because it simplifies construction and quality control. However, due to project-specific geometric constraints, nail spacing may need to be irregular, with reduced spacing at some locations; for instance, in areas where the bottom of the excavation or the top of the wall is not horizontal. In such cases, it is more convenient to install one or two nail rows parallel to the non-horizontal edge and then establish a transition zone where nails have a closer vertical spacing until a horizontal nail row is achieved (Figure 9-35a). It is also customary to reduce horizontal spacing at the vertical edges of the wall to accommodate transition zones (Figure 9-35a). Soil Nail Inclination

Soil nails are typically installed at an inclination ranging from 10 to 20 degrees from horizontal with a typical inclination of 15 degrees. This recommended range of soil nail inclination assures that grout will flow readily from the bottom of the hole toward the nail head for typical borehole and soil nail dimensions and conventional grout mixtures. Steeper nail inclinations may be required, particularly for the upper row of nails, if a significantly stronger soil zone is located at a greater depth and a more effective anchorage in the stiffer layer is desired. Such evaluations can be readily made during design. Nail inclination smaller than about 10 degrees should not be used because the potential for creating voids in the grout increases significantly. Voids in the grout will affect the load capacity of the nail and reduce the overall corrosion protection provided by the grout. Project conditions may, however, require that other nail inclinations be used. For example, Figure 9-35b shows a case in which utilities or other underground structures are located within the proposed soil nail zone. In most cases, this situation only occurs for the upper first and second rows of nails. Another situation where different nail inclinations may be used is at exterior wall corners. To avoid intersecting nails behind exterior corners of a wall, nail inclination on one side of the corner could be installed with a different inclination. An alternative layout for exterior corners is to splay the nails on a plan view (Figure 9-35c). Overhead space restrictions may require that the nail inclination be smaller than 15 degrees. This might be the case for road widening at embankment bridge abutments. Logistical limitations due to location of nailing equipment (i.e., operating at the bottom of a narrow excavation) may require a steeper nail inclination.


9-76


TOP OF WALL NAIL LOCATION (TYP)

SVO

SV SH

ALL S v AND S h MUST BE SMALLER THAT MAXIMUM SPECIFIED

BOTTOM OF EXCAVATION

(a) EXAMPLE OF NAIL ARRANGEMENT FOR NONHORIZONTAL GROUND

UTILITIES

1

NAILS 1 AND 2 ARE OFFSET HORIZONTALLY TO AVOID INTERSECTION

>15°

2

MANHOLE

<15°

SV

SH SH

15°

SH

SH

SH SH

SH

SH

CROSS-SECTION

PLAN VIEW

(b) CHANGE OF NAIL DECLINATION AROUND UTILITIES

(NOT TO SCALE)

(c) NAIL SPLAYING IN CORNERS

Figure 9-35. Varying Soil Nail Patterns (after Lazarte, 2003).

The effect of nail inclination should be considered in global and local stability analyses of the soil nail wall system because stability factors of safety for the system, particularly for sliding wedge analyses in the upper portion of the wall, can decrease significantly as the nail inclination increases below the horizontal.


9-77


Soil Nail Length and Distribution

The distribution of soil nail lengths in a soil nail wall can be selected as either uniform (i.e., only one nail length is used for the entire wall), or variable, where different nail lengths may be used for individual soil nail levels within a wall cross section. Additional information on nail distribution is provided below. •

Uniform Nail Length: When the potential for excessive wall deformation is not a concern (e.g., soil nail walls constructed in competent ground or in an area without nearby structures), it is beneficial to select a uniform length distribution because it simplifies construction and quality control. Additionally, a slightly smaller total length of nails is obtained with a uniform soil length pattern. This pattern provides commonly a high sliding stability safety factor. Uniform patterns should be used in most projects.

•

Variable Nail Lengths: Occasionally, a variable nail length distribution may be used if wall deformations need to be controlled. The global equilibrium and deformation pattern of a system with different nail lengths will be different from a system with uniform nail lengths. Field measurement data from constructed soil nail walls indicate that wall displacements can be significantly reduced if the nail lengths in the upper two-thirds to three-quarters of the wall height are greater than those in the lower portion. Placing additional reinforcement (i.e., length of soil nail) near the top of the wall will provide more resistance to wall movement at the critical areas near the top of the wall. As lower nails are shorter in non-uniform length patterns, this distribution tends to produce a lower sliding stability safety factor.

Performance of soil nail walls has shown that larger displacements are observed when the upper nails are too short. The deformations in soil nail walls can be significantly reduced when nails at the top of the structure are longer than required by stability analysis. In general, the higher the global factor of safety of a soil nail wall, the smaller the wall deformations. Nail lengths have been installed successfully with a uniform nail length in the upper twothirds to three-quarters of the wall, with progressively shorter nails to a minimum value, not smaller than 0.5 H (H is the wall height), at the bottom of the wall in dense cohesionless soils that provide relatively large sliding stability. In general practice, nail length in the lower rows should never be shorter than 0.5 H. Nail lengths less than 0.5 H will not likely satisfy sliding stability requirements. In all cases, and especially where reducing the nail lengths in FHWA NHI-07-071 Earth Retaining Structures

9-78


the lower reaches of the wall are considered, stability analysis considering sliding need to be performed as part of a detailed design. In general, variable nail lengths result in a more complicated installation and require more nail materials. Nevertheless, as many soil nail projects are specified based on performance criteria, contractors may prefer to use longer nails in the upper rows to reduce deflections. Project specifications must provide ROW constraints, locations of underground utilities and substructures (or requirements that the contractor locate these), and specific deformation criteria (i.e., maximum wall deflection and location where this deflection is to be measured). Based on the discussion presented in this section, the following recommendations are made concerning soil nail length and distribution. •

Select uniform length pattern whenever possible.

•

Select longer nails than required by the target factor of safety as a means to reduce wall deformations in the upper portions of the wall.

•

Avoid the use of “short” nails in top portion of wall.

•

Avoid the use of too “short” nails in lower portion of wall. Evaluate if shorter nails in bottom rows installed in competent ground satisfy sliding stability requirements. Shorter nails at the bottom should be not smaller than 0.5 H.

•

Non-uniform nail length patterns may be used if soil layers with very dissimilar conditions are encountered.

For feasibility evaluations, soil nail length can be initially assumed to be 0.7 H, where H is the height of the wall. The length of the nails may be greater than 0.7 H if large surcharge loads are expected or if the wall is very high [greater than 30 ft high.] Soil Nail Materials

Select appropriate grade of steel for the soil nail bar. For most applications Grade 60 steel is used, however information on the selection of steel grade can be found in Section 9.2.3.1. Soil Properties

The ultimate bond strength for the grout-ground interface can be selected using Table 9-3.


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9.2.7.3 Step 4: Preliminary Design Using Simplified Charts Introduction

Nail length, diameter, and spacing typically control external and internal stability of a soil nail wall. Therefore, these parameters may be adjusted during design until all external and internal stability requirements are satisfied. A series of charts was developed by Lazarte et al. (2003) as a design aid to provide preliminary nail length and maximum tensile forces (Figures 9-36 through 9-41). The charts were developed using the computer program SNAIL, which was selected because it is public domain software, readily available, and free of charge. In preparing these charts, the following main assumptions were made: • • • • •

homogenous soil; no surcharge; no seismic forces; uniform length, spacing and inclination of nails; and no groundwater.

When the conditions of a new analysis case do not match the assumptions listed above, it is recommended that interpolations or extrapolations be made to estimate the soil lengths from these charts. Alternatively, the use of a preliminary nail length between 0.7 to 1.0 times the wall height can be made. The upper range of soil nail length is used for less favorable soil conditions, wall heights greater than 30 ft, and where large surcharge loads need to be resisted by the wall. The charts were developed for different values of face batter (α); backslope (β); effective friction angle (φ’); and ultimate bond strength (qu). Table 9-8 presents the geometric and material conditions used for the development of the design charts. The first type of chart was developed to evaluate the nail length (Figures 9-36a through 941a) for combinations of α and β. Using these charts, the required nail length, L, (normalized with respect to the wall height, H) to achieve a global safety factor of 1.35 is obtained as a function of the normalized allowable pullout resistance (µ). The normalized allowable pullout resistance is defined as:


9-80


Face Batter α=0

L

Backslope β=0

µ = (qaDDH)/(γ SH SV) qa = qu/FSP

H

c* = c / γH

γ, c, φ

SV

c* = 0.02 t max-s = Normalized Maximum Design Force in Nails = Tmax-s/γ H SH SV FS=1.35 DDH= 100 4 in.mm

For other FS, c*, and DDH, see Figure B7

1.5 Friction Angle (degrees) 31 35 39 27

L/H

1

0.5

Normalized Design Nail Force, tmax-s

(a) 00 0.4

0.3

0.2

0.1

(b)

0 0

Nail forces for FSG = 1.0 0.1

0.2

Normalized Bond Strength, µ =

0.3

0.4

q aD DH γ SH SV

Figure 9-36. Batter 0o – Backslope 0o (after Lazarte et al., 2003).


9-81


Backslope β = 10O

Face Batter α=0

µ = (qaDDH)/(γ SH SV)

L

qa = qu/FSP

H

SV

γ, c, φ

c* = c / γH

c* = 0.02

t max-s = Normalized Maximum Design Force in Nails = Tmax-s/γ H SH SV

FS=1.35 4 in.mm DDH= 100


1.5

Friction Angle (degrees) 27 31 35 39

L/H

1

0.5

Normalized Max. Design Force, tmax-s

(a) 00 0.4

0.3

0.2

0.1 NOTE: Nail forces are for FSG = 1.0

(b) 0 0

0.1

0.2


0.3

0.4

q aD DH γ SH SV



9-82


Backslope β=0

Face Batter α = 10O


L

qa = qu/FSP

SV

H

γ, c, φ

c* = c / γH

c* = 0.02


FS=1.35 4 in.mm DDH= 100



L/H

1

0.5


(a) 00 0.4

0.3

0.2

0.1 NOTE: Nail forces are for FSG = 1.0

(b) 0 0

0.1

0.2


0.3

0.4

q aDDH γ SH SV



9-83


Backslope β = 10O



L

qa = qu/FSP

SV

H

γ, c, φ

c* = c / γH

c* = 0.02


FS=1.35 DDH= 4 100 in.mm



L/H

1

0.5

Normalized Max. Design Nail Force, tmax-s

(a) 00 0.4

0.3

0.2

0.1

NOTE: Nail forces are for FSG = 1.0

(b) 0 0

0.1

0.2

0.3

0.4

q aD DH Normalized Bond Strength, µ = γ SH SV



9-84


Face Batter α=0

Backslope β = 30O µ = (qaDDH)/(γ SH SV)

L

qa = qu/FSP

H

SV

γ, c, φ

c* = c / γH

c* = 0.02

t max-s = Normalized Maximum Design Force in Nails

FS=1.35 DDH=4100 in. mm

= Tmax-s/γ H SH SV For other FS, c*, and DDH, see Figure B7

1.5 Friction Angle (degrees) 39

L/H

1

0.5

(a) Normalized Max. Design Force, tmax-s

00 0.4

0.3

0.2

0.1


(b) 0 0

0.1

0.2


0.3

0.4

q aD DH γ SH SV



9-85



Backslope β = 30O µ = (qaDDH)/(γ SH SV)

L

qa = qu/FSP c* = c / γH

γ, c, φ

SV

H

c* = 0.02 FS=1.35

t max-s = Normalized Maximum Design Force in Nails

DDH=4100 in. mm

= Tmax-s/γ H SH SV For other FS, c*, and DDH, see Figure B7

1.5 Friction Angle (degrees) 39

L/H

1

0.5


(a) 00 0.4

0.3

0.2

0.1


(b)

0 0

0.1

0.2


0.3

0.4

qaD DH γ SH SV



9-86


Table 9-8. Variable Parameters Used in Design Charts (Lazarte et al., 2003). Parameter

Units

Values

Face Batter

degrees

0, 10

Backslope

degrees

0, 10, 20, 30

Effective Friction Angle

degrees

27, 31, 35, 39

Ultimate Bond Strength

psi

8,15,25,40,100

µ =

q u D DH FSP γ SH SV

(9-37)

where FSP is the factor of safety against pullout (typically 2.0); DDH is the drillhole diameter; γ is the total unit weight of the soil behind the wall; and SH and SV are the horizontal and vertical nail spacing, respectively. The nail lengths in these charts were computed based on the most critical failure surface (i.e., considering base and toe failures) for the selected geometry and material properties, and assuming that failure of the nail (i.e., tensile breakage) and/or failure of the facing would not take place. Therefore, the pullout failure is implicitly assumed. Equation 9-37 is based on a drillhole diameter of 4 in. Also, the use of Equation 9-37 inherently assumes that the soil has a cohesion intercept c′ such that c* = c′/γH = 0.02. If the drillhole diameter or cohesion intercept values being considered are different than the assumptions stated here, then adjustments to the calculated nail length and maximum tensile forces are made in the final step. These adjustments are discussed subsequently. The second type of charts (Figures 9-36b through 9-41b) provides the corresponding maximum normalized design tensile force of all nails (tmax-s) as a function of µ calculated for a global safety factor of 1.0. The maximum normalized design tensile force in the bar is defined as: t max − s =

Tmax − s γ H S H SV

(9-38)

With tmax-s read from the design charts, the maximum nail tensile force, Tmax-s can be calculated using Equation 9-38. These design charts are developed for the case in which all nail bars are the same length. These design charts do not provide information on the distribution of tensile load in individual soil nails or the maximum load in any particular nail. FHWA NHI-07-071 Earth Retaining Structures

9-87


Preliminary Design Procedure

A step-by-step procedure for preliminary design using the charts shown on Figures 9-36 through 9-41 is presented in this section. 1.

For a specific project application, evaluate batter (α), backslope (β), effective friction angle (φ′), and ultimate bond strength (qu). Calculate normalized pullout resistance (µ) using Equation 9-37.

2.

Obtain normalized length (L/H) from the first set of charts (Figures 9-36a through 9-41a).

3.

Obtain normalized force (tmax-s ) from the second set of charts (Figures 9-36b through 9-41b).

4.

Using Figure 9-42, evaluate correction factors for: (a) normalized length to account for a drillhole diameter other than 4 in. (correction factor C1L), (b) a c* value other 0.02 (correction factor C2L), and (c) a global factor of safety other than 1.35 (correction factor C3L).

5.

Using Figure 9-42, evaluate correction factors for normalized maximum nail force to account for: (a) a drillhole diameter other than 4 in. (correction factor C1F), and (b) a c* value other 0.02 (correction factor C2F).

6.

Apply correction factors to normalized length and/or normalized force. Calculation method is provided on Figure 9-42.

7.

Multiply the normalized length by the wall height to obtain the soil nail length.

8.

Calculate the maximum design load in the nail Tmax-s using the value of tmax-s and Equation 9-38.

9.

Calculate the required cross-sectional area (At) of the nail bar according to: At =

Tmax − s FST fy

(9-39)

where fy is the steel yield strength and FST is the factor of safety for nail bar tensile strength. 10. Select closest commercially available bar size using Table 9-9 that has a crosssectional area of at least that evaluated in the previous step. 11. Verify that selected bar size fits in the drillhole with a minimum grout cover thickness of 1 in. 12. If the length and/or nail diameter are not feasible, select another nail spacing and/or drillhole diameter, recalculate the normalized pullout resistance, and start the process again. FHWA NHI-07-071 Earth Retaining Structures

9-88


Corrections Of Normalized Soil Nail Length L L (corrected ) = C1L × C 2L × C 3L × ( from charts for D DH = 4 in., c * = 0.02 , FS G = 1.35) H H where: C1L = Correction for Drillhole Diameter C2L = Correction for Soil Cohesion C3L = Correction for Global Factor of Safety

see chart (a) below see formula (b) below see formula (c) below

Corrections of Normalized Maximum In-Service Nail Force

t max - s (corrected ) = C1F × C 2F × t max -s ( from charts for DDH = 4 in., c * = 0.02) C1F= Correction for Drillhole Diameter C2F= Correction for Soil Cohesion

see chart (a) see formula (b)

(a) Correction for Drillhole Diameter 21 C

1.8 0.9

0.9 1.8

C1F1F

1.6 0.8

0.8 1.6

1.4 0.7

0.7 1.4

C1L

0.6 1.2 0.5 1 100 4

5

150 6

7

200 8

1.2 0.6

C1L

9

250 10

Drillhole Diameter, DDH (mm)

11

Correction for Correction forLength, Force,C 1L C1F

Correctionfor for Length, Force, C 1FC1L Correction

21

10.5 300 12

Drillhole Diameter, DDH (in.)

Figure 9-42. Correction Factors for Use in Design Chart Solutions (after Lazarte, 2003). FHWA NHI-07-071 Earth Retaining Structures

9-89


Table 9-9. Threaded Bar Properties [ASTM A615, Grades 60 and 75 ksi] Nominal Bar Designation English

Cross-Sectional Area in.2

Nominal Unit Weight lbs/ft

Max. Diameter w/Threads in.

#6

0.44

1.50

0.86

#7

0.60

2.04

0.99

#8

0.79

2.67

1.12

#9

1.00

3.40

1.26

#10

1.27

4.30

1.43

#11

1.56

5.31

1.61

#14

2.25

7.65

1.86

ASTM Grade English 60 75 60 75 60 75 60 75 60 75 60 75 60 75

ksi

Max. Axial Load kips

60 75 60 75 60 75 60 75 60 75 60 75 60 75

26.4 33.0 36.0 45.0 47.4 59.3 60.0 75.0 76.2 95.3 93.6 117.0 135.0 168.8

Yield Strength

9.2.7.4 Step 5: Evaluate External Stability For this step (and all subsequent design steps), it is necessary to perform a final design in which the actual wall geometry, stratigraphy, loads, variation of engineering parameters (if present), and other conditions are considered. The preliminary design procedure described earlier should not replace the findings and results obtained with the final design presented herein. Global Stability

•

Select a well-established computer program for design of soil nail walls that considers heterogeneous soils, groundwater, general loading conditions, seismic forces, and diverse nail characteristics. In this section, the computer program SNAIL (see Lazarte et al., 2003 for a description of input and output capabilities of SNAIL) is selected.

•

Select the factor of safety against pullout failure (FSp) (SNAIL requires the value of bond stress factor (BSF) equal to 1/FSp (e.g., for FSP = 2, the corresponding BSF is 0.5).


9-90


•

Use the nail diameter calculated in preliminary design (if available) as the input nail hole diameter for SNAIL. If no preliminary design was performed, assume a relatively large nail diameter [e.g., > 4 in.] to ensure that pullout failure controls the design. Also, select a high punching shear strength [e.g., RFP > 220 kip] to ensure that pullout failure controls the design. It is noted that the tensile strength and punching failure mechanisms are explicitly considered subsequently in this chapter.

•

For the first SNAIL analysis, use nail length estimated previously (or calculated in the preliminary design) and perform global stability analysis using SNAIL.

•

After selecting an initial nail length, perform the following iterative procedure using SNAIL: (1) calculate the global factor of safety using the selected nail length; (2) compare the calculated global factor of safety to the recommended minimum factor of safety; and (3) increase or decrease the nail length if the calculated factor of safety is lower or higher than the recommended value and start the process again.

•

If the length of the nail needs to be reduced without reducing the factor of safety, then increase the nail hole diameter or reduce the nail spacing.

Sliding Stability

Evaluate the potential for sliding failure using the equations and procedures outlined below: •

Calculate the horizontal resisting forces (ΣR) as follows (see Figure 9-20). ∑ R = c b B + (W + Q D + PA sin β ) tan φb

(9-40)

where cb soil cohesion strength along the base, BL is the length of horizontal failure surface where cb is effectively acting, W is the weight of soil nail block, QD is permanent portion of total surcharge load QT, PA is active lateral earth pressure, β is backslope angle, and φb angle of internal friction of the base. •

Calculate the horizontal driving forces as follows: o If necessary, calculate the equivalent backslope angle (βeq in Figure 9-20); o Evaluate the active earth pressure coefficient KA using Rankine or Coulomb theories. o Calculate the active lateral earth force (PA) as:


9-91


PA =

γ H 12 KA 2

(9-41)

where Hl is the effective height over which the earth pressure acts [Hl = H + (β + tan α) tan βeq]. o Assume that the active lateral earth force is applied a distance of H1/3 from the elevation of the bottom of the soil nail wall (Figure 9-20); and o Calculate the horizontal driving force (ΣD) as: ∑ D = PA cos β

•

Evaluate the factor of safety against sliding failure (FSLS) as: FSSL =

•

(9-42)

∑R ∑D

(9-43)

If the factor of safety against sliding is lower than the specified minimum, increase the length of the lower nails and reevaluate sliding stability.

Bearing Capacity

If soil nail wall is constructed in soft soils, evaluate the factor of safety against bearing capacity failure (FSH) using Equation 9-44. FSH =

Su N c S   H eq  γ − u  B'  

(9-44)

where Su is undrained shear strength of soil, Nc is bearing capacity factor (see Figure 921), γ is the unit weight of soil behind wall, Heq is equivalent wall height [Heq = H + ∆H], B’ is width of influence [B’ = Be /

2 ].

9.2.7.5 Step 6: Evaluate Internal Stability Nail Pullout Resistance (Rp)

The procedure described in “Global Stability” under external stability calculations automatically satisfies the condition for pullout resistance (i.e., Tmax ≤ Rp/FSp). In a FHWA NHI-07-071 Earth Retaining Structures

9-92


SNAIL analysis, the calculated nail lengths (corresponding to an acceptable global factor of safety for a given critical failure surface) are based on pullout capacity values, which have already been reduced by the factor of safety with respect to pullout, FSP. Nail Tensile Resistance (RT)

•

The SNAIL analysis provides (at the end of the output file) the average nail tensile force calculated for a case with FSG = 1.0 (Tavg-s).

•

Calculate the average nail load (Tavg) as the sum of the individual nail forces calculated by SNAIL divided by the number of nails in the analyzed cross section.

•

The SNAIL analysis provides the maximum nail tensile force (Tmax).

•

Calculate the maximum nail tensile force (Tmax-s) as:  T Tmax − s =  avg − s  Tmax  Tavg   

(9-45)

•

Select the nail yield strength (fy).

•

With Tmax-s, fy and the factor of safety against tensile failure (FST), calculate the required cross sectional area of a steel nail bar (At) according to: At ≥

Tmax −s FST fy

(9-46)

•

Select the closest commercially available nail bar size (see Table 9-9)

•

Verify that the bar fits in the drillhole subject to a minimum grout cover thickness of 1 in. and the required corrosion protection.

9.2.7.6 Step 7: Perform Facing Design Calculate design nail head tensile force at the face (To)

To = Tmax -s [0.6 + 0.057 (S v [ft ] − 3)]


9-93

(9-47)


Select wall facing thickness

Temporary facing thickness (h): [e.g., 4, 6, 8 in.]. Permanent facing thickness (h): [e.g., 8 in.]. Select soil nail wall materials

Steel reinforcement: Grade (fy), WWM dimensions, and Rebar dimensions (see Tables 9-9 and 9-10) Table 9-10. Welded Wire Mesh Dimensions (after WRI, 2001)

Mesh Designation (1), (2)

Wire Cross-Sectional Area per Unit Length(3)

Weight per Unit Area

English (in.2/ft)

English (lbs/ft2)

0.042 0.060 0.087 0.120 0.028 0.040 0.058 0.080

3.1 4.4 6.2 8.8 2.1 3.0 4.2 5.8

English (in. x in.-in2/100 x in.2 /100) 4x4 - W1.4xW1.4 4x4 - W2.0xW2.0 4x4 - W2.9xW2.9 4x4 - W4.0xW4.0 6x6 - W1.4xW1.4 6x6 - W2.0xW2.0 6x6 - W2.9xW2.9 6x6 - W4.0xW4.0

Notes: (1) The first two numbers indicate the mesh opening size, whereas the second pair of numbers following the prefixes indicates the wire cross-sectional area. (2) Prefix M indicates metric units. Prefix W indicates plain wire. If wires are pre-deformed, the prefix D is used instead of W. (3) This value is obtained by dividing the wire cross-sectional area by the mesh opening size.

Concrete Strength (fc’): 3,000 and 4,000 psi. Select headed-stud characteristics (see Table 9-11). Select bearing plate geometry: min. 8 × 8 in. and 0.75 in. thick.


9-94


Verify facing flexural resistance (RFF) for temporary and permanent facing

a. The minimum reinforcement ratio is calculated as:

ρ min [%] = 0.24

f c' [psi]

(9-48)

f y [ksi]

The maximum reinforcement ratio is calculated as:  f c' [psi]  90   ρ max [%] = 0.05 f y [ksi]  90 + f y [ksi] 

(9-49)

b. Select reinforcement area per unit length of WWM for temporary/permanent facing at the nail head (an) and at mid-span (am) in both the vertical and horizontal directions. Typically, the amount of reinforcement at the nail head is the same as the amount of reinforcement at the mid-span (i.e., an = am) in both vertical and horizontal directions. For temporary facing, if waler bars are used at the nail head in addition to the WWM, recalculate the total area of reinforcement at the nail head in the vertical direction (see Equation 9-50) and horizontal direction (change Equation 9-50 appropriately). an = am +

A vw SH

(9-50)

c. Calculate the reinforcement ratio (ρ) at the nail head and the mid span as:

ρn =

an 100 b h/2

(9-51)

ρm =

am 100 b h/2

(9-52)

d. Verify that the reinforcement ratio of the temporary and permanent facing at the midspan and the nail head are greater than the minimum reinforcement ratio (i.e., ρmin ≤ ρ), otherwise increase the amount of reinforcement (an and/or am) to satisfy this criterion. e. Verify that the reinforcement ratio of the temporary and permanent facing at the midspan and the nail head are smaller than the maximum reinforcement ratio (i.e., ρ ≤ FHWA NHI-07-071 Earth Retaining Structures

9-95


Table 9-11. Headed-Stud Dimensions (Metric and English Units) (after Byrne et al., 1996) Nominal Length

Head Diameter

Shaft Diameter

Head Thickness

Ls

DH

DS

tH

in.

in.

in.

in.

1

4.125

0.5

0.25

4.7

4.0

0.75

3

4.125

0.75

0.38

7.1

4.0

0.75

3

6.125

0.75

0.38

7.1

4.0

0.75

1

/2 x 41/8

4.125

1

0.5

7.9

4.0

0.62

1

/2 x 55/16

5.3125

1

0.5

7.9

4.0

0.62

1

/2 x 61/8

6.125

1

0.5

7.9

4.0

0.62

5

/8 x 69/16

7.875

1.3

0.625

7.9

4.0

0.50

3

/4 x 311/16

15.5

1.3

0.750

9.5

2.8

0.75

3

4.1875

1.25

0.75

9.5

2.8

0.75

3

5.1875

1.25

0.75

9.5

2.8

0.75

3

6.1875

1.25

0.75

9.5

2.8

0.75

7

4

1.4

0.875

9.5

2.5

0.75

7

5

1.4

0.875

9.5

2.5

0.75

7

6

1.4

0.875

9.5

2.5

0.75

Headed-Stud Size

/4 x 41/8 /8 x 41/8 /8 x 61/8

/4 x 43/16 /4 x 53/16 /4 x 63/16 /8 x 43/16 /8 x 53/16 /8 x 63/16

Head Thickness/ Head Area/ Shaft Area (Head Diameter-Shaft Diameter) in.

Nominal length indicated is before welding.

• •

For DS ≤ 1/2”, LS is approximately 1/8” shorter after welding. For DS > 5/8”, LS is approximately 3/16” shorter after welding. DSH

tSH LS


DSC

9-96


ρmax), otherwise reduce the amount of reinforcement (an and/or am) to satisfy this criterion. f. Using Table 9-4, select factor CF (typically 1 for permanent facings) to take into account the non-uniform soil pressures behind facing. g. Calculate facing flexural resistance (RFF) for the temporary and permanent facing as:

R FF

 230 × C F × (a vn + a vm ) [in 2 /ft]× h[ft]    [kip] = min imum of    230 × C × (a + a ) [in 2 /ft]× h[ft]  F hn hm  

(9-53)

Alternatively, calculate the total reinforcement ratio ρtot as: ρ tot = ρ n + ρ m

(9-54)

and use Table 9-12a (interpolate for ρtot if necessary) and calculate RFF for the temporary/permanent facing. h. Using the recommended factor of safety for facing flexure (FSFF), verify that the temporary and permanent facing flexural resistance is higher than nail head tensile force (To): RFF ≥ FSFF To

(9-55)

i. If the capacity of the temporary and/or permanent facing is insufficient, increase the thickness of facing, steel reinforcement strength, concrete strength, and/or amount of steel and repeat the facing flexural resistance calculations. Verify Facing Punching Shear Resistance (RFP) a. Temporary Facing: With the values of concrete strength (fc’), facing thickness (h), and bearing plate length (LBP), use Table 9-12b to obtain the punching shear resistance (RFP) for the temporary facing. b. Permanent Facing: With the values of concrete strength (fc’), headed-stud geometric characteristics and spacing, use Table 9-12c to obtain the punching shear resistance (RFP) for the permanent facing.


9-97


Table 9-12. Facing Resistance For Various Failure Modes (after Lazarte et al., 2003).


9-98


(a) FACING RESISTANCE FOR FLEXURE, RFF (TEMPORARY FACING) h(1)

Nail Spacing Ratio(2)

ρtot (%) (4) 0.5

1.0

2.0

[-] RFF in kip 0.67 12 24 48 4 1 18 36 71 0.67 20 40 81 6 1 30 60 120 0.67 24 48 95 8 1 36 71 143 Based on a reinforcement yield strength, fy, of Grade 60. For fy = Grade 75, multiply the values in the table by 1.24. For permanent facing, for h = 4 in., divide RFF by 2; for h = 6 in., divide RFF by 1.5; for h = 8 in., use same RFF.

in.

ksi 3 4 3 4 3 4

4 6 8

in.

kip

Notes:

(b) FACING RESISTANCE FOR SHEAR PUNCHING, RFP (TEMPORARY FACING) f’c(3)

RFH

3/8 28 ½ 48 5/8 75 ¾ 108 7/8 146 Based on 4 headed-studs and a yield strength, fy, of Grade 60. For fy = 5Grade 75, multiply the values in the table by 1.24

in.

h(1)

Head Shaft Diameter, DS

Bearing Plate Length, LBP in. 8 9 10 RFF in kip 32 35 37 37 40 43 56 60 64 65 69 74 85 91 96 99 105 111

(1)

h is the facing thickness.

(2)

Nail space ratio is the lowest of either SV/SH or SH/SV, resulting in a value less than or equal to 1.

(3)

f’c is the concrete nominal compressive strength.

(4)

ρtot is the total reinforcement ratio calculated as ρtot = ρn+ρm, where ρn and ρm are the nail head and mid-span reinforcement ratios, respectively. ρI = aij/0.5h, where aij = cross sectional area of reinforcement per unit width in “i” direction (vertical or horizontal) and at location “j” (nail head or midspan).

(5)

(c) FACING RESISTANCE FOR SHEAR PUNCHING, RFP (PERMANENT FACING)

hc = Ls – tH + tP where: Ls is the effective headedstud length (Table 9-11); tP is the bearing plate thickness [typically 0.75 in.]; tH is the headed-stud head thickness (Table 9-11).

Headed Stud Spacing, SHS in. 4 5 6 in. ksi RFP in kip 3 21 21 21 4 4 25 25 25 3 30 33 33 5 4 35 39 39 3 40 44 48 6 4 46 51 55 (d) FACING RESISTANCE FOR HEADED STUD, TENSILE FAILURE, RFH, (PERMANENT FACING) hc(5)

f’c(3)


9-99


c. Using the recommended factor of safety for punching shear (FSFP), verify that that capacity for the temporary/permanent facing is higher than the nail head tensile force: RFP > FSFP To

(9-56)

d. If capacity for the temporary/permanent facing is not adequate, then implement larger elements or higher material strengths and repeat the punching shear resistance calculations. Facing Headed Stud Resistance (RFH) – Permanent Facing a. Calculate the maximum tensile resistance due to headed-stud tensile failure (RHT) using Table 9-12d, or alternatively as: R FH = N AS f y

(9-57)

b. Verify that that capacity is higher than nail head tensile force: RFH > FSHT To

(9-58)

c. Verify that compression on the concrete behind headed-stud is within tolerable limits by assuring that: AH ≥ 2.5 AS

(9-59)

tH ≥ 0.5 (DH – DS)

(9-60)

where AH is the cross-sectional area of the stud head; AS is the cross-sectional area of the stud shaft; tH is head thickness; DH is diameter of the stud head; and DS is diameter of the headed-stud shaft. d. Provide sufficient anchorage to headed-stud connectors and extended them at least to the middle of the facing section and preferably behind the mesh reinforcement in final facing. e. Provide a minimum 2 in. of cover over headed-studs. f. If capacity is not enough, adopt larger elements or higher strengths and recalculate.


9 – Soil Nail and Micropile Walls 9-100

June 2008

Other Facing Design Considerations To minimize the likelihood of a failure at the nail head connection, use the recommended minimum specifications for the hardware elements provided below. Additional information can be found in Chapter 5 of Geotechnical Engineering Circular No. 7 (Lazarte et al., 2003). • • •

Bearing Plates: Bearing plates should be mild steel with a minimum yield stress, fy, equal to 36 ksi (ASTM A-36/A36M). Nuts: Nuts should be the heavy-duty, hexagonal type, with corrosion protection (oversized when epoxy-coated bars are used). Beveled Washers: Beveled washers (if used) should be steel or galvanized steel. If the plate and other hardware elements are not within the ranges recommended, a formal calculation of capacities should be performed. Note that some proprietary systems employ spherical seat nuts that do not require washers.

9.2.7.7 Step 8: Estimate Maximum Wall Deformations 1) Use Figure 9-32 as a guide to estimate the magnitude of vertical and horizontal displacements. 2) Obtain wall height (H) and batter angle (α) (see Figure 9-32 for a description of variables). 3) Identify ground conditions (i.e., weathered rock/stiff soil, sandy soil, clayey soil). 4) Estimate horizontal and vertical displacements δh and δv at the top of the wall using Table 9-5 5) Calculate zone of influence, DDEF, where noticeable ground deformations occur using Equation D DEF = C (1 − tan α ) H

(9-61)

6) Verify estimated displacement against maximum deformation criteria.



June 2008

9.2.8

Soil Nail Load Testing

The design of a soil nailing system is usually tested in the field to verify that: (a) the design loads can be carried by the nails without excessive movements, (b) the contractor’s equipment and installation procedures are adequate, and (c) the long-term behavior of the nails is as anticipated. Four types of tests are usually performed: an ultimate test, a verification test, a proof test and a creep test. Table 9-13 presents a brief discussion on various aspects of these tests (Porterfield et al., 1994 and Clouterre, 1991).

Figure 9-43. Details of Typical Soil Nail Test Set-Up (Porterfield et al., 1994). Figure 9-43 illustrates a typical set-up for a soil nail test. A hydraulic jack and pump are used to apply the load to the nail. A jacking frame or reaction block is usually installed between the shotcrete (or the excavated face) and the jack. Once the jack is centered, an alignment load is applied to the jack to secure the equipment. Movement of the nail head is measured by one or preferably two dial gages attached to a rigid support independent of the jacking set-up. The dial gages are zeroed after the alignment FHWA NHI-07-071 Earth Retaining Structures


June 2008


9-103

June 2008

9 – Soil Nail and Micropile Walls

Table 9-13. Soil Nail Load Test Types

Figure 9-44. Typical Data Sheet for Soil Nail Load Testing (after Porterfield et al., 1994). FHWA NHI-07-071 Earth Retaining Structures


June 2008

Figure 9-45. Example of Data Reduction from Soil Nail Load Testing (after Porterfield et al., 1994) FHWA NHI-07-071 Earth Retaining Structures


June 2008

Figure 9-46. Example of Data Reduction from Soil Nail Creep Testing (after Poterfield et al., 1994).



June 2008

load has been applied. These gages should be capable of measuring movement to the nearest 0.8 mil. The load is applied to the nail by a calibrated hydraulic jack. The jack should have a minimum travel of 6 in. A calibrated load cell, which should be aligned with the axis of the nail and the jack, is used to detect small changes in load and allow maintenance of constant load during creep testing. Figure 9-44 shows a data log sheet that can be used for the load testing of soil nails. Figure 9-45 presents an example of data reduction of soil nail load testing to calculate elastic movement. Figure 9-46 presents an example of data reduction of soil nail load testing to calculate creep movement between 1- and 10-minute readings. 9.2.9

Construction Inspection of Soil Nail Walls

Inspection responsibilities for soil nail walls are summarized in Table 9-14. Table 9-14. Inspector Responsibilities for a Typical Soil Nail Wall. CONTRACTOR SET UP Review Plans and Specifications Review Contractor’s schedule Discuss anticipated ground conditions and potential problems with Contractor Review Contractor’s methods for surface water control and verify adequacy throughout construction Review corrosion protection requirements from the Specifications and confirm that Contractor is following these requirements If specified, obtain test samples from steel components, centralizers, and drainage materials and check all Mill test certificates for compliance with Specifications NAIL STORAGE AND HANDLING Nails, cement, and bars must be kept dry and stored in a protected location Nails and bars should be placed on supports to prevent contact with the ground EXCAVATION Prior to starting excavation, check for any variance between actual ground surface along the wall line and that shown on the Plans Collect excavated soil samples and perform visual identification. Inform Engineer of the results for comparison against the assumed soil type for design. Confirm that stability of excavated face (i.e., stand-up time) is maintained at all stages of construction Confirm that excavations are constructed within Specification tolerances of the design line and grade For each excavation lift, confirm that Contractor is not over excavating Enforce specific excavation sequencing plan provided on the Plans as they relate to lift thickness, length of open unsupported excavation, and, if required use of stabilizing FHWA NHI-07-071 Earth Retaining Structures


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berms Identify areas of excessive seepage and report to Engineer Confirm that excavated face profile is sufficiently smooth to facilitate shotcrete placement and to minimize overages in shotcrete quantities DRILLING OF NAIL HOLES Confirm that drilling technique used is consistent with ground conditions Document drilling procedures and report to the Engineer if drilling method unsuitable for actual ground conditions encountered Confirm that soil nail hole is drilled within acceptable tolerances of the specified location, length, and minimum diameter Observe and document locations of excessively hard drilling Visually inspect for loss of ground or drill hole interconnection and confirm that neither of them are occurring during drilling; subsidence of ground above drilling location or large quantities of soil removal with little or no advancement of the drill head should not be permitted TENDON INSTALLATION AND GROUTING Inspect open soil holes for caving or loose cuttings using a high intensity light Inspect all soil nail bars and reinforcing steel for damage and defects prior to installation Confirm consistency of epoxy coated or encapsulated tendons and inspect for any damage to corrosion protection prior to installation into drill hole Confirm mix design compliance of soil nail grout and take grout samples as required Record volume of grout placed for each drill hole Confirm that nail bars are inserted to the minimum specified length Confirm that centralizers are installed at specified intervals Confirm that all required hardware is appropriately affixed at the soil nail head Confirm that no damage occurs to corrosion protection components during installation Confirm that grout is injected by tremie pipe starting at the bottom of the hole and that the end of tremie pipe always remains below the level of the grout as it is extracted Confirm that grout is continued to be pumped as the grout tube, auger, or casing is removed Confirm that the Contractor does not reverse the auger rotation while grouting except as necessary to initially release the tendon Confirm that grout is batched in accordance with approved mix designs Observe Contractor’s methods to place grout/shotcrete just behind the soil nail head and confirm continuous coverage Confirm that any required testing for grout strength is conducted in accordance with specified testing methods LOAD TESTING Obtain all required calibration certifications of Contractor’s load testing equipment Check all deformation gauges and confirm movements during load testing Confirm that load testing of individual nails does not commence until minimum grout curing time has passed Confirm that the load test is performed consistently with Specifications and all required load test data is provided to permit comparison to acceptance criteria outlined in Specifications If the soil nail fails, report to the Engineer and do not allow any retesting until the FHWA NHI-07-071 Earth Retaining Structures


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Contractor modifies the installation procedures DRAINAGE INSTALLATION Confirm compliance of drainage materials with Specifications Confirm that geocomposite drain strips and weep hole outlet pipes are installed as specified and Plans and that the drain elements are sufficiently interconnected and provide continuous drainage paths WALL FACING Confirm shotcrete mix design consistent with Specifications Confirm that steel reinforcing is appropriately positioned within temporary shotcrete facing Confirm that exposed soil face is covered with shotcrete within specified time limits Confirm that minimum shotcrete thickness is maintained at all sections of the work Confirm that shotcrete installation methods used in the field are consistent with the Specifications and as approved by the Engineer Confirm that construction joints are clean and acceptable for shotcrete placement Confirm that shotcrete is batched in accordance with the approved mix design Confirm that wall finish line and grade is in accordance with Plans and Specifications If specified, confirm that shotcrete test panels are prepared, cured, and transported to the Testing Laboratory POST INSTALLATION Verify pay quantities

9.3

MICROPILE WALLS

9.3.1

Introduction

Micropiles are small diameter (less than 12 in.) drilled piles constructed with steel reinforcement, and bonded to the ground with grout using gravity or pressure grouting techniques. Micropiles may be used for structural support, slope stabilization, and retaining systems. Information on the design and construction of micropiles for structural support and slope stabilization is provided in Sabatini and Tanyu (2006). Micropile walls may be used for temporary shoring and permanent earth retaining systems. In general, micropiles are relatively expensive compared to other forms of deep foundation elements such as driven piles or drilled shafts. Inasmuch as drilled shafts and driven pile elements are used as vertical wall elements (e.g., secant pile walls and driven steel soldier piles), the use of micropiles for wall systems will likely only be a viable and cost-effective system where driven piles or drilled shafts cannot be installed. The principal components of a micropile wall consist of vertical micropile elements installed from the ground surface at or near the final excavated wall face line and subhorizontal FHWA NHI-07-071 Earth Retaining Structures


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elements installed from the ground surface which resembles a ground anchor. Figure 9-47 shows a cross section of a micropile retaining wall. The A-frame system formed by the vertical and subhorizontal micropiles is structurally connected with a reinforced concrete grade beam. The advantages of micropile walls include: • • • • • •

can be constructed in remote areas or where there is restricted access; can be installed in particularly difficult and variable ground conditions (e.g., karstic areas, uncontrolled fills, cobbles, boulders, etc.); unlike other drilled shaft systems, the construction of micropiles is less affected by soft clays, running sands, or high groundwater table; vibration and noise is minimal; no significant spoil generated during construction of the micropiles; and due to high tension and compression capacities of micropiles, a relatively tight Aframe configuration can be used allowing for construction under limited right-of-way constraints.

Some of the limitations of micropile walls include: • • •

•

underground easement is required for installation of micropiles; vertical micropiles have limited lateral capacity; micropiles may not be suitable in areas where liquefaction may occur due to concerns of buckling resulting from loss of lateral support (although this effect can be evaluated); and design methods are not well-developed primarily because of limited performance data for wall applications.

9.3.2 Construction Materials and Methods

Micropile retaining walls are constructed from the top-down and generally follow this sequence: • • •

at the ground surface, excavate an area wide and deep enough to accommodate the cap beam; install the formwork for the cap beam and place the cap beam steel reinforcement; place corrugated plastic sleeves for installation of the micropiles through the cap beam;



June 2008

24 ft

12 in.

4 in.

10 ft

15 ft

Figure 9-47. Micropile Wall Cross for Wall 600, Portland, Oregon.

• • • •

pour the concrete cap; install micropiles through the plastic sleeves; excavate, apply temporary shotcrete facing, and install geocomposite drain and other drainage systems until final excavation grade is reached; and install cast-in-place (CIP) wall facing (if required).

Excavation in front of the wall is performed in lifts (typically no more than 6 ft thick). During excavation, shotcrete is applied to the excavation face to temporarily prevent raveling of the soil face. Connection to the micropiles is performed via head studs that are welded to the front line micropiles. Following completion of the excavation, a leveling pad is poured to allow erection of one-sided forms. Once the leveling pad is completed the wall face is constructed from CIP concrete. Headed studs welded to the micropiles are embedded in the CIP grade beam and wall face to provide connection of the micropile structure to the CIP wall face. FHWA NHI-07-071 Earth Retaining Structures


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The typical construction sequence for simple gravity grouted and pressure-grouted micropiles (Figure 9-48) includes drilling the pile shaft to the required tip elevation, placing the steel reinforcement, placing the initial grout by tremie, and placing additional grout under pressure as applicable. In general, the drilling and grouting equipment and techniques used for the micropile construction are similar to those used for the installation of soil nails and ground anchors. The amount of steel reinforcement placed in a micropile is determined by the loading it supports and the stiffness required to limit elastic displacements. Reinforcement may consist of a single reinforcing bar, a group of reinforcing bars, a steel pipe casing or rolled structural steel.

Figure 9-48. Typical Micropile Construction Sequence Using Casing.



June 2008

Reinforcement may be placed either prior to grouting, or placed into the grout-filled borehole before the temporary casing (if used) is withdrawn. It must be clean of deleterious substances such as surface soil and mud that may contaminate the grout or coat the reinforcement, impairing bond development. Suitable centralizers should be firmly fixed to maintain the specified grout cover. Pile cages and reinforcement groups, if used, must be sufficiently robust to withstand the installation and grouting process and the rotation and withdrawal of the temporary casing. 9.3.3

Micropile Wall Design

9.3.3.1 Overview No generally accepted procedure is available to design micropile retaining walls; however, these systems may be analyzed using soil-structure interaction analyses in which the axial stiffness and bending stiffness of the vertical and battered micropiles are explicitly modeled. Also, all stages of excavation in front of the wall can be modeled. With this approach, other potential failure mechanisms need to be considered separately including the potential for soil to squeeze in-between the small-diameter micropiles and the potential for structural failure of the vertical micropiles due to buckling. Buckling is checked because the relatively smalldiameter vertical micropiles will experience compressive loads and they are close to the exposed ground surface. For the Wall 600 project in Portland (i.e., micropile wall shown in Figure 9-47), detailed soilstructure interaction analyses were performed to verify a more simplified model in which a design total pressure diagram (including earth pressures, seepage pressures, seismic forces, and traffic barrier impact loading) was applied proportionally to the two vertical micropiles and the two battered micropiles for each micropile section (see Ueblacker, 1997). 9.3.3.2 Design Approach for Temporary Shoring Systems Micropiles were used to provide temporary excavation support for a project involving improvements to State Highway 82 near Aspen, Colorado (Macklin et al., 2004). In this mountainous region, construction is hampered by difficult site access and slope instability risks. This site consists of loose to very dense silty to gravelly sand with cobbles and boulders, which were deposited as debris flow, sheet wash, and colluvium over dense alluvial sandy gravel with cobbles. Originally, temporary shoring using combinations of soil nails and tiebacks was considered for the project. However, in an effort to improve the construction schedule and phasing, micropiles were selected as an alternative temporary shoring system. The micropile shoring implemented on this project was essentially a hybrid between a soldier pile and lagging system and a soil nail stabilization system (Figure 9-49). Phasing of the project required constructing a number of bridges in a specific sequence. The FHWA NHI-07-071 Earth Retaining Structures


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micropile shoring system allowed the contractor to transition access road grades where convenient, depending upon the excavation and access requirements for any phase of the work. For this project, the internal and external stability of the temporary micropile shoring system was analyzed using a combination of gravity wall calculations, free earth support methods, lateral pile, and slope stability calculations. Also, finite element and finite difference models were used to predict deformation behavior, stresses in the micropiles, and to analyze the potential for soil flow around the micropiles. The design approach for this project is described in the following six steps.

Note: 1 meter = 3.28 ft Figure 9-49. Cross Section Showing Steep Canyon Slope and Temporary Micropile Shoring (after Macklin et al., 2004). •

Micropile sections were assumed and the ultimate bending capacity of the micropile sections was calculated. The flexural rigidity (EI) of the micropiles was calculated and these values were later used in numerical analyses. Micropiles that consisted of a centralized reinforcing bar in a drilled and grouted hole were analyzed using LPILE’s ultimate bending analysis module. The tensile and compressive capacity of each section was also calculated.



June 2008

•

The micropile wall system was next analyzed as a rigid gravity wall. The wall geometry was defined as the ground enclosed by the micropile system envelope. Earth pressures were calculated using classical earth pressure theories, assuming that the wall deformed sufficiently to allow the soil to reach the active state. Sliding of the system was analyzed and included the shear capacity of the front micropile. The embedment of the micropile was checked to evaluate whether sufficient passive resistance could be developed in front of the micropile to mobilize the required micropile shear strength. Overturning was checked by summing overturning moments about toe of micropile wall. The required bond length of the rear row of battered micropiles to resist the overturning moment with respect to tensile rupture and pullout failure was computed.

•

The micropile wall system was next analyzed using the free earth support method (as is commonly used for anchored bulkhead design). For this evaluation, the front row of closely spaced micropiles is considered to be analogous to a sheet pile wall and the battered rows of micropiles are analogous to the deadman anchors. The analysis was modified in that it was assumed that one-half of the calculated active earth loads was applied to the vertical micropile row as a triangular pressure distribution and one-half of the calculated active earth load was applied to the rear, battered micropile row. A lateral pile analysis was then completed for both the vertical and battered micropiles using LPILE.

•

Global stability of the micropile system was evaluated.

•

Finite element analyses were performed to predict deformations in the structure. Worst case geometries and construction stages of the temporary excavation support for the walls were analyzed.

•

The potential for soil flow in-between the relatively closely-spaced micropiles was evaluated.

9.3.4

Load Testing

Load testing is performed on micropiles in the field to verify that: (a) the design loads can be carried by the micropiles without excessive movements, (b) the contractor’s equipment and installation procedures are adequate, and (c) the long-term behavior of the micropiles is as anticipated. Micropiles are tested individually using the same conventional static load testing procedures as are used for driven piles and drilled shafts. These tests include incremental loading (which may be applied in compression, tension, or laterally and which may be cycled (i.e., FHWA NHI-07-071 Earth Retaining Structures


June 2008

load/unload)) until the micropile reaches the selected maximum test load, structural displacement limit, or ground creep (i.e., movement under constant load) threshold. Unlike ground anchors and soil nails for which well-defined testing programs, consistent with a well-developed design approach, are available, load testing program protocols for micropiles used for earth retaining systems have not been developed. Regardless, however, load testing does need to be performed to verify the displacement response and capacity of micropiles used for wall systems. Such a testing program will need to be developed on a project-specific basis. Details of micropile load testing can be found in Sabatini and Tanyu (2006). It is also recommended that performance data (including micropile load transfer, axial loads, bending moments, and displacements) be collected for micropile wall systems to enable design methods to be updated. 9.3.5 Construction Inspection

The responsibilities of the inspector for micropile wall construction are summarized in Table 9-15. Table 9-15. Inspector Responsibilities for a Typical Micropile Wall Project. CONTRACTOR SET UP Review Plans and Specifications Review Contractor’s schedule Review in-situ property test results (i.e., grain size, Atterberg limits, unit weight, and shear strength) Discuss anticipated ground conditions and potential problems with Contractor Confirm that Contractor’s grout pump is consistent with Plans (i.e., positive displacement pump) and grout equipment is capable of producing uniform grout Confirm that the dimensions of micropiles consistent with Specifications Confirm that cement, reinforcement steel, and micropile are handled and stored consistently with Specifications Review corrosion protection requirements of metallic units and confirm their consistency with Specifications CONCRETE CAP INSTALLATION Confirm that site preparation for the wall construction is consistent with Specifications Confirm that the area excavated for the cap beam is consistent with Plans Confirm that formwork for the cap beam is installed according to Specifications and Plans Confirm that corrugated plastic sleeves are installed through the cap beam according to the Plans Confirm that the concrete cap is poured according to Specifications and Plans FHWA NHI-07-071 Earth Retaining Structures


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MICROPILE INSTALLATION Confirm that layout of micropiles consistent with Plans Visually inspect the micropiles to asses any possible damage prior installation Confirm that micropiles are installed through the plastic sleeves to the minimum grouted lengths as shown on Plans Confirm that micropiles are installed with the sequence specified in Specifications Confirm that the minimum hole diameter is consistent with Plans Confirm that the steel bar is consistent with Specifications and is inserted into the hole to the depths shown on Plans Confirm that the grout is injected at the lowest point of each micropile and the hole is filled in a continuous operation Confirm that disposal of excavated material is performed according to Specifications Confirm that tops of all micropiles are surveyed and the results are provided to the Engineer LOAD TESTING Inspect calibration certifications for all jacks, gauges, and load cells according to the requirements specified in Specifications Inspect all deformation gauges to confirm movements Confirm that the load test is performed consistently with Specifications and test results are consistent with load test acceptance criteria as outlined in Specifications. If the micropile fails, report to the Engineer and do not allow any retesting until the Contractor modifies the installation procedures EXCAVATION FOR WALL FACE Confirm that excavation in front of the vertical micropiles is performed in lifts according to the Specifications Confirm that the excavated face is shotcreted according to the Specifications and no further excavation is performed until shotcreting is completed Report significant excavation face sloughing to Engineer immediately DRAINAGE INSTALLATION Confirm compliance of drainage materials with Specifications Confirm that geocomposite drain strips are installed as specified and Plans and that the drain elements are sufficiently interconnected and provide continuous drainage paths WALL FACE Confirm that the construction of leveling pad is consistent with Plans Confirm that CIP concrete wall face is constructed according to Specifications Confirm the alignment of the wall face by visual inspection and by using leveler POST INSTALLATION Verify pay quantities



June 2008



June 2008

CHAPTER 10 WALL SELECTION 10.1

INTRODUCTION

Prior to 1970, the predominant types of earth retaining walls for permanent structures were gravity and cantilever. Both gravity and cantilever wall types had decades of successful use in both cut and fill situations. Therefore, selection of a wall type consisted of choosing between gravity and cantilever walls. Selection of a wall system in 2007 is considerably more complex because the number of wall types available has increased significantly. A systematic evaluation process should be used to select the most appropriate wall type for the project. This chapter presents a systematic wall system evaluation and selection process. The objective of wall selection is to determine the most appropriate wall type that is costeffective, practical to construct, stable, and aesthetically and environmentally consistent with its surroundings. As part of the evaluation process, walls are scored using a wall selection matrix based on wall selection factors. The wall with the highest score is chosen as the wall for the project. A flowchart illustrating wall selection is presented in Figure 10-1. This flowchart is intended to serve as a guide for highway design and construction specialists for evaluating and selecting wall system alternatives for a project application. Aspects of wall selection that are outside the scope of this flowchart, but which may be part of a formal review and acceptance program, include review of approved wall system lists by an owner agency, establishment of technical guidelines and criteria by which feasible wall system alternates can be judged, and performance of a life cycle cost analysis for candidate wall systems. These aspects are generally agency-specific and are not discussed in this chapter.

10.2

STEP 1: IDENTIFY NEED FOR AN EARTH RETAINING SYSTEM

The first step in the selection process is to identify the need for an earth retaining system for the project. The function of a retaining system is to form a nearly vertical face through confinement and/or strengthening of a mass of earth material. Typically, an earth retaining system is needed for projects that require abrupt changes in slope grades that cannot be achieved by simply grading the slopes. As a rule of thumb, it can be stated that the more restricted or congested the site, the greater the need for an earth retaining system. This is FHWA NHI-07-071 Earth Retaining Structures

10-1

10 – Wall Selection June 2008

Step 1:

Identify Need for an Earth Retaining System

Step 2:

Identify Site Constraints and Project Requirements

Step 3:

Evaluate Project Requirements Against Factors Affecting Wall Selection

Step 4:

Evaluate Wall Alternatives Against Wall Selection Factors

Step 5:

Select An Acceptable Wall Type

Figure 10-1. Wall Selection Flow Chart. because congested sites typically have very limited right-of-way (ROW) which does not allow abrupt changes in slope grades by just grading the slopes. The ROW restrictions of the site can be evaluated by visiting the site for preliminary review. During the site visit, the site geometry can be evaluated and ROW restrictions, such as existing structures and utility lines, can be noted. Typically, the observations from the site visit when compared against the proposed project requirements are enough to make an initial decision on the need for an earth retaining system for the project.

10.3

STEP 2: IDENTIFY SITE CONSTRAINTS AND PROJECT REQUIREMENTS

The second step in wall selection involves identifying site specific constraints and project requirements. This information can be obtained during a preliminary site review. Items affecting wall selection include, but are not limited to the following: (1) site accessibility and space restrictions that may include limited ROW and headroom, availability of on-site storage for wall materials, access for specialized construction equipment, and restrictions on traffic disruption; (2) location of above-ground utilities and nearby structures; (3) aesthetic requirements imposed by project surroundings; (4) environmental concerns that may include local policies concerning construction noise, vibration, and dust, on-site stockpiling and/or transport and disposal of excavated material, discharge of large volumes of water, and encroachment on existing waterways; and (5) exposed wall face height. The relative importance of each of the above items should be assessed for the specific project under consideration so that the more important items are given priority during the selection process. FHWA NHI-07-071 Earth Retaining Structures

10-2


Once all the necessary site constraints and project requirements are identified then the next step (i.e., Step 3) should be to evaluate these project requirements against wall selection factors.

10.4

STEP 3: EVALUATE PROJECT REQUIREMENTS AGAINST FACTORS AFFECTING WALL SELECTION

For this step, eleven wall selection factors are considered. These factors include: (1) ground type; (2) groundwater; (3) construction considerations (i.e., availability of material, equipment, and etc.); (4) speed of construction; (5) ROW; (6) aesthetics; (7) environmental concerns; (8) durability and maintenance; (9) tradition; (10) contracting practices; and (11) cost. This list is not project specific and is presented as a tool to help evaluate the importance of project requirements determined in step 2 in a systematic manner. For a given project, more specific factors such as, for example, lateral movements, cost of maintenance, and the availability of a standard design may be critical in the process. The evaluation is performed by the party that is responsible in wall selection. Depending on the contracting policy adopted for the project, this evaluation may be performed by the owner, the consulting engineer representing the owner, or the contractor. Each wall selection factor is evaluated based on its relevancy and importance to the project requirements and site constraints and they are assigned a rating number between one (1) and three (3). This is termed the weighted rating (WR) for a given wall selection factor. Three is assigned to the most relevant or important factors and one is assigned to the least relevant ones. The evaluation results should be tabulated as shown in Table 10-1. Typically, cost, speed of construction, and durability are the most important wall selection factors for permanent wall systems. In general, any issues which are given a WR = 1 need not be considered further.


10-3


Groundwater

Construction Considerations

Speed of Construction ROW

Aesthetics

Environmental Concerns

Durability and Maintenance

Tradition

Contracting Practice

Cost


10-4


Note: 1Rating of the importance of each wall selection factor based on project requirements and site constraints. Each factor should be rated between 1 and 3, where 1 is the least important or relevant factor and 3 is the most important factor.

WR1

Ground

Factors That Affect Wall Selection

Table 10-1. Wall Selection Factors.

10.5

STEP 4: EVALUATE WALL ALTERNATIVES AGAINST WALL SELECTION FACTORS

10.5.1 General From this step forward, various wall systems are evaluated for the project. As a logical first substep, obviously inappropriate wall systems should be eliminated. For this elimination process, project constraints related to wall geometry and wall performance (or design requirements) should be considered. Constraints related to cost should not be addressed at this stage; cost will be addressed consistently for all wall systems that are not eliminated as part of this step. The factors affecting the selection of cut and fill wall systems are summarized below. In some projects both cut and fill wall systems may be suitable and in these types of projects both wall systems should be evaluated. 10.5.2 Selection Issues for Cut Wall Systems Cut walls are constructed from the top down and most cut walls can be used for both temporary and permanent applications. The permanent systems are typically designed with greater corrosion protection measures and are constructed with permanent facing elements such as cast-in-place or precast concrete panels. Cut walls are either drilled or cut into the in-situ soil or rock and are constructed with specialized equipment and labor. For cut walls without ground anchors (i.e., nongravity cantilever walls) little or no ROW is required. For anchored walls and soil-nailed walls, significant ROW or permanent easements may be necessary. Costs can vary significantly for cut walls depending on the specific wall being constructed (i.e., complexity of the construction process) and the availability of experienced contractors and equipment in the project location. For cut walls, the unit cost of the wall increases as the height of the wall increases. For wall heights greater than approximately 15 to 30 ft, an anchored wall or a soil-nailed wall is necessary. Additional cost results from material procurement, drilling, installation, corrosion protection, and testing of the anchors or soil nails. For all permanent wall systems, factors affecting costs include constructing an aesthetically pleasing wall finish, fabricating and installing special connections for the facing panels, and, if necessary, providing adequate long-term corrosion protection and constructing drainage systems.


10-5


The advantages and disadvantages, cost-effective height range, required ROW, speed of construction, and representative tolerances for lateral movements and water tightness for cut wall systems are summarized in Table 10-2. 10.5.3 Selection Issues for Fill Wall Systems Fill walls are constructed from the bottom up and are typically used for permanent wall applications due to the high cost of their facing components, although temporary MSE systems are used without permanent facings. All fill walls require ROW and their costeffective height range is less than the cost-effective height range of cut walls. For permanent highway applications, fill wall systems generally require granular, nonplastic, free-draining backfill. The cost effectiveness of a MSE wall system, which typically requires a greater quantity of select backfill than rigid gravity and semi-gravity wall systems, may be reduced if select backfill is unusually expensive at a specific project site. The advantages and disadvantages, cost-effective height range, required ROW and representative tolerances for differential settlement for fill wall systems are summarized in Table 10-3. 10.5.4 Wall System Alternatives Evaluation The performance of wall alternatives that are not eliminated earlier in this step should be evaluated against wall selection factors that are defined here using a rating between one (1) and four (4). For each factor, four is assigned to the most suitable factor for the wall being evaluated and one is assigned to the least suitable factor for the wall being evaluated. At this stage, constraints related to cost should be addressed for all wall systems that are not eliminated previously. The rating for each wall type should be tabulated for use in Step 5. A brief summary of how each of the wall selection factors affects the performance of each wall type is given below. 10.5.4.1

Ground Type

An earth-retaining system is influenced by the earth it is designed to retain, and the one on which it rests. The influence of the earth is particularly important in “earth walls” where the retained earth itself has a major load-carrying function. In MSE walls, for instance, which usually involve some sort of reinforcement, the pull-out force in the reinforcement is resisted by (1) the friction along the soil-reinforcement interface and (2) the passive resistance along the transverse members of the reinforcement, if any (grid reinforcement). Therefore, these systems are best suited for soils with high internal friction such as sands and gravels. FHWA NHI-07-071 Earth Retaining Structures

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9

9

9

9

9

9

9

9

9

9

9

Tangent pile wall

Secant pile wall

DMM wall

Anchored wall

Soil-nailed wall

Micropile wall

None(5)

0.6 H + anchor bond length 0.6 - 1.0H

15 – 70 ft(3)

N/A

Varies

None(5)

20 – 80 ft(2)

10 – 70 ft

None(5)

10 – 30 ft 20 – 80 ft(2)

20 – 80 ft(2)

N/A

smallmedium

smallmedium

small

small

small

None(5)

20 – 80 ft(2)

10 – 30 ft

small

None

N/A

N/A

N/A

fair

Fair

poor

good

⋅

⋅ ⋅

⋅ ⋅

⋅

⋅ ⋅

⋅ ⋅ ⋅ ⋅

⋅

Lateral Water Movements Tightness large fair ⋅ ⋅ medium poor ⋅ ⋅

up to 16 ft

Required ROW(4) None Advantages


5

ROW required if wall includes anchors.

2

Disadvantages

10-7


rapid construction ⋅ difficult to construct in hard ground or through obstructions readily available rapid construction ⋅ difficult to maintain vertical tolerances in hard ground soldier beams can be drilled or driven ⋅ potential for ground loss at excavated face can be constructed in all soil types or ⋅ requires specialty contractor weathered rock ⋅ significant spoil for disposal watertight ⋅ requires specialized equipment wide range of wall stiffness adaptable to irregular layout ⋅ difficult to maintain vertical tolerances in hard ground can control wall stiffness ⋅ requires specialized equipment ⋅ significant spoil for disposal adaptable to irregular layout ⋅ difficult to maintain vertical tolerances in hard ground can control wall stiffness ⋅ requires specialized equipment ⋅ significant spoil for disposal adaptable to irregular layout ⋅ requires specialized equipment ⋅ relatively small bending capacity can resist large horizontal pressures ⋅ requires skilled labor and specialized equipment adaptable to varying site conditions ⋅ anchors may require permanent easements rapid construction ⋅ nails may require permanent easements adaptable to irregular wall alignment ⋅ difficult to construct and design below water table does not require excavation ⋅ requires specialty contractor

Perm. and Temp. indicate permanent and temporary wall applications respectively. Height range given is for wall with anchors. 3 For soldier pile and lagging wall only. 4 ROW requirements expressed as the distance (as a fraction of wall height, H) behind the wall face where wall anchorage components (i.e., ground anchors and soil nails) are installed.

1

9

9

Slurry (diaphragm) wall

Notes:

9

9

Perm.1 Temp.1 Cost Effective Height Range 9 9 up to 16 ft

Soldier pile/ lagging wall

Wall Type Sheet-pile wall

Table 10-2. Summary of Evaluation Factors for Cut Walls (after Sabatini et al., 1997).

6 – 35 ft

9

9

9

9

9

9

9

9

9

9

9

9

9

9

9

9

9

9

Concrete crib wall

Metal bin wall

Concrete module wall

Gabion wall

MSE wall (precast facing)

MSE wall (modular block facing)

MSE wall (geotextile/geogrid/ welded wire facing) MSE wall (vegetated soil face)

6 – 40 ft

10 – 100 ft

6 – 50 ft

6 – 60 ft

10 – 100 ft

6 – 30 ft

6 – 25 ft

1/500

1/500

0.4 - 0.7H(4)

0.4 - 0.7H(4)

0.5 – 0.7H

0.5 – 1.0H

0.7 - 1.0H

0.7 - 1.0H

0.7 - 1.0H

0.5 - 0.7H

0.5 – 0.7H

0.5 - 0.7H

1/100

1/60

1/60

1/200

1/100

1/50

1/300

1/300

1/300

1/500

0.5 - 0.7H(4)

0.5 - 0.7H

Differential Settlement Tolerance(3)

Required ROW(2)

⋅ ⋅ ⋅ ⋅

⋅ ⋅ ⋅

⋅ ⋅

⋅ ⋅ ⋅

⋅ ⋅

⋅ ⋅ ⋅ ⋅ ⋅ ⋅ ⋅ ⋅

⋅ ⋅ ⋅

⋅ ⋅ ⋅

⋅ ⋅

Disadvantages

10-8


durable ⋅ deep foundation support may be necessary requires smaller quantity of select backfill as compared to MSE ⋅ relatively long construction time walls concrete can meet aesthetic requirements durable ⋅ deep foundation support may be necessary requires smaller quantity of select backfill as compared to MSE ⋅ relatively long construction time walls concrete can meet aesthetic requirements durable ⋅ deep foundation support may be necessary requires smaller quantity of select backfill as compared to MSE ⋅ relatively long construction time walls concrete can meet aesthetic requirements does not require skilled labor or specialized equipment ⋅ difficult to make height adjustments in field rapid construction does not require skilled labor or specialized equipment ⋅ difficult to make height adjustments in field rapid construction ⋅ subject to corrosion in aggressive environment does not require skilled labor or specialized equipment ⋅ difficult to make height adjustments in field block are easily handled does not require skilled labor or specialized equipment ⋅ need adequate source of stone ⋅ construction of wall requires significant labor does not require skilled labor or specialized equipment ⋅ requires use of select backfill flexibility in choice of facing ⋅ subject to corrosion in aggressive environment (metallic reinforcement) does not require skilled labor or specialized equipment ⋅ requires use of select backfill flexibility in choice of facing ⋅ subject to corrosion in aggressive environment (metallic reinforcement) blocks are easily handled ⋅ positive reinforcement connection to blocks is difficult to achieve does not require skilled labor or specialized equipment ⋅ facing may not be aesthetically pleasing flexibility in choice of facing ⋅ geosynthetic reinforcement is subject to degradation in some environments ⋅ facing may not be aesthetically pleasing does not require skilled labor or specialized equipment ⋅ geosynthetic reinforcement is subject to degradation in flexibility in choice of facing vegetation provides ultraviolet light protection to geosynthetic some environments ⋅ vegetated soil face requires significant maintenance reinforcement ⋅ requires high quality backfill does not require skilled labor or specialized equipment ⋅ fairly new technology and limited case histories can be built on a batter ⋅ difficult to make height adjustments in the field blocks are easily handled requires smaller quantity of select backfill as compared to MSE walls

Advantages

Perm. and Temp. indicate permanent and temporary wall applications respectively. ROW requirements expressed as the distance (as a fraction of wall height, H) behind the wall face where fill placement is generally required for flat backfill conditions, except where noted. Ratio of the difference in vertical settlement between two points along the wall to the horizontal distance between the points. 4 ROW requirement given is the typical wall base width as a fraction of wall height, H.

3

2


Notes:

1

30 – 60 ft

9

Concrete counterforted wall

T-wall

6 – 30 ft

9

Concrete cantilever wall

6 – 35 ft

3 – 10 ft

Cost Effective Height Range

9

Perm.1 Temp.1

Concrete gravity wall

Wall Type

Table 10-3. Summary of Evaluation Factors for Fill Walls (after Sabatini et al., 1997).

When in situ reinforcement is used to support excavations, such as in soil nail walls, the possible saturation and creep of the in situ soil can have a large negative impact on the longterm performance of the system. Therefore, these systems are less suited for highly plastic and creep-prone clayey soils than for granular materials. For the same reason, the anchored wall also is not frequently used in these soils. Gravity-type structures are less influenced by the type of soil than the systems involving soil reinforcement. For soils with large vertical and horizontal deformations, a very flexible system such as a gabion wall may be chosen in lieu of a more rigid system that attempts to resist such deformations. In DMM and jet grouted walls, the compatibility of the ground and the structure depends largely on the type of chemicals used, and hence on the specific system employed. While DMM walls, for instance, are normally used in sandy and liquefiable soils, the lime-column walls are mostly suitable for use in deep clay deposits rich in pozzolans. In jet-grouted walls, the in situ properties and structure of the soil are not as important since the concept of jet grouting is to break down the soil structure and replace it with a self supporting composite mass of soil and grout sometimes called soilcrete. The strength of the soilcrete, however; and the permeability of the wall may be influenced by certain soil elements such as peat or boulders. 10.5.4.2

Groundwater

Generally, the groundwater table behind an earth-retaining structure is lowered for the following reasons: • • •

to reduce the hydrostatic pressure acting on the structure; to reduce the likelihood of corrosion of any metal reinforcing and facing elements used in the system; and to prevent saturation of the soil, which may significantly increase displacements and cause instability during or after excavation, particularly in systems involving staged excavation and support, such as soil nailing.

To reduce the negative impact of groundwater, a free-draining system such as a MSE wall can be used. Sometimes, it is desirable to keep the water table high to prevent settlement of adjacent structures or protect existing untreated timber pile foundations from fungus decay due to exposure to oxygen. In these cases, a relatively rigid watertight structure is used (slurry wall, tangent/secant piles, jet-grouted wall, etc.). These structures usually are designed to support the full hydrostatic pressure.


10-9


10.5.4.3


Availability of construction material, site accessibility, equipment availability, temporary dewatering requirements, and labor considerations are important factors affecting the selection of an earth retaining system. Material availability is an important factor affecting the selection of a retaining wall type. For example, where rock is abundant, a gabion wall may be suitable and economical. This may not be the case, however, where suitable rock backfill has to be hauled a long distance to the project site. Where suitable aggregate is not available, structures built of concrete usually are avoided. Site accessibility is also an important factor. Depending on the terrain, the mobilization of heavy equipment might not be possible. This would limit some of the available wall systems that could be built. For example, crib walls or bin walls can be preassembled and mobilized to the sites that are difficult to access with heavy equipment. However, a sheet pile or soldier pile and lagging wall require the use of a hammer and/or drilling equipment and could not be built unless such equipment could be mobilized to the site. A major construction-related advantage of earth walls, as compared to conventional cast-inplace concrete structures, is that no form work is required on the job, and construction can be accomplished regardless of temperature. However, the behavior of some walls, such as DMM walls, soil-nail walls, and bin walls may be influenced by the soil temperature during and after construction. The temperature at the site, for instance, may affect the hydration of the cement in jet grouting. Also, repeated freeze-thaw cycles may soften up frost-susceptible in situ or backfill soils and cause ground deformations that may have negative impacts on the structure. 10.5.4.4

Speed of Construction

Speed of construction is one of the most important factors in wall selection. Precast concrete module walls and precast concrete crib walls can be constructed relatively quickly. These walls are shipped to the site assembled. Each wall face panel covers a relatively large area, which makes the construction even faster. The cells of the bin walls are manufactured and delivered to the site ready for assembly. Each steel member is bolted together at the site and then filled with backfill soils. Speed of construction for all gravity wall systems is affected by the condition of the wall foundation materials. If the foundation soils consist of unsuitable material, such material will need to be supported on a deep foundation, removed, or improved (e.g., surcharge fill) before the construction of the walls, which will increase construction time significantly.


10-10


Sheet pile walls, soldier pile and lagging walls, and soil nail walls are constructed relatively rapidly. Sheet pile walls can be constructed even more rapidly if a vibratory hammer can be used instead of an impact hammer. However, installing ground anchors with sheet pile walls and soldier pile and lagging walls can slow down the construction dramatically. As compared to anchored walls, construction of soil nail walls is rapid because the installation of structural wall elements is not required. Walls that require a relatively long time to construct include concrete gravity walls, concrete cantilever walls, and concrete counterfort walls. Preparation of foundation and pouring concrete are the two important factors that affect the speed of construction for these types of walls. These walls have very little tolerance to differential settlement, therefore the foundation has to be prepared to provide a relatively stiff and uniform bearing surface. Once the foundation is prepared, concrete should to poured carefully to eliminate cold joints and no further construction should be performed until the concrete is allowed to cure and reach a specified strength. Concrete cantilever and counterfort walls also require reinforcing elements. Gabion walls require a relatively long time to construct because they require significant labor. Each gabion basket module has to be assembled in the field and tied to one another to form the face of the wall. After that, each gabion basket has to be filled with stone and the lid of the gabion basket module has to be closed and tied. As with any other modular gravity wall, unsuitable foundation material has to be removed or treated before the construction of the gabion baskets. 10.5.4.5

Right-of-Way and Space Requirements

In MSE walls, a relatively large space is required behind the structure face as compared to that needed for construction of conventional walls (the length of the reinforcing elements is typically 0.7 times the wall height). To support an excavation in a very tight space, a top-down staged excavation and support system, such as soil nailing or an anchored wall, may be the most suitable. The feasibility of such a structure, however, is influenced by the presence of utilities and buried structures nearby and the additional cost of permanent underground easement for placement of the reinforcing elements. Soil nail walls may also allow for construction underneath a bridge without the need to disrupt traffic on the bridge. Site congestion may be a drawback for some systems such as slurry walls. When low headroom does not allow the operation of conventional construction equipment, walls which can be implemented in a limited operating space (i.e., soil nails, micropiles) or from a remote FHWA NHI-07-071 Earth Retaining Structures

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operating area (i.e., jet grouting) are preferred. If the earth retaining system is to be located adjacent to the ROW line, the space needed in front of the wall for construction should be considered as part of the ROW restrictions. 10.5.4.6

Aesthetics

In addition to being functional and economical, permanent earth retaining structures, in most cases, have to be aesthetically pleasing. Different types, shapes and color facings are used in construction of earth walls. The types of facings range from built on-site continuous facings (shotcrete, welded wire mesh, cast-in-place concrete) to prefabricated concrete or steel panels. Cast-in-place facing and precast panels usually are more attractive than shotcrete or soldier pile and lagging walls. For permanent drilled shaft walls (i.e., tangent/secant pile walls), an architectural wall facing usually is provided, but at an additional cost to the structure. The aesthetic factor is very important when building a retaining system in parks, forests and natural habitat. A number of attractive wall systems (Criblock, Evergreen, etc.) are usually considered for those areas because of their aesthetic, acoustic and anti-graffiti advantages. The Evergreen wall, for instance, consists of precast concrete units with open spaces at the face into which are planted shrubs, vines, etc. With adequate water supply for the foliage, the concrete facing will no longer be visible a few years after construction. 10.5.4.7


Like most structures, the selection of an earth retaining system is influenced by its potential environmental impact during and after construction. Excavation and disposal of contaminated material at the project site, and discharge of large quantities of water or slurry fluids generated during jet grouting and slurry wall constructions are of primary concern. MSE walls which allow construction of roadway embankments with vertical sides to minimize encroachment on wetlands have positive environmental and ecological benefits. The environmental advantages of these walls can be further enhanced by the use of wood chips, and other recycled waste materials in construction. To reduce noise and vibration impacts, the systems which use pile driving or heavy construction machinery may be rejected. To reduce traffic noise in environmentally-sensitive areas, the gravity-type gabion and Evergreen walls offer specific advantages. The open nature of the face and the presence of foliage covering are effective in absorbing the noise hitting their facings, making these walls acoustically superior as compared to other earthretaining structures where the traffic noise is reflected on hard or smooth continuous surfaces. FHWA NHI-07-071 Earth Retaining Structures

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10.5.4.8


An earth retaining structure built of concrete has a higher durability against corrosion and deterioration effects than a structure constructed of metal, or which uses metal or synthetics for reinforcement and/or facing. The durability factor is extremely important when selecting a maintenance-free earth retaining structure in highly corrosive surroundings, or when the structure is subjected to attack by non-conventional elements such as waves, chemicals or marine borers. Corrosion of the reinforcement, for instance, is one of the major design issues of MSE walls that use metal reinforcement. Although, these walls are sometimes avoided in higher than normal corrosive environments, corrosion-protection measures can be provided if necessary, but at an added cost to the structure. Gabion walls have durability concerns similar to those of MSE walls with metal reinforcement. When geosynthetics are used for reinforcement, their long-term creep behavior and resistance to deterioration due to chemical attack and exposure to ultraviolet light are major considerations that have to be addressed in the design. Selection of the types of geosynthetic is sometimes dictated by durability. The durability of a concrete structure (gravity wall, slurry diaphragm wall, etc.) is influenced by the quality of the aggregates and water used in the mix, and by the casting procedures. As indicated before, concrete walls are not recommended in areas where quality aggregates are not economically available. 10.5.4.9

Tradition

Tradition may dictate or prevent the use of a certain-type of structure, irrespective of its technical rating. Although earth walls are very popular in certain states they are rarely built in others. While still considered novelty in certain parts of the U.S.A., slurry walls are heavily used in construction of underground facilities in other areas such as Boston where drilled caissons are traditionally used as deep foundations, bored-pile walls are popular and generally economical since the local contractors are equipped for, and experienced with, that type of construction. Tradition plays a greater role in construction in underdeveloped countries. 10.5.4.10


The contracting policies and procedures followed in the United States may discourage or even preclude the use of certain types of walls, particularly those involving patented equipment, materials or procedures. Contracting issues are discussed in detail in Chapter 11. FHWA NHI-07-071 Earth Retaining Structures

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10.5.4.11

Cost

The total cost of an earth-retaining system has many components including the structure, ROW, temporary or permanent easement, excavation and disposal of unsuitable material, and drainage. The construction costs of specific types of retaining systems have been discussed in previous chapters and are summarized here in Tables 10-4 and 10-5 for cut and fill wall systems respectively. It should be noted, however, that a structure with the least construction cost does not necessarily mean an economical alternative, as the ultimate cost of the system is influenced by many indirect cost factors such as those listed above as well as schedule, permitting, maintenance, and wall face requirements. For cut walls, the most common permanent wall facing is either cast-in-place (CIP) or precast concrete. As of 2004, the cost of precast concrete facing is $25 to $30 per ft2 of facing. The cost of CIP concrete facing is typically less than the cost of precast concrete facing and as the area increases the cost decreases. At about 10,000 to 15,000 ft2, the cost of CIP concrete wall facing may drop to $15 to $25 per ft2 of facing.

10.6

STEP 5: SELECT AN ACCEPTABLE WALL TYPE

This is the final step where walls are compared to each other in a wall selection matrix format and the wall (or walls) that have the highest score is selected for the project. The scoring of each wall type is obtained for each wall selection factor by multiplying WR from Table 10-1 with the 1 through 4 rating for each wall. The wall that scores the highest may be developed as the base design and other high scoring walls may be included in the Contract Documents as acceptable alternates.

10.7

WALL SELECTION EXAMPLE

An earth retaining wall is needed to allow construction of a road adjacent to a creek. The road is proposed for construction to provide temporary access to remote areas in a U.S. national forest for about 5 years while a permanent highway is built nearby. Therefore, the earth retaining wall is proposed for temporary support for 5 years with minimum maintenance requirements.


10-14


Table 10-4. Summary of Cut Wall Costs. Cost(1) in $

Wall Type

per ft2 of entire wall(2)

Sheet-pile wall(3, 4)

15 – 40 (3, 5)

Soldier pile and lagging wall

20 – 35(6)

Slurry (diaphragm) wall(3)

70 – 120

(3)

Tangent pile wall

25 – 45

Secant pile wall(3)

30 – 50

(3)

DMM wall

45 – 60 (3)

Jet grouted wall

60 – 90

Soil-nailed wall(7)

25 - 60

Anchors(8)

30

Permanent Concrete Facing

20 – 30

1

Note: Total installed costs in 2007 U.S. dollars; 2 Costs include whole wall include embedded portion, unless otherwise noted. 3 Cost shown do not include permanent facing or anchors; 4 Lower cost is associated with renting sheet piles for temporary applications; 5 Cost shown for walls up to 16 ft; 6 Costs shown are based on wall face area above excavation bottom; 7 Cost shown do not include permanent facing; and 8 Assume average 50 ft long ground anchor “affecting” 75 ft2 of wall.

Table 10-5. Summary of Fill Wall Costs Cost(1) in $

Wall Type

per ft2 of exposed wall face

Concrete gravity wall (10 ft)

25 – 35

Concrete cantilever wall (20 ft)

20 – 35

Concrete counterforted wall

20 – 35

Concrete crib and metal bin walls

25 – 35

Gabion wall

30 – 50

Concrete module wall

30 – 35

MSE wall (precast facing)

20 – 35

MSE wall (cast-in-place facing)

25 – 45

MSE wall (modular block facing)

15 – 25

MSE wall (geotextile/geogrid/welded wire facing)

10 – 25

T-wall

30 – 40

1

Note: Total installed costs in 2007 U.S. dollars. FHWA NHI-07-071 Earth Retaining Structures

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Based on the roadway profile and the site topography, the required wall height along the alignment is estimated to be 30 ft in certain sections. The soils at the site consist of medium dense silty sand, with zones of soft compressible clays that may cause long-term differential settlement problems, even over the relatively short service life of the wall. Wall selection for this example is performed following the wall selection flow chart provided in Figure 10-1. Step 1 and Step 2 are already given in the problem statement. Step 3: Evaluation of Project Requirements against Factors Affecting Wall Selection The owner’s evaluations of the factors that affect the wall selection are presented in Table 10-6. The ground represents a critical factor because the site consists of soils that have a potential to undergo differential settlement. Environmental concerns are rated high because the site is located in a U.S. national forest, which requires that any construction satisfy specific construction permitting requirements (which may involve limitations on noise and restrictions on working in the creek). Durability and maintenance are not considered important due to the temporary nature of the wall. Groundwater is rated as one of the least important factors for the project because no temporary dewatering is needed during construction. Contracting practice and tradition are not important issues overall, however the Owner has requested that a concrete cantilever wall be considered since standard design details are available for this system. Step 4: Selection of Wall Systems For this project both cut and fill wall systems could be constructed, however due to costs and the need for specialized equipment and labor associated with cut wall systems, fill wall systems are chosen as the most viable for this project. Based on an initial review of the project data, concrete walls can be eliminated immediately because of the required wall height, relatively slow construction, and poor performance where the potential for differential settlement exists. Furthermore, MSE walls with modular block or precast concrete facing are also eliminated because the project calls for a temporary wall and other facing types (for MSE walls) are more appropriate for temporary applications. Concrete module wall, gabion wall, bin wall, and MSE wall with a wrap-around geotextile face are chosen to be further evaluated. As presented in Step 3, the owner decided to keep tradition as one of the factors affecting the wall selection. Therefore, to satisfy traditional construction practice in the area, a concrete cantilever wall is also added to the list for further evaluation. FHWA NHI-07-071 Earth Retaining Structures

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The performance of each wall type against the factors previously identified are evaluated and tabulated in Table 10-6. A brief discussion on the rating of some of the wall selection factors is presented below. Ground conditions at the site are known to be problematic due to their potential for differential settlement. Therefore, walls that can typically tolerate higher differential settlement are rated higher and walls that are more rigid and susceptible to damage caused by differential settlement. Construction considerations are evaluated based on the availability of material and the effort it takes to deliver these materials to the site. Walls that require no special backfill material and consist of wall materials that could be delivered to the site fairly easy, such as geotextile, bin, and concrete modules, are rated high. Walls that require special material such as gabions or suitable aggregate for concrete are rated lower than the other wall types. The ROW for the wall is evaluated based on the space required behind the structure. Walls that require less space are rated higher than the walls that require more space behind the structure. The importance of aesthetics for the wall selection is a bit subjective. Walls that have a concrete facing are typically considered more aesthetic however, because the site is located in a U.S. national forest, walls with a more natural appearance, such as gabions and vegetation are rated high. Walls that could be constructed with less potential for encroachment on the creek and less potential for noise are rated high. Concrete walls are known to be more durable and they are rated higher than geotextile walls where the geotextile may degrade due to exposure to sunlight. Step 5: Selection of an Acceptable Wall System The final wall selection matrix is shown in Table 10-7. The scores range from 58 for concrete cantilever wall to a high of 77 for MSE wall with a wrap-around geotextile face. The wall type with the highest score is selected for the project and second and third ranked wall types are considered as alternates.


10-17


4 3 1 4

4

2

1

3

3

Groundwater

Ground

2

1

3


Gabion Wall Concrete Module Wall Concrete Cantilever Wall MSE Wall (Geotextile wrap)

Bin Wall

Wall Types

WR

4

3

3

2

3

1

3

2

3

Speed of Construction


4

3

2 Aesthetics

2

4

4

3

3

3

10-18

2

4

2

3

1

Initial Rating

ROW

2

3

3

3

3

3


3

3

4

4

3

4

4

3

4

3


1


1

4

3

2

2

Tradition


4

1

2

Table 10-6. Initial Rating of Each Wall Selection Factor for Each Wall Alternative.

10.8

HYBRID WALL SYSTEMS

10.8.1 General Occasionally hybrid wall systems are used for an earth retaining system project. Hybrid walls combine elements from typical fill and cut wall systems and are effective alternatives to fill or cut walls when they can be built higher, with less ROW restrictions, and/or less expensively than cut or fill wall systems alone. Hybrid walls always require special attention by the design engineer because: •

•

•

The hybrid wall may combine systems whose components require differing magnitudes of deformation to develop resistance to loading. These differences may lead to incompatible deformations at the wall face; The relatively well-defined measures available to provide internal drainage of fill and cut wall systems may require special detailing for a hybrid wall system to provide drainage continuity; While design information and performance requirements are available for individual components of a given cut or fill wall system, these values may need to be modified to address the performance of the complete hybrid system.

In the remainder of this section, specific issues related to the design of hybrid wall systems are identified and selected examples of hybrid wall systems that have been constructed for highway facilities are provided. It is recognized that other hybrid systems have been constructed in the U.S. 10.8.2 Compatibility of Deformations Systems designed for cut support characteristically require much smaller strains to mobilize the restraint mechanism than systems designed for fill support. Attachment of multiple systems with incompatible strain characteristics to a common wall face can result in overstress of the low strain elements and damaging and/or aesthetically unpleasing differential movement of the wall face.


10-19



1 4

3 9

12

Gabion Wall 3

4

6

Bin Wall

6

3

Ground

Concrete Module Wall Concrete Cantilever Wall MSE Wall (Geotextile wrap)

Groundwater

Wall Types

Construction Considerations 8

6

6

4

8

Speed of Construction 12

4

12

8

12

ROW 8

8

6

6

6

Aesthetics 4

8

4

6

2

Environmental Concerns 10-20

9

9

9

9

9

Durability and Maintenance 6

8

8

6

8

1

4

3

2

2

Tradition

Risk Factors That Affect Wall Selection

4

4

3

4

3

Contracting Practice 12

3

9

6

12

77

58

69

67

71

Total Score

Table 10-7. Wall Selection Matrix and Total Score of Each Wall Alternative

Cost


Cheney (1990) described a failure of a 34-ft high hybrid wall system. The wall was designed with a flexible face of precast reinforced concrete segments interconnected with vertical steel rods. Lateral earth pressures were resisted by a hybrid system consisting of rock anchors in the lower 12 ft, a gravity bin wall in the middle 8 ft and deadman anchors in the upper 14 ft. Following construction, cracks were noted in the lower portion of the wall and the wall face had bulged in the middle of the wall. It was surmised that movements in the upper two thirds of the wall had reduced pressures in that region to active conditions. Because of the relative rigidity of the lower portion of the wall with rock anchors, stresses in that region increased to levels probably consistent with at-rest earth pressures causing the reinforcements to break out in that region. 10.8.3 Overall Stability of Hybrid Systems In many cases, hybrid wall systems may consist of two walls effectively stacked on one another to achieve a given total wall height. Two examples of such systems are shown in Figures 10-2 and 10-3. The wall in Figure 10-2 is a combination of a cast-in-place cantilever wall and MSE wall. MSE walls or cantilever walls may require a slope cut behind the wall to provide for the necessary area to construct the wall from the bottom-up. In locations where ROW restrictions would not otherwise permit the construction of a full-height MSE wall, this hybrid wall system can be built to reduce the ROW requirements of the wall because the length of the footing for the lower wall can be significantly reduced as compared to a case where the full wall height requirement is satisfied with one wall. Typically this type of hybrid wall is built for wall heights greater than 30 ft. This hybrid wall can also be used for road widening applications where a road exists on top of the cantilever wall. The hybrid wall shown in Figure 10-3 is a combination of a MSE wall and cast-in-place Lwall. This type of wall may be used for bridge abutment applications. The MSE wall serves as the foundation for the L-wall. Without the MSE wall, the full height of the wall would be reinforced concrete. A cost savings is therefore achieved because less concrete is required for the wall. In general, the design of the lower wall system for those shown in Figures 10-2 and 10-3, requires consideration of the weight of the upper wall system. Typically, the design of the lower wall system considers the upper wall system as a surcharge loading for stability computations. The design of the upper wall system must consider the potential effects of settlement of the lower wall and backfill because the lower wall provides foundation support for the upper wall.


10-21


REINFORCEMENT

CANTILEVER WALL

Figure 10-2. MSE Wall on Top of Cantilever Wall (after CDOT, 2003).

L - WALL REINFORCEMENT

Figure 10-3. L-Wall on Top of MSE Wall (after CDOT, 2003). In many cases, hybrid systems such as these can be analyzed using slope stability limit equilibrium methods as compared to earth pressure concepts usually used for conventional cut and fill walls. Limit equilibrium analyses allow numerous potential slip surfaces to be analyzed and can allow the various restraining forces available in a hybrid system to be modeled within one analysis. With this, however, the design engineer needs to select an acceptable factor of safety for the hybrid wall system which may be different from those commonly specified for individual fill or cut walls.


10-22


10.8.4 Example Hybrid Systems Gabion Walls Anchored with Geogrids This hybrid wall is a combination of a gabion wall and geogrid reinforcement of a MSE wall (Figure 10-4). In this application, the gabion baskets may be considered as “facing” elements for an MSE wall. Alternatively, stacked stones can also be used as facing elements. In the stacked stone application, stones are grouted together. These types of walls are typically used in mountainous areas where there is an abundant source of stone. These walls are used for projects that have strict ROW requirements (i.e., available base width < 0.5 H) which would preclude the use of gravity or MSE wall systems alone. With geogrid reinforcement, gabion baskets can be placed on top of each other to heights greater than 26 ft. The differential settlement tolerance of this wall is similar to the tolerances for gabion and MSE walls, which is 1/50.

Figure 10-4. Gabion Wall Anchored with Geogrid (after CDOT, 2003).

Anchored L-Wall This hybrid wall is a combination of a modular precast L-wall (or modular bin wall) and geogrid reinforcement of an MSE wall (Figure 10-5). This wall is similar to a MSE wall with precast facing elements. The modular precast L-wall facing may be a proprietary product. The L-wall units are 2 ft tall which results in a larger reinforcement spacing than for a typical MSE wall. Walls with L-wall facings are typically used for permanent applications.


10-23


PRECAST L-WALL

REINFORCEMENT

Figure 10-5. Anchored L-Wall (after CDOT, 2003).

Anchored Cantilever Wall This hybrid wall is a combination of CIP cantilever wall and ground anchors (Figure 10-6). This type of wall is used for rehabilitation and for roadway widening applications. If the existing CIP cantilever wall shows any signs of movement, or cracks, ground anchors may be installed to stabilize the wall. Ground anchors are only lightly post tensioned to prevent damage to the existing face of the CIP cantilever wall. Depending on the site constraints, ground anchors may be installed using an over hanging drill. The post tensioning and load tests are performed from a working platform. Composite T-wall This hybrid wall (Figure 10-7) is a combination of several structural systems with elements of: • • • •

A modular precast facing unit (i.e., T-wall); A horizontal tension anchor such as a rock bolt or a soil screw; A vertical support such as a pile acting in compression or tension; and A structural fill tying all the components together such as concrete or select granular soil.


10-24


GROUND ANCHOR

CANTILEVER WALL

Figure 10-6. Anchored Cantilever Wall (after CDOT, 2003).

Composite T-walls may be considered in the following situations: •

•

•

Where insufficient space is available for a conventional wall, a composite T-wall of narrow cross section can be built with vertical compression piles, vertical tension anchors and narrow facing units that can be filled with concrete. Where a mixed face of soil and rock is encountered, a composite T-wall of narrow cross section can be built in front of the rock with horizontal rock anchors and narrow facing units can be filled with concrete. Above the rock a standard wall of wider cross section can be built. Where existing utilities limit the space behind the wall, a composite T-wall of narrow cross section can be built in front of the utilities. If built on soil, the wall can consists of piles, tension anchors, and concrete filled facing units. If built on rock, the wall can consists of horizontal or vertical rock anchors, and concrete filled facing units.

Combination Temporary Soil Nails and Permanent Anchored Wall For some cut wall applications, construction is performed underneath an existing bridge. Because headroom and equipment access is limited, soil nails can be an effective system to provide support. However, in some states, permanent soil nails may still not be acceptable. In that case, an effective hybrid system consists of installing temporary soil nails in the upper FHWA NHI-07-071 Earth Retaining Structures

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Figure 10-7. Composite T-wall (after the Neel Company, 2004).

portion of the cut wall followed by permanent ground anchors for the lower reaches where access is sufficient. If soldier beams cannot be installed, a horizontal wale beam can be used to support anchor lock-off loads. With this system, the temporary nails are assumed to carry no loads for the permanent system. Soil Nail Wall and MSE Wall A landslide stabilization system using tiered soil nail walls and a MSE wall was used to stabilize an unstable slope and allow for the construction of a roadway widening for a project FHWA NHI-07-071 Earth Retaining Structures

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in Wyoming (Turner and Jensen, 2005). A typical cross section is shown in Figure 10-8. The design includes two soil nail walls and a MSE wall. The lower soil nail wall reinforces the existing embankment and provides foundation support for the MSE wall. The upper soil nail wall provides support for the existing roadway during and after construction of the MSE wall. The design results in the roadway being partly supported by the upper soil nail wall and partly by the MSE wall and with soil nails crossing the failure plane of the existing slide. Soil Nail Wall and CIP Wall with Geofoam Backfill Figure 10-9 shows a hybrid system constructed for an Arizona DOT project in which a geofoam backfill and fascia wall was constructed in front of a permanent soil nail wall. For this project, the existing fill slope was stabilized with soil nailing. The section of the slope below the bottom elevation of the wall fascia was very loose and unstable. An MSE wall was considered to achieve the roadway grades, but it would have overstressed the foundation soils. Rather than improve the foundation soils, the widening was achieved using lightweight geofoam blocks (with an approximate unit weight of 2 pcf).

Excavate and replace with MSE wall

NEW ROADWAY MSE WALL TEMPORARY FACING

SOIL NAIL PERMANENT FACING

SLIP SURFACE Figure 10-8. Soil Nail and MSE Wall (after Turner and Jensen, 2005).


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ROADWAY EMBANKMENT

CIP OR PRECAST FASCIA WALL

SOIL NAIL GEOFOAM BACKFILL Figure 10-9. Soil Nail and Fascia Wall with Geofoam Backfill (after Samtani, personal communication).

Shored Mechanically Stabilized Earth (SMSE) Wall Federal Lands Highway (FLH), a program of the FHWA, is responsible for design and construction of roadways in rugged, mountainous terrain. Where the terrain is steep, retaining walls are frequently required to accommodate widening of existing roads, or construction of new roadways. Recently, FLH investigated the design and performance of a hybrid system that includes a composite MSE and shoring wall, i.e., a SMSE wall. The report titled “Shored Mechanically Stabilized Earth (SMSE) Wall Systems” (Morrison et al, 2006) provides a design procedure for SMSE systems. Figure 10-10 shows a schematic of a SMSE wall. It is specifically noted that actual wall dimensions including length of reinforcement are evaluated on a project-specific basis. As previously described in Chapter 7, a minimum bench width of 0.7 H is required for MSE walls. Also, toe embedment is proportional to the steepness of the slope below the wall toe. In some cases, the excavation requirements for construction of an MSE wall become substantial and unshored excavation for the MSE wall is not practical, particularly if traffic must be maintained during construction of the MSE wall. Shoring walls, often soil nail walls, have been used to stabilize the backslope (or back-cut) for construction of the MSE wall, with the MSE wall being designed and constructed in front of the shoring wall. When a composite MSE and shoring wall system is proposed, the MSE wall component of the system should consider the long-term retaining benefits of the shoring wall, including reduction of lateral loads on the MSE wall mass and contributions to global stability. FHWA NHI-07-071 Earth Retaining Structures

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0.6H

Shoring Wall H MSE Wall Facing

MSE Reinforcement

0.3H Minimum

Figure 10-10. Example SMSE Wall.


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CHAPTER 11 CONTRACTING APPROACHES 11.1

INTRODUCTION

The purpose of this chapter is to describe contracting approaches that are commonly used in developing construction contract documents for retaining structures. Three contracting approaches may be used for retaining structures and are described herein. These include: (1) method approach; (2) performance approach; and (3) contractor design/build approach. The responsibilities of the owner and the contractor with respect to design, construction, and performance of the wall vary for each of these approaches. All contracting approaches should use performance based acceptance criteria for retaining structures. •

Method Approach: Method specifications are used for owner or material-supplier designs. In the contract documents, wall construction materials and the execution of construction are explicitly specified. A variant to this method allows the contractor to select specific elements of the work (e.g., drilling and grouting methods for anchors or soil nails to achieve a specified resistance).

•

Performance Approach: This type of contracting specification also referred to as an end result specification, uses approved or generic wall systems or components. Included in the contract documents are lines and grades, as well as specific geometric, design, and performance criteria. For this approach, the contractor submits projectspecific design calculations and plans for owner review in conjunction with normal working drawing submittals.

•

Contractor Design/Build Approach: This type of contracting approach is similar to the performance approach, except the responsibility for design, construction, and performance of the completed retaining wall is placed solely on the specialty contractor. This method requires a strict prequalification process as part of the selection of the specialty contractor. Performance-based submittals should be provided to the owner at key times during design and construction.

All contracting approaches are valid for most earth retaining systems, if properly implemented. Often the approach will be selected based on the experience of the owner and their engineering consultants with wall systems, the complexity of the project (e.g., coordination of wall project with other contracts, high risk if unsatisfactory wall performance, elaborate design analyses required), the availability of specialty contractors or material suppliers, and the local highway agency philosophy with respect to contracting methods. FHWA NHI-07-071 Earth Retaining Structures

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11 – Contracting Approaches June 2008

In 2007, most State DOTs have a formal policy with respect to selection, design, construction, and contracting of wall systems. The general objectives of such a policy are to: •

obtain local highway agency uniformity in selection of wall systems;

•

establish standard policies and procedures for technical review and acceptance of proprietary and generic earth retaining systems;

•

establish internal agency responsibility for the acceptance of new retaining systems and/or components; and for the design, preparation of bid documents, and construction control and, if necessary, performance monitoring, of such systems;

•

develop uniform design and performance criteria standards and construction and material specifications for earth retaining systems; and

•

establish guidelines for the selection of method, performance, or contractor design/build contracting approaches.

11.2

METHOD CONTRACTING APPROACH

11.2.1 Introduction The method contracting approach involves the development of a detailed set of plans and construction specifications for inclusion in the bidding documents. Depending on the specific wall type, certain components of the work will be the responsibility of the Contractor. The advantage of this approach is that it enables the design engineer to examine various earth retaining system options during design, with impartiality that cannot be expected from the contractor. The design engineer also has more time to optimize the design and develop technical details that would minimize uncertainties and disputes during construction. A disadvantage of the method approach is that for alternate bids, more systems must be evaluated, and more sets of design must be developed. Therefore, the owner’s resources may be expended, even though only one wall system will be constructed. Another disadvantage is that the designer may be unfamiliar with newer and potentially more cost-effective systems and thus may not consider them during the design stage. Similarly, the proprietary wall systems, may have technical details known only to the proprietors, thus, the owner or the consulting engineer may not feel comfortable enough to use them.


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Under this contracting procedure, the agency is fully responsible for the design and performance of the wall system, as long as the contractor has installed each component in strict accordance with the contract documents. The agency assumes all risks and is responsible for directing the work if changes to the design are required based on actual field conditions. The method contracting approach is best suited for walls supporting fill where the available technology is either traditional or widely disseminated and reasonably well-established. Knowledgeable contractors and material suppliers of fill-type wall systems are widespread throughout the United States. Detailed plans and special technical provisions are often furnished to the design engineer, at no expense, by specialty contractors and proprietary material suppliers, especially those involved in construction of MSE wall systems. The use of a variant to this method, in which the contractor is responsible for developing, for example, required anchor or soil nail capacity by varying the drilling and grouting methods, drill hole diameter, and length of anchors or nails from specified minimums, has several advantages. It empowers contractors to maximize the use of their experience and specialized equipment and allows the agency to share the major risk (i.e., pile capacity for a specified length) with the contractor. 11.2.2 Contract Documents for Method Approach The contract documents in the method approach consist of plans, specifications, and bidding items and quantities. In the method approach, the contract can be bid on a lump sum basis or following a detailed unit price list. Bidding documents (i.e., Plans and Specifications) prepared using the method approach should typically include at least the following items: •

•

•

horizontal alignment of the wall identified by stations and offset from the horizontal control line to the face of the wall and all appurtenances that affect construction of the wall; elevation at the top and bottom of the wall, beginning and end stations for wall construction, horizontal and vertical positions at points along the wall, and locations and elevations of the final ground line; cross sections showing limits of construction, existing underground interferences such as utilities or piles supporting adjacent structures, any backfill requirements, excavation limits, as well as mean high water level, design high water level, and drawdown conditions, if applicable;


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•

• • • • •

notes required for construction including general construction procedures and all construction constraints such as staged construction, vertical clearance, right-of-way limits, construction easements, noise and air quality requirements, etc.; typical sections and special details; dimensional and alignment tolerances during construction; all details for connections to traffic barriers, copings, parapets, noise walls, and attached lighting; drainage system details; and payment limits and quantities.

In addition to the general requirements listed above, the following items are required for specific wall types under the method contracting approach: Cast-in-Place Gravity and Semi-Gravity Walls (Chapter 5) • • • • • • •

location of the wall footing; footing size, depth, and reinforcing steel details; leveling and proof-rolling requirements for the foundation area; location and details for deep foundations, if required, including pile elevations and pile cap details; location of construction and expansion joints and applicable details; typical sections of the wall showing concrete dimensions and, for cantilever and counterfort walls, reinforcing steel details; and special technical specifications such as pile driving.

Modular Gravity Walls (Chapter 6) • • • • • • • •

length, size, and type of the gravity unit (concrete module, gabion basket, bin cell, crib cell, etc.), and positions for which unit dimensions change; wall cross-section showing unit arrangements and positions for which different unit sizes are used; footing location, depth, dimensions and details (removing unsuitable materials, proof rolling, leveling, etc.); limits of any required wall excavations; properties and methods of placement of infill and backfill materials; corrosion protection requirements of metallic units; arrangement of wall face and tolerance on alignment; and planting or seeding requirements in the facing blocks.


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Mechanically Stabilized Earth Walls (Chapter 7) • • • • • • • • • • •

length, size, and type of soil reinforcement, and positions for which the reinforcing elements change in length or size; layouts, dimensions and elevations of the footings and/or leveling pads; backfill soil property requirements and, if required, requirements for backfill placed just behind the wall facing; horizontal alignment of wall face and offset from the horizontal control line to the face of wall; alignment and elevation of internal drainage systems, and method of passing reinforcing elements around the drainage systems; construction constraints, such as staged construction, vertical clearance, right-of-way limits, etc.; details of facing panels and panel connections with reinforcing elements; details of wire-mesh reinforcement for shotcrete facing and steel reinforcement for cast-in-place facing; details of architectural treatment or surface finish of the facing; details for construction along curved alignments, and around drainage facilities, overhead sign footings or other structures; and corrosion protection requirements/details for reinforcing elements.

Nongravity Cantilevered Walls (Chapter 8) Sheet Pile Walls • • •

size, location, and minimum embedment depth of sheet piles; corrosion protection, if required, of steel sheeting; and requirements for sheet-pile installation equipment (if necessary based on Owner experience with given ground type).

Soldier Pile and Lagging Walls • • • • • • •

size, location, and minimum embedment depth of soldier piles; soldier pile spacing; requirements for installation methods for piles (i.e., driven or drilled in); excavation requirements between soldier piles; requirements for size of lagging; details for temporary and final facing connections to the soldier piles and/or walers; corrosion protection requirements of soldier piles;


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• •

alignment, installation, and elevation requirements for drainage systems; and in-situ and backfill soil property requirements.

Slurry Walls • • • • • • • • •

dimensions of the slurry wall panel, top and bottom wall elevations, panel joints and depth of embedment below the bottom of the excavation; bentonite slurry and concrete mix and concreting details in slurry walls; panel excavation requirements and disposal of excavated material; details of reinforcing steel, soldier piles, or precast concrete panels in slurry walls; guide wall details, width, and height and requirements for the finished face of the guide wall methods of placing structural steel shapes, reinforcing steel, and concrete; field testing and inspection requirements; requirements for cleaning, patching, and sealing of leaks during construction; and verticality requirements of the slurry wall.

Tangent and Secant Pile Walls • • •

• • • • • •

diameter, tip elevation, reinforcement arrangement and construction sequence; requirements for piling equipment (especially for secant pile walls); excavation requirements (i.e., wet method, dry method, temporary casing method, etc.), details of slurry operations (i.e., methods to mix, circulation, desanding, etc.), inspection and testing responsibilities; center to center pile spacing and minimum overlap requirements (for secant piles); methods of handling and disposal of spoil excavation, waste slurry, and waste concrete; concrete placement and testing requirements; acceptance requirements for constructed piles and correction requirements for piles installed outside these acceptance requirements; field demonstration requirements for trial wall panel; and testing program for pile integrity and acceptance.

Jet Grouted and Deep Mix Walls • • •

jet grouting system, e.g., single, double or triple fluid, etc.; minimum required design parameters, e.g., grout pressure, w/c ratio, strength, permeability, etc. minimum column diameter;


11-6


• • • • • • •

layout of stabilized columns; installation sequence of deep soil mixing columns; vertical and horizontal tolerances for stabilized columns including minimum overlap requirements; field and laboratory testing requirements and frequency; reinforcing details or size and type of soldier piles, if any; special considerations where peat, boulders, utilities, etc. may be encountered; and quality control procedures and field acceptance check list.

Ground Anchors (Chapter 8) • •

location, inclination, minimum unbonded length, and anchor bond length; anchor capacity and lock off load for each anchor;

•

corrosion protection requirements for the anchor head, the unbonded length, and the anchor length; requirements for anchorage devices, drilling, and tendon insertion; requirements for bondbreaker and centralizers; acceptable cement grout types, water for mixing grout and grout tubes; requirements, details, frequency and acceptance criteria of anchor testing; details for facing treatment or permanent facing installation including drainage requirement and water proofing.

• • • • •

Soil Nail Walls (Chapter 9) • • • • • • • • • •

nail size, spacing, length, and inclination; nail installation procedure, including minimum drill hole diameter; requirements for soil nail components (i.e., centralizer, nail grout, etc); requirements for bearing plates, nuts, and shear connectors; required nail pullout resistance; corrosion protection requirements; minimum thickness of the temporary and/or permanent facing with details for facing reinforcement and nail cover plates; details for facing treatment or permanent facing installation; nail testing requirements, including methods and frequency of testing; and sequence and dimensions of the step-by-step top-down construction of the soil-nailed wall.


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11.3

PERFORMANCE CONTRACTING APPROACH

11.3.1 Introduction For the performance contracting approach, the owner establishes the scope of work and prepares drawings showing the geometric requirements of the retaining wall, design loadings and factors of safety, material specifications or components that may be used, performance requirements, and any instrumentation or monitoring requirements. The performance approach offers several benefits over the method approach when used with appropriate specifications and prequalification of suppliers, specialty contractors, and materials. Design of the structure is the responsibility of the contractor and is usually performed by a trained and experienced contractor or engineering consultant. This enables engineering costs and manpower requirements for the owner to be decreased since the owner’s engineer is not preparing a detailed design, and transfers some of the design cost to construction. The disadvantage of the performance approach is that if the owner’s engineer is not experienced with designed wall system technology, he/she may not be fully qualified to review and approve the wall design and any construction modifications. Newer and potentially more cost-effective methods and equipment may be rejected due to the lack of confidence of owner personnel to review and approve these systems. Three principal methods have been used to implement the performance approach for walls. These methods are referred to as pre-bid wall design, pre-bid typical section design, and post-bid design and are described in subsequent sections. Differences between these methods are associated with the required time to perform the design. 11.3.2 Implementing Performance Contracting Approach 11.3.2.1

Pre-bid Wall Design

Contract documents for pre-bid wall designs are prepared to allow for various retaining wall alternates. With this method, the owner contacts specialty contractors and informs them that a retaining wall is being proposed for a site. The owner requests that the contractors prepare detailed wall designs prior to the advertisement of the bid. The designs are based on ownerprovided line and grade information, geotechnical and subsurface information, and design requirements. Approved designs are then included in the bid documents. This approach allows the owner to review design details based on submittals from several contractors. Because of the detail that must be provided with this type of a submission, only those contractors who have significant expertise and experience in proposed retaining wall are FHWA NHI-07-071 Earth Retaining Structures

11-8


likely to prepare the required submission. The owner should prepare and include a generic wall system design in the bid documents to enable general contractors to decide whether they want to use the generic design or a design from a specialty contractor. 11.3.2.2

Pre-bid Typical Section Design

With pre-bid typical section design, schematic or conceptual plans are developed by prequalified specialty contractors based on geometric and performance requirements specified by the owner. Sufficient detail must be provided by the specialty contractor to enable the owner to judge whether the approach of the contractor is acceptable. Contractors will typically exclude details which they believe are unique to their design. The advantage of this approach compared to pre-bid wall design is that specialty contractors are more likely to submit their solutions for review and inclusion in the bid documents. With this approach, only limited preparation effort is required by the contractor, and development of a detailed design and working drawings is only necessary if they are the successful bidder. The disadvantage of this approach is that total project requirements are less well defined and may lead to misunderstandings and claims. In cases where the general contractor will not be constructing the proposed retaining system, the apparent lack of detail using this approach may result in problems during construction because the general contractor does not fully understand the design. 11.3.2.3

Post-bid Wall Design

Like pre-bid wall design and pre-bid typical section design, the post-bid wall design approach allows for various prequalified contractor-designed wall alternates. In the bid documents, each wall and acceptable alternates are identified. Design requirements for each wall type are contained in the special provisions or standard agency specifications. General contractors receive bids from prequalified specialty contractors and subsequently select a specialty contractor-prepared wall design and wall price to include in their bid. Once the contract is awarded, if the general contractor decides to build the wall system, he/she then requests that the selected specialty contractor prepare detailed design calculations and a complete set of working drawings for owner review and approval. Upon approval, the walls are built in accordance with the working drawings. When an owner uses this type of contract, they benefit from the experience of the wall contractors or supplier. However, they do not have as much control over the finished product as they do when they require the prebid approval of the working drawings. Also, since the general contractor wants to minimize risk, he/she will likely not select an alternate design unless the construction cost savings is significant.


11-9


11.3.3 Contract Documents for Performance Approach Regardless of which performance approach is used, the owner must prepare and include as part of the contract documents geometric and site data, design guidelines, and performance requirements. Also, for performance specifications, an instrumentation and monitoring program may be required in which minimum levels of instrumentation to be used by the contractor and threshold values against which the monitoring data will be evaluated are included. Required information for inclusion in the bid package is listed below: 11.3.3.1 •

•

•

• • •

horizontal alignment of the wall identified by stations and offset from the horizontal control line to the face of the wall and all appurtenances that affect construction of the wall; elevation at the top and bottom of the wall, beginning and end stations for wall construction, horizontal and vertical positions at points along the wall, and locations and elevations of the final ground line; cross sections showing limits of construction, any backfill requirements, excavation limits, as well as mean high water level, design high water level, and drawdown conditions, if applicable; all construction constraints such as staged construction limitations, vertical clearance, right-of-way limits, construction easements, etc.; location of utilities, signs, etc., and any loads that may be imposed by these appurtenances; and data obtained as part of a subsurface investigation and geotechnical testing program;

11.3.3.2 •

• • • •

Geometric and Site Data

Design Guidelines

reference to specific governing sections of appropriate agency design manuals (materials, structural, hydraulic, and geotechnical), construction specifications, and special provisions; if none are available, reference to current AASHTO LRFD Specifications may be used; magnitude, location, and direction of external loads due to bridges, overhead signs and lights, and traffic surcharges; limits and requirements of drainage features beneath, behind, above, or through the structure; seismic design requirements; infill and backfill requirements, including material specifications and placement wall facing;


11-10


•

• • •

resistance factors for applicable limit states such as limiting eccentricity (i.e., overturning), sliding, stability of temporary construction slopes, overall rotational stability, bearing resistance, uplift resistance, pull-out resistance, etc.; geotechnical design parameters such as friction angle, cohesion, and unit weight, as well as electrochemical properties of the soils to be utilized; type, size, and architectural treatment of permanent facing; and erosion protection requirements, if applicable, and methods of shore protection design.

11.3.3.3 • • • • • • •

Performance Requirements

design life for the earth retaining system and, if applicable, required corrosion protection; tolerable horizontal and vertical movements of the structure and acceptable methods of measuring these movements; required pile axial resistance and static and dynamic driven pile and drilled shaft testing requirements for walls supported on deep foundations; acceptance criteria, and performance and proof testing requirements for ground anchors , soil nails, and micropiles; anticipated creep behavior of anchors, soil nails, and micropiles, and methods of measuring creep movements; durability requirement of jet grouted or DMM walls, and methods of testing longterm behavior of soilcrete; and permissible range of variation in groundwater levels, and methods of groundwater level measurement.

With respect to nongravity cantilevered walls, anchored walls, and soil nail walls which may be used for temporary and/or permanent support in urban or congested areas, there are other specific project elements that must be considered and addressed in the Contract Documents. These include: • •

•

performance requirements (if Contractor design) for temporary groundwater and surface water controls and dewatering systems; requirements for coordination between excavation contractor and wall contractor including maximum excavation below brace/anchor level and requirements for minimal disturbance near base of excavation; and requirements for protection of adjacent structures (e.g., underpinning, responsibility for preconstruction survey of adjacent structures, requirements for pile driving).


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11.3.4 Review and Approval If a performance contracting approach is used, the contractor’s submittals are reviewed and approved by the agency, or its consultant, before construction can commence. The evaluation by the agency’s structural and geotechnical engineers must be rigorous and must consider, as a minimum, the following items: • •

• • • • • • • •

11.4

conformance to the project line and grade; conformance of the design calculations to the agency standards or special provisions or codes such as the current AASHTO LRFD Specifications with respect to design methods; adequacy of the resistance factors used in the design; corrosion protection details; development of design details at obstructions such as drainage structures or other appurtenances; external and internal drainage features and details; architectural treatment of the wall face; monitoring methods as required by the performance specifications; conformance of the bidding documents to the performance requirements of the system; and adequacy of the QA/QC program developed by the contractor, including monitoring methods and testing details.

CONTRACTOR DESIGN/BUILD APPROACH

For the contracting approaches previously described, the owner and contractor share responsibility in the design and construction of the wall system. With the contractor design/build method, the owner outlines the project requirements, obtains complete subsurface and geotechnical information, and provides construction quality assurance. The design-build contractor is responsible for the complete design, construction, and performance of the wall system. A design/build proposal may be submitted either before the bid advertisement (pre-bid) or after the contract award (post-bid). This method is most often used for securing bids on temporary retaining wall projects. The key elements for a successful contract are communication of basic design concepts to the owner and the joint development of a quality assurance plan prior to construction.


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11.5

RECOMMENDATIONS

Regardless of the contracting approach selected, a performance based approach should be used for the construction details of the retaining walls. Specialty contractors and wall system suppliers have developed various special systems, construction equipment, and construction methods which are appropriate for specific soil/site conditions. It is in the competitive interest of the specialty contractor and wall system suppliers to remain current on latest innovations in the field. Public agencies can, therefore, benefit from these innovations by specifying performance requirements of the wall rather than specific components of the wall system. The owner must specify certain minimum system performance requirements such as corrosion protection and acceptance/rejection criteria for the wall system and its components. It is recommended that construction details should be the sole responsibility of the contractor. For walls requiring specialized equipment, materials, and contractor experience, prequalification of specialty contractors is recommended. Alternatively, where prequalification is not allowed, specifications should require that bidders demonstrate that they meet specific experience requirements. Minimum qualifications should include demonstration of previous successful experience in the installation (and design) of the wall components. Depending on the nature of the work, specific experience of the reviewing engineer and drill operators should also be included. Qualifications requirements information for ground anchors and soil nails is provided in the Specifications provided in GEC No. 4 (anchors) (Sabatini et al, 1999) and GEC No. 7 (soil nails) (Lazarte et al., 2003). In 1994, the Civil Engineering Research Foundation (CERF) established the Highway Innovative Technology Evaluation Center (HITEC). HITEC’s purpose was to accelerate the introduction of technological advances in products, systems, services, materials, and equipment to the highway and bridge markets. The evaluation of new and more costeffective retaining wall systems was performed through HITEC’s nationally-focused, earth retaining system (ERS) group evaluation program. While the HITEC program is still available, FHWA funding of the program has been reduced. Wall system suppliers are encouraged to conduct an independent review of newly developed components and/or systems related to materials, design, construction, performance, and quality assurance. As many public agencies, especially state DOTs, require HITEC evaluations or independent evaluations of wall components or wall systems, suppliers should consider obtaining such reviews as it will be beneficial in securing acceptance of their system.


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11.6

SPECIFICATION RESOURCES

Various wall specifications are available from state DOTs either as part of their standard specifications or from Special Provisions from individual projects. Table 11-1 provides a partial listing of websites and sources to obtain specifications for wall types covered in this manual (at the time of the Manual preparation). This list is by no means exhaustive.


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Table 11-1. Identified Sources for Wall Specifications

Wall Types Cast-in-Place Gravity and Semi-Gravity Walls Crib Walls

Gabion Walls

Specification Sources DOT Agencies

Others

Wisconsin

-

New York Wisconsin California Connecticut Montana Wisconsin

-

-

Bin Walls

Wisconsin

Contech (http://www.contech-cpi.com)

Concrete Module Walls (Doublewal)

Georgia New York

Doublewal Corporation

MSE Walls

Illinois Montana Wisconsin

NHI Course Manual (FHWA-NHI-00-043)

T-Walls

-

Neel Company (www.neelco.com)

Sheet Pile Walls

Illinois Wisconsin California

-

Soldier Pile and Lagging (Post and Panel) Walls

Montana Wisconsin Illinois

Central Artery Project (www.bigdig.com)

Slurry (Diaphragm) Walls

Port Authority of New York and New Jersey

Tangent/Secant Pile Walls

Wisconsin (Marquette Interchange Project)

-

DMM Walls

-

Central Artery Project Central Artery Project

Jet Grouted Wall

-

Nicholson Construction www.nicholson-rodio.com Port Authority of New York and New Jersey

Ground Anchors and Anchored Walls

Illinois Wisconsin Georgia

Soil Nail Walls

-


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GEC No. 4 GEC No. 7 11 – Contracting Approaches June 2008


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CHAPTER 12 REFERENCES

AASHTO (2007). American Association of State Highway and Transportation Officials. LRFD Bridge Design Specifications. Washington, D.C., 4th Edition. AASHTO (2004). American Association of State Highway and Transportation Officials. Deep Foundation Design and Construction Guidelines for AASHTO LRFD. Washington, D.C. AASHTO (2002). American Association of State Highway and Transportation Officials Standard Specifications for Highway Bridges. Washington, D.C., 17th Edition. AASHTO (1997). American Association of State Transportation and Highway Officials. Standard Specifications for Geotextiles – M 288, Standard Specifications for Transportation Materials and Methods of Sampling and Testing. Washington, D.C: 18th Edition. ACI (1998). American Concrete Institute. Code Requirements for Nuclear Safety-Related Concrete Structures (ACI 349-97) and Commentary. Farmington Hills, MI: p.129. ADAMA Engineering Inc. (1998-2000). Mechanically Stabilized Earth Walls: Program MSEW, developed under contract with Federal Highway Administration, www.msew.com, Newark, DE. ASCE (1997). American Society of Civil Engineers. Guidelines of Engineering Practice for Braced and Tied-Back Excavations. Geotechnical Special Publication No. 74. ASCE (1994). American Society of Civil Engineers. Guidelines for Braced and Tie-back Excavations. Allen, T.M. (2005). “Development of Geotechnical Resistance Factors and Downdrag Load Factors for LRFD Foundation Strength Limit State Design”, Federal Highway Administration, Report No. FHWA-NHI-05-052. Ashour, M., Norris, G.M., and Piling, P. (1998). “Lateral Loading of a Pile in Layered Soil Using the Strain Wedge Model.” Journal of Geotechnical and Geoenvironmental Engineering, ASCE, Vol. 124, No. 4, pp. 303-315. FHWA NHI-07-071 Earth Retaining Structures

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12 – References June 2008

Barton, N.R. (1973). “Engineering Geology.” Review of a New Shear Strength Criteria for Rock Joints. Vol. 7, pp. 189-236. Becker, D.E., Crooks, J.H.A., Been, K. and Jefferies, M.G. (1987). “Work as a Criterion for Determining In-Situ and Yield Stresses in Clays.” Canadian Geotechnical Journal Vol. 24, No. 4, pp. 549-564. Bell, F. G. (1992). Engineering in Rock Masses. Boston, MA: Butterworth Heinemann Limited. Bieniawski, Z.T. (1974). “Geomechanics Classification of Rock Masses and its Application in Tunnelling.” Proceedings of the 3rd International Congress on Rock Mechanics, Denver, Vol. 2, No. 2, pp 27-32. Bolton, M.D. (1986). “The Strength and Dilatancy of Sands.” Geotechnique Vol. 36, No. 1, pp. 65-78. Bowles, J. E. (1977). Foundation Analysis and Design. New York, McGraw-Hill Publication. Bowles, J. E. (1988). Foundation Analysis and Design, Fourth Edition. New York, McGrawHill Book Company. Brandl, H. (1992). Retaining Structures for Rock Masses, Engineering in Rock Masses. Boston, MA, Buttewroth-Heinemann Limited, Chapter 26, pp. 530-572. Broms, B.B. (1965). “Design of Laterally Loaded Piles.” Journal of the Soil Mechanics and Foundations Division Vol. 91, No. SM3, Proceedings Paper 4342. ASCE, pp. 79-99. Bruce, D. A. (1994). “Jet grouting - Chapter 8.” Ground control and improvement. New York, NY: John Wiley & Sons, Inc. Byrne, R.J., Cotton, D., Porterfield, J., Wolschlag, C., and Ueblacker, G. (1998). Manual for Design and Construction Monitoring of Soil Nail Walls. Federal Highway Administration, FHWA-SA-96-069. Canadian Geotechnical Society (1992). Canadian Foundation Engineering Manual Third Edition. Richmond, British Columbia: Biotech Publishers Ltd. Caquot, A., and Kerisel, F. (1948). “Tables for the calculation of passive pressures.” Active Pressure and Bearing Capacity of Foundations, Paris, France: Gautheir-Villars.


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Appendix STUDENT EXERCISES AND SOLUTIONS CONTENTS #

Title

S1

Active Earth Pressure Calculation

S2

Lateral Load Distribution Due to Strip Load

S3

Sliding Resistance

S4

MSE Wall Design External Stability

S5

MSE Wall Internal Stability

S6

Corrosion Calculation

S7

Apparent Earth Pressure Diagrams

S8

Single-Tier Anchored Soldier Beam and Lagging Wall

S9

Design of a Two-Tier Anchored Soldier Pile and Lagging Wall

S10

Lateral Wall Movement

S11

Soil Nail Wall Design

S12

Fill Wall Selection

S13

Cut Wall Selection


i

Appendix – Exercises & Solutions June 2008


ii

Appendix – Exercises & Solutions June 2008

STUDENT EXERCISE 1 Active Earth Pressure Calculation

Assume a 20 ft deep excavation is to be supported permanently by the retaining wall as shown on Figure S1-1. Calculate the following: a)

The active earth pressure diagram and total resultant active force considering drained strength parameters for the lower clay layer. b) Same as (a) except consider undrained strength parameters for the lower clay layer.

Manual Reference: Sections 3.2.1, 3.2.3, 3.13 Figures 3-13 Example Problem 1

S1-1

Top

10 ft

Layer 1

γ1 = 121 pcf

(Cohesionless Soil)

10 ft

Layer 2

φ1′ = 33°, c1′ = 0

Middle

φ2′ = 28°, c2′ = 0 Su = 630 psf

γ2 = 120 pcf

(Cohesive soil) Bottom

Figure S1-1. A 20 ft deep excavation through two distinct soil layers.

S1-3

STUDENT EXERCISE 2 Lateral Load Distribution Due to Strip Load

The stratigraphy is the same as for Student Exercise 1. A 5 ft wide strip load of 800 psf is located 3 ft behind the wall as shown in Figure S2-1. Use drained strength parameters (for permanent wall). Calculate the following: a)

The lateral earth pressure due to the strip load at a depth of 2 ft and 20 ft. b) The hydrostatic pressure diagram. Manual Reference: Section 3.2, 3.6, 3.7, 3.8 Figures 3-13, 3-20, 3-24 Example Problem 3

S2-1

3 ft

10 ft

10 ft

Layer 1 Layer 2

800 psf

5 ft

γ1 = 121 pcf γ2 = 120 pcf

φ1′ = 33°, c1′ = 0 φ2′ = 28°, c2′ = 0

1

2

3 Note: Figure is NOT to scale.

Figure S2-1. Excavation through two layers of soil with water table located at 10 ft below the ground surface.

S2-2

STUDENT EXERCISE 3 From Station 5+850 to 6+50 (i.e., L = 200 ft), a fill wall is needed to maintain the grade of the highway. A cantilever reinforced cast-in-place (CIP) wall shown in Figure S3-1 was selected. For this wall, compute the following: Group 1: Evaluate sliding resistance. Use heel width of 8 ft (neglect passive resistance in front of the wall). Group 2: Evaluate sliding resistance by increasing heel width from 8 ft to 10 ft. Group 3: Evaluate sliding resistance by increasing φ’b from 29° to 32°. All Groups: Evaluate bearing resistance based on Group 1’s parameters. Manual Reference: Sections 5.5.3 through 5.5.7 Figures 5-11, 5-12, 5-13

S3 - 1

γconc =145pcf

β = 10°

γb =115pcf ∆h φ’b = 29° (Groups 1 and 2) φ’b = 32° (Group 3)

DC

EV

PV

EH δ = 10°

h = 20 ft

Not to Scale

(DCstem)

1.5 ft

γf =115pcf φ’f = 35°

5 ft

PH

θ = 90° 2.5 ft

(DCftg) 2 ft 2 ft (heel width)

Clean fine to medium sand

B

Figure S3-1: Group 1 (with forces)

S3 - 2

2.5 ft

Summary of Equations Sliding Equations Active Earth Pressure Coefficient: sin 2 (θ + φ ' ) b Ka = Γ sin 2 θ sin(θ − δ ) ⎡ ⎢ Γ = ⎢1 + ⎢ ⎣

(Figure 3-5)

sin(φ ' +δ ) sin(φ ' −β ) ⎤⎥ b b ⎥ sin(θ − δ ) sin(θ + β ) ⎥

2

⎦

β = Angle of backfill with horizontal = δ (Assumed) δ = Orientation of the resultant active force. It is assumed that resultant active force is oriented at the same angle as backfill) φ’b = Effective internal friction angle of backfill θ = Angle of back wall face to horizontal

Active Earth Pressure: EH = FT = 0.5γ b H 2 K a

(Figure 5-13)

γb = total unit weight of backfill H = height of pressure surface

Factored resistance against sliding: RR = ϕτ Rτ + ϕ ep Rep ϕτ = resistance factor for sliding resistance Rτ = Nominal sliding resistance = V tan δ

(Eq 5-7)

(Eq 5-8) V = Total factored vertical force (see Figure 5-11 and Chapter 4 for load factors) tan δ = tan φ ' f (for concrete cast against soil) φ’f = Effective internal friction angle of foundation soil ϕep = Resistance factor for passive resistance Rep = Nominal passive resistance S3 - 3

Bearing Equations Vertical stress calculation: σv =

∑V

B − 2e

(Eq 5-2) B

where: ΣV = Sum of factored vertical loads B = width of wall footing B e = − Xo B 2 Xo =

M

VTOT

−M

HTOT

V

MVTOT = Sum of moments due to vertical forces MHTOT = Sum of moments due to horizontal forces

Bearing Resistance:

(Eq 7-7) q = ϕ qn R b ϕb = Bearing resistance Factor (B’ is based on AASHTO C10.6.3.1.1) qn = 0.5γ B' NγmC wγ f γf = total unit weight of foundation soil B’= B – 2eB Nγm = Nγ sγ iγ (From AASHTO section 10.6.3)

Nγ = Bearing capacity factor ⎛B⎞ Sγ = 1− 0.4⎜⎜ ⎟⎟ = footing shape factor ⎝L⎠ L = length of wall iγ = load inclination factor Cwγ = ground water table correction factor

S3 - 4

STUDENT EXERCISE 4 MSE Wall Design External Stability The ground profile between 6+150 to 6+650 is shown in Figure S4-1. A 23 ft high MSE wall was selected to retain the highway in this stretch. The traffic load is applied parallel to and up to the wall face. The unit weight of the retained fill is 127 pcf, and the frictional strength of this fill and the foundation soil has been estimated at 30 degrees, with no cohesion. Evaluate the following: a) Preliminary length of reinforcement b) Eccentricity (at Strength I limit state) c) Sliding (at Strength I limit state) Manual Reference: Section 7.8.1, 7.8.2 Figure 7-18 Example in Section 7-9

S4- 1

23 ft

Figure S4-1: Proposed highway cut and MSE wall

S4 - 2

STUDENT EXERCISE 5 MSE Wall Internal Stability From Student Exercise 4, it was determined that the 23 ft high wall is externally stable with 16.5 ft long reinforcements. Consider that linear ribbed reinforcements will be used and that the frictional strength of the select fill was determined to be at least 34 degrees and the maximum factor F*=1.5 Compute at a depth of 11.5 ft the following: a)

The length of reinforcement in the resisting zone for internal stability computations. b) The coefficient K for internal stability computations. c) The coefficient F* for internal stability computations. Manual Reference: Section 7.8.3 Example in Section 7-9

S5 - 1

STUDENT EXERCISE 6 Assume a Mechanically Stabilized Earth (MSE) wall is being constructed using metal strip reinforcements and mildly corrosive backfill. Given: Anticipated design life (lifetotal): 75 years Thickness (tn) = 160 mil (0.16 in.) Reinforcement width (b) = 2 inch Galvanization thickness (tg) = 3.4 mil (0.0034 in.) Yield strength of steel (Fy) = 65 ksi Compute: Factored long-term reinforcement design strength of each reinforcement (kips) at the end of the anticipated design life. Manual Reference: Sections 7.8.3 Example from section 7-9

S6 - 1

STUDENT EXERCISE 7 Apparent Earth Pressure Diagrams

Three soil profiles are provided. Develop appropriate apparent earth pressure diagrams for a 20 ft high anchored (or internally braced) wall in these profiles: a) Sand b) Soft Clay c) Stiff to Hard Clay

Manual Reference: Section 8.11.2.1 Figures 8-67, 8-68, 8-69

S7-1

STUDENT EXERCISE 8 Single-Tier Anchored Soldier Beam and Lagging Wall A single-level 25-ft high (H) anchored wall is proposed to be built as shown in Figure S8-1. The anchor will be constructed at a depth of 8.5 ft below the top of the wall (H1) at an inclination of 30° (i) below horizontal. Calculate the following: a)

Anchor force based on apparent earth pressure diagram for sands. b) Size of prestressing steel (assume steel strand will be used for anchor tendon). c) Length of anchor bond zone assuming nominal anchor bond stress (τa) of 4.4 ksf. Assume drill hole diameter (d) of 6 inches.

Manual Reference: Sections 8.11.2 through 8.11.7 Figures 8-67, 8-70, 8-74 Table 8-16

S8-1

Summary of Equations a) Horizontal anchor force (Tn) using Tributary Area Method: Tn

(23H =

)

− 10 HH1 p (spacing ) (Figure 8-74) 54(H − H1 ) H = height of wall H1 = distance from top of wall to anchor p = K aγH (Figure 8-67) φ' ⎞ ⎛ K a = tan 2 ⎜ 45 − ⎟ (Eq 3-2) 2

⎝

2⎠

φ' = effective friction angle

γ = total unit weight of sand

b) Tensile Resistance (Fn): ϕt Fn ≥

γ pTn

cos (i ) ϕ t= Resistance factor for tensile resistance (See Section 8.11.6) γp = Load factor for Apparent Earth Pressure (See Table 4-2) i = inclination of anchor below horizontal

c) Anchor bond length Lb min ≥

T ϕ p Qn

T=

(Eq 8-6)

γ pTn

(Eq 8-4) cos(i ) ϕ p= Resistance factor for pullout (cohesionless soils)

Nominal pullout resistance = Qn = πdτ a (Eq 8-7) d = drill hole diameter τa = nominal anchor bond stress S8-2

H1

Driven soldier pile w/ timber lagging facing

Cohesionless Soil φ=33 φ’ =33°° γm=18 =115kN/m pcf 3

H-H1

H= 25 ft

T

Pile spacing = 2.43 mcenter O.C. 8 ft on

Figure S8-1. Cross Section at Location of SingleLevel Anchored Wall

S8-3

STUDENT EXERCISE 9 Design of a Two-Tier Anchored Soldier Pile and Lagging Wall A 25 ft deep excavation is to be supported by a driven soldier pile and lagging wall and two levels of ground anchors as shown in Figure S9-1. The soils to be retained by the anchored wall are predominantly medium to dense sands to a depth of more than 20 ft below the proposed bottom of the wall. Evaluate the following: a) b) c) d) e) f) g)

Anchor forces and subgrade reaction forces Load factors and resistance factors Tendon type and tensile resistance Anchor bond length Flexural resistance of solider pile Axial loads in wall Design of steel section for solider pile

Manual Reference: Sections 8.11 Figures 8-67, 8-71, 8-73 AASHTO (2007) Section 6.5.4.2, 6.9.2.2 S9-1

Driven soldier pile w/timber lagging facing

H1

q=0.2 γ [[psf ]

H2

T2

Cohesionless Soil φ=33° γm=115 =18 kN/m pcf 3 8 ft O.C. Pile spacing = 2.43 m O.C.

d

H3

H= 25 ft

T1

Figure S9-1. Geometry, soil properties, and loading for driven soldier pile anchored wall design problem.

S9-2

STUDENT EXERCISE 10 Lateral Wall Movement Assume a 42 ft deep excavation through sand and clay is supported by a wall and three levels of ground anchors as shown in the cross-section in Figure S10-1. The width of the excavation is 40 ft and the length is 400 ft. Estimate maximum lateral wall movement for a sheetpile wall and for a structural concrete wall. Calculate lateral wall movement for: Sheetpile wall with Su of soft to medium clay of 600 psf b) Structural concrete wall with Su of soft to medium clay of 600 psf c) Sheetpile wall with Su of soft to medium clay of 950 psf

a)

Manual Reference: Section 8.11.10, 8.11.11 Figures 8-69, 8-76d

S10-1

Sand (SP) N=15-26 φ=33° γ=121 pcf

4.7 ft

9.3 ft

3.9 ft

13.1 ft

Soft to Medium Clay Su=600 - 950 psf φ=28° γ=120 pcf

11 ft

d = 12 ft 18 ft

Stiff Clay Su=1500 - 2300 psf φ=32° γ=125 pcf

Note: Drawing not to scale.

Figure S10-1. Geometry of excavation.

S10-2

Assume the following material properties, PZ-27 Sheeting (Reference Skyline Cut Sheet) Section Modulus – 30.2 in3/ft Moment of Inertia – 184.2 in4/ft Modulus of Steel (E) = 29,000 ksi Geometry of excavation as shown in Figure S10-1. Structural Slurry Wall Thickness = 2.5 ft Steel Reinforcements = #8 & #6 #8 – Vertical Direction (Diameter = 1”) #6 – Horizontal Direction (Diameter = .75”) 6” Spacing Ec = 57,000 f 'c f’c= 3000 psi Ec = 57,000 3000 = 3,100 ksi

S10-3

S10-4

STUDENT EXERCISE 11 Soil Nail Wall Design To maintain the grade of the proposed highway, a cut in the existing hillside between Stations 6+600 and 6+900 was necessary. The profile of the cut is shown in Figure S11-1. A soil nail wall was selected to support the cut slope. Perform a preliminary soil nail wall design using the simplified design chart procedure to estimate nail length and maximum design load.

Manual Reference: Section 9.2.7.3

S11-1

20 ft 10°

Medium Desnse to Dense Silty Sand Soil Parameters: 32 ft 10°

γ = 115 pcf (unit weight) f u= 34° (friction angle) cu = 100 psf (cohesion) Qu = 1200 psf (nail pullout resistance based on field pullout testing)

Assumptions: •Permanent, Non-Critical Structure •Ground water table below bottom of cut •AASHTO Group I loading condition governs •Soil nail inclination angle = 15° (typical) •Initial trial nail spacing: SH = SV = 5 ft (typical)

Figure S11-1. Wall Cross Section.

STUDENT EXERCISE 12 Fill Wall Selection Problem Statement An earth retaining wall is needed to allow construction of a road adjacent to a creek. The road is proposed for construction to provide temporary access to remote areas in a U.S. national forest for about 5 years while a permanent highway is built nearby. Therefore, the earth retaining wall is proposed for temporary support for 5 years with minimum maintenance requirements. Based on the roadway profile and the site topography, the required wall height along the alignment is estimated to be 30 ft in certain sections. The soils at the site consist of medium dense silty sand, with zones of soft compressible clays that may cause long-term differential settlement problems, even over the relatively short service life of the wall. Wall selection for this example is performed following the wall selection flow chart provided in Figure 10-1.

Assignment Select the appropriate wall alternative for construction at the above described embankment in the national forest. Select weighting factors from 1 to 3 for the wall selection factors listed on the selection matrix in Table S12-1. Using Tables 10-3 and your own table that you have created during class, perform initial screening and select potential alternatives. For each wall alternative considered, assign an initial qualitative rating from 1 to 4 based on each wall selection factor. Calculate the weighted ratings by multiplying the initial rating by the weighting factors and summarize the results in Table S12-2. Assign a final score for each wall alternative.

Manual Reference: Chapter 10 Figure 10-1 Tables 10-3 and 10-5 Example Section 10.7

S12 - 1

STUDENT EXERCISE 13 Cut Wall Selection Problem Statement A permanent retaining wall is required for a major interchange in a large Midwestern city. The wall needs to be built rapidly and may need to be constructed during the winter. Existing structures including pile foundations and utilities are in close proximity to the wall line. The wall system, therefore, needs to be relatively watertight to prevent groundwater drawdown and potential resulting settlements of nearby structures. The contract documents will require that the maximum lateral wall movement be less than 1 percent of the wall height. The maximum height of the wall is 33 ft as shown in Figure S13-1, and is approximately 200 ft in length. The general soil profile along the alignment of the wall includes a silty clay and sand fill overlying silty clays that increase in strength with depth. The majority of the wall will be constructed in a medium stiff to stiff silty clay layer with sands and gravel. SPT blowcount values range from 20 to 30 blows per ft in this layer. Prior to wall construction, the upper 10 ft of fill will be excavated to the elevation of the top of the wall.

Fill

Medium Stiff to Stiff Clay

Exposed Face of Future Wall ~33 ft

Very Stiff to Hard Clay

Dense Silty Sand

Figure S13-1. General stratigraphy and wall geometry.

S13-1

Assignment Select the appropriate wall alternative for construction at the above described interchange project using the procedure described in the wall selection flow chart in Figure 9-1. Select weighting factors from 1 to 3 for the wall selection factors listed in the selection matrix. For each wall alternative assign an initial qualitative rating from 1 to 4 based on each wall selection factor. Assign a final score for each wall alternative.

Manual Reference: Chapter 10 Figure 10-1 Tables 10-2 and 10-4

S13-2

FHWA NHI-07-071 Earth Retaining Strucutures June 2008 (Without Solutions)

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